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COLOPHON

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GEOTECHNIEKSpecial 16th European Conference on Soil Mechanics and Geotechnical Engineering.

Geotechniek is the leading independent journal for geotechnical

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CONTENTS

GEOTECHNIEK IS POWERED BY:

A coarse sand barrier as an effective piping measure4 Van Beek, V.M. / Koelewijn, A.R. / Negrinelli, G. / Förster, U.

Interpretation of TA and DSS test on organic soft soil to derive strength parameters for dike design.8Lengkeek, H. J. / Brunetti, M.

Working Through Water16 Corney, N.

Geotechnical foundation design for onshore wind turbines18Cools M.Sc., J.E.

Reinforced soil walls over compressible soils23 Suk, S.

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IntroductionBackward erosion piping is a process whereby shallow pipes are formed at the interface of a co-hesive layer and a sand layer, due to the removal of sand particles under the action of water flow. Ongoing pipe development can lead to severe erosion and finally failure of the water-retaining structure. The foundation that is susceptible to this mechanism, a combination of a uniform sandy layer covered by a cohesive layer is often encountered below river dikes in deltaic areas. Rise of the water level results in the formation of sand boils as a first sign of backward erosion. Numerous sand boils have been observed in the past, but failure due to backward erosion piping is not very common. Nevertheless, several dike failures in the Netherlands, China and the U.S. are attributed to this mechanism (Vrijling et al., 2010, Yao et al., 2009).

In the Netherlands, backward erosion piping is predicted with the Sellmeijer model (Sellmeijer, 1988, Sellmeijer et al. 2011, TAW, 1999). This model predicts the critical head on the basis of the groundwater flow towards the pipe, the viscous flow through the pipe and the limit-state equilibrium of particles at the pipe bottom. The model has been validated using experiments, but application in the field proves to be complex, as the required parameters are difficult to deter-mine and show large fluctuation in the field. The uncertainty with respect to input parameters leads to the selection of conservative estimates, such that considerable reinforcements are due. The more stringent safety standards, due to a recent validation of the model, the inclusion of the length-effect and the risk approach, recently embraced in the Netherlands, lead to a further

increase of the dike length to be reinforced.

Traditional measures, such as a berm, are not attractive when large seepage lengths are re-quired and often houses are situated closely behind the dikes. Sheet pile walls are an alter-native, but are economically unfeasible when it comes to application for long dike stretches.

Innovative or alternative piping measures are therefore becoming more and more popular. An example of an innovative measure is the vertical sand-retaining geotextile (Bezuijen et al., 2013, Förster et al., 2015). Using this method nearby the toe of the dike a vertical geotextile is inser-ted into a trench. Above the sand layer the trench is refilled with clay, such that upward seepage is not possible. An optimisation of this innovative solution is proposed here: a coarse sand barrier. In this solution the pipe formation is resisted by coarse sand instead of by the geotextile.

The conceptThe coarse sand barrier relies on the concept that coarse sand provides more resistance to pipe formation than fine sand. The coarse sand is brought into the subsurface by creating a trench and simultaneous filling with coarse sand. Once a pipe forms, it will develop along the top of the sand bed and will collide with the barrier. As the coarse sand provides more resistance to erosion and the water flow is controlled by the overall properties of the aquifer, the water forces acting on the coarse particles are not sufficient for pipe development, unless the head across the struc-ture is raised.

A coarse sand barrier as an

effective piping measure

Van Beek, V.M.Deltares

Negrinelli, G.University of Brescia / Deltares

Koelewijn, A.R.Deltares

Förster, U.Deltares

Figure 1 - The coarse sand barrier concept.


AbstractRecent safety studies indicate a significant risk of backward erosion piping as a failure mechanism for dikes. Traditional measures, such as berms or sheet pile walls take a lot of space or are expensive, in particular in areas with a high infrastructural density. Consequently, there is an urgent need for alternative piping measures. An innovative solution is the verti-cal sand-retaining geotextile, which has been developed and tested in the

field. An optimisation of this method is a coarse sand barrier: a trench filled with coarse sand, covered by a clay layer. The coarse sand provides more resistance to erosion than the inherent sand of the aquifer with its uncertain composition with respect to d70, so that initiated pipes cannot continue to develop. Laboratory and field scale tests suggest that this is an effective and economically feasible method.

The permeable barrier will deflect excessive vertical seepage below the pipe tip, such that fluidisation of the sand bed below the pipe is less likely than for an impermeable structure like a sheet pile wall.

As in the method with the geotextile, above the sand the trench is filled with clay, to pre-vent upward seepage. Due to the clay filling above the barrier, the method is different from a more common filter, which aims for control-led discharge of water. For the coarse sand bar-rier the discharge is not expected to increase, not more than would be the case for any other barrier-type solution. Clogging is not expected, as long as the barrier is continuously below the water level.

Laboratory evidenceLaboratory experiments have been performed to illustrate the functioning and potential of a coarse sand barrier. A small-scale box (descri-bed in more detail in Van Beek et al., 2011 and Van Beek et al., 2015) with transparent cover

(simulating the dike) was used for the experiments. Two confi-gurations were used, one with an open exit representing a 2D exit with unconstrained flow towards the surface (also des-cribed in Van Beek et al., 2008) and one with a circular exit in the cover, representing a 3D exit with concentrated flow towards a single point (Figure 3).

In both configurations the box was filled with fine sand with a band of (medium) coarse sand and a head difference was applied to the sand until pipe formation occurred. The sand types used in the slope-type configuration were Playground sand and Masonry sand (with a d50 of 0.191 and 0.454 mm respectively) and the sand types used in the hole-type configuration were Baskarp sand and Itterbeck fraction 431 μm sand (with a d50 of 0.132 and 0.342 mm respectively). All sand types are uniform (d60/d10 1.5 – 2.6). The observed process was similar in both con-

figurations: the head drop was increased until a pipe formed. This pipe developed up to the coarse sand and started to develop parallel to the direction of flow and adjacent to the coarse sand barrier. Upon significant increase of head the pipe passes through the coarse sand barrier and develops towards the upstream side.

The critical hydraulic heads obtained in the experiments are indicated in Figure 4 for the samples with a coarse sand barrier and their homogeneous equivalents. It is noted that the upstream filter resistance causes quite some head loss in the slope-type configuration, such that the actual head loss across the sand bed is

Figure 3 - Slope type exit viewed from above (above) and circular exit covered

by a sand boil (below).

Figure 2 - Laboratory set up, showing the small-scale model with circular exit.


smaller. Nevertheless, the difference between the critical heads in the experiments with and without sand barrier is very large, illustrating the potential of the method. Field evidenceAs the critical head of piping is affected by scale, it is relevant to also test the method at larger scale.

In 2009 and 2012 large-scale experiments were performed at the location of the IJkdijk. In these tests an actual dike was built on top of a sand bed placed in a basin. The test dike was 3.5 m high, 15 m long and 15 m wide at its base. It was constructed of compacted clay. The base con-sisted of a 3 m thick sand layer which extended 15 m beyond the test dike both at the upstream side and at the downstream side. Homogene-ous tests without measures were performed in 2009. One of the purposes of the experiments in 2012 was to test piping measures, of which one of them was a coarse sand barrier. This coarse sand barrier was applied as an obstructing bar underneath the dike, 0.5 m wide and 0.5 m deep, at about one quarter of the seepage length from the downstream toe, see Figure 5 and 6. The coarse sand filter has been applied in the same basin where the first IJkdijk piping test had been carried out (Van Beek et al., 2011). In the upper 0.5 m a new, comparable sand was placed, with d50 =0.180 mm, Cu =1.7, d70 =0.207 mm. For the selection of a suitable sand for the filter, three criteria should be met: the filter should have a sufficient permeability, it should retain the finer material of the test sand and it

should be internally stable. Based on filter cri-teria by Terzaghi and given by Giroud (2010) and Burenkova (1993) the coarse sand was selec-ted, with d50=1.331 mm, Cu=1.3, d70=1.52 mm, d10=1.054 mm, d15=1.085 mm, d90=1.79 mm. The first sand boil appeared at a head of 1.60 m. The head was increased until a level of 3.49 m, which equalled the height of the levee and was therefore the maximum head that could be ap-plied in this experiment. In the homogeneous equivalent, performed in 2009, the dike failed at a hydraulic head difference of 2.1-2.3 m. These experiments indicate that at this scale the dike with the coarse sand barrier can at least with-

stand a head that is 1.6 times that of a dike wit-hout the barrier. Discussion and ConclusionsThe laboratory and field experiments illustrate that the application of a coarse sand barrier as a piping measure is promising. In the small-scale experiments an increase by a factor of 3-4 was established. An increase in strength of at least 1.6 is obtained in the field experiments at which failure due to piping did not even occur. Design rules should be based on filter criteria, heave criteria and the horizontal resistance against piping, which still requires investigation. Due to the experience with the vertical geotextile as a

Figure 5 - IJkdijk test schematisation showing the location of the coarse sand barrier.

Figure 4 - Critical gradients obtained in the experiments.


piping measure, which is in many ways similar, practical issues for application in the field are likely to be resolvable.

References- Bezuijen, A., Van Beek, V., Förster, U., (2013).

Geotextiel als pipingremmend scherm, hoe werkt het? Geotechniek 18(1): 38-41, katern Geokunst.

- Burenkova, V.V. (1993). Assessment of suffu-sion in non-cohesive and graded soils, Filters in geotechnical and hydraulic engineering, Brauns, Heibaum& Schuler, Balkema, Rotter-dam, 357-360.

- Förster, U., Bezuijen, A, Van den Berg, S. G., (2015), : Vertically inserted geotextile used for strengthening levees against internal erosion

Category: B3 Earthworks, Dams and Dykes- Giroud, J. P. (2010). Development of criteria for

geotextile and granular filters, Proceedings 9th International conference on Geosynthetics, Guaruja, Brazil, 20 pp.

- Sellmeijer, J.B. (1988). On the mechanism of piping under impervious structures. Doctoral dissertation, TU Delft, The Netherlands.

- Sellmeijer J.B., Lopéz de la Cruz J., Van Beek V.M., Knoeff J.G. (2011). Fine-tuning of the piping model through small-scale, medium-scale and IJkdijk experiments. European Jour-nal of Environmental and Civil Engineering 15(8): 1139-1154.

- TAW (Technische Adviescommissie voor de Waterkeringen) (1999). Technisch rapport Zandmeevoerende wellen. Technische Advies-commissie voor de Waterkeringen, Delft, The Netherlands.

- Van Beek, V.M., Koelewijn, A., Kruse, G., Sell-meijer, H., Barends, F. (2008). Piping pheno-mena in heterogeneous sands – experiments and simulations, Proceedings of the 4th Inter-national Conference on Scour and Erosion, p. 453-459, http://scour-and-erosion.baw.de/conferences/icse4/.

- Van Beek, V.M., Knoeff, J.G., Sellmeijer, J.B.

(2011). Observations on the process of back-ward piping by underseepage in cohesionless soils in small-, medium- and full-scale expe-riments. European Journal of Environmental and Civil Engineering 15(8): 1115-1137.

- Van Beek, V.M., Van Essen, H.M., Vandenboer, K., Bezuijen, A. (2015). Developments in mo-delling of backward erosion piping. Géotechni-que, to be published.

- Vrijling, J.K., Kok, M., Calle, E.O.F., Epema, W.G., Van der Meer, M.T., Van den Berg, P., Schweckendiek, T. (2010). Piping - Realiteit of Rekenfout? Technical report, Dutch Expertise Network on Flood Protection (ENW).

- Yao, Q., Xie, J., Sun, D., Zhao, J. (2009). Data collection of dike breach cases of China. Si-no-Dutch Cooperation Project Report. China Institue of Water Resources and Hydropower Research.

A COARSE SAND BARRIER AS AN EFFECTIVE PIPING MEASURE

Figure 6 - Digging the trench for the barrier (left) and the coarse sand barrier (right). At the background the dike that is to be placed is drawn at the side of the basin.


Interpretation of TA and DSS test on organic soft soil

to derive strength parameters for

dike design.

H. J. (Arny) LengkeekWitteveen + Bos, Deventer,

The Netherlands

M. (Matteo) BrunettiWitteveen + Bos, Deventer,

The Netherlands

1. IntroductionFor a study into undrained stability calculations of flood defences on behalf of STOWA and water-board of Delfland, Witteveen+Bos investigated the differences between drained and undrained stability calculations for a local flood defence.

The reference site is a local flood defence along the Berkelsche Zweth channel located between Rotterdam and Delft in the west of the Nether-lands. The Berkelsche Zweth channel functions as a discharge for the water pumped out of the surrounding polders. The studied section is the flood defence between the channel (with a wa-ter level of -0.4 m NAP) and the low lying polder Schieveen (level approximately -5.0 m NAP).

The analyzed flood defence has been previously studied in 1972-1973 in the context of the syste-matic local flood defence study (“systematisch kade-onderzoek COW”). Based on those results it has been improved in the 1980s. In the design of that improvement, particular attention was paid to limiting the “driving” force (e.g. limiting the toe ditch) and limiting the surface subsi-dence. In 2006 another improvement of the flood defence has resulted in its current stable state. From archives it is known that in 1806 a slope instability occurred at about 600m from the pilot location. Further collapse have been prevented by strengthening the dike with a berm.

2. Site investigation program and soil profileAn extensive geotechnical investigation program has been carried out to gain reliable data on soil profile and parameters. In a single cross-secti-on, eight CPT’s and three boreholes with conti-nuous sampling have been carried out (see Fi-gure 1). The CPT’s were made to a level of -19 m NAP, while the boreholes reached a depth of 5 m in the crest of the dike and 10 m in the berm and in the polder. After the field work the following

Figure 1 - Test locations in the topview of the dike.

Figure 2- Soil profile, water level in channel and polder level.


This paper presents the findings on the behaviour of organic soft soil (clay and peat) using a variety of state-of-the-art laboratory tests. Moreover it addres-ses the importance of critical evaluation of the each test and the parameter determination. The tests are performed within a research program on behalf of research foundation “STOWA” to verify the stability of a local flood defence of the waterboard “Hoogheemraadschap van Delfland”. Oedometric tests (OED) and constant rate of strain tests (CRS) have been carried out to deter-mine the preconsolidation stress, after which anisotropically K0-consolidated undrained triaxial tests (TA-ACU) and direct simple shear tests (DSS) have been executed with three different consolidation paths. All tests have been performed on undisturbed samples to obtain reliable soil parameters.

The interpretation of laboratory tests is made in terms of pre-consolidation stress (s’vy), undrained shear strength (Su) and effective friction angle (φ’). This paper focuses on determination of the strength parameters at large strains, the so called critical state or ultimate state parameters, further on called ul-timate state parameters (“ult”). The ultimate parameters can be used for a effective stress approach with Mohr-Coulomb strength model using effective friction angle and no cohesion. The ultimate parameters can also be used for a critical state soil mechanics (CSSM) approach with undrained shear strength related to the effective stress and overconsolidationratio (OCR).

The undrained shear strength (Su) is normalised by the vertical consolidation stress (s’vc) to derive the undrained shear strength ratio (S). By plotting S ver-sus the OCR, both the normally consolidated value (S;nc) as the exponent (m) to describe the overconsolidated behaviour are derived. The OCR is defined as s’vy divided by s’vc. This paper presents additional criteria to define the ul-timate state of Su.

To obtain the ultimate state effective friction angle (φ’), the interpretation of TA-ACU is made by plotting t against s’. For the DSS tests, Su is plotted against the ultimate vertical effective stress (σ’ult), which represents the vertical ef-fective stress acting on the sample at the moment of failure. This paper pre-sents additional criteria to define the ultimate state of φ’.

Lastly a comparison is made between two consolidation procedures to derive the undrained shear strength at in-situ stress conditions. Two identical sam-ples have been tested, one sample is consolidated to the in-situ stress level before undrained shearing. The other samples is first consolidated to the pre-consolidation stress, then consolidated to the in-situ stress before undrained shearing, the so called SHANSEP method.

Abstract

laboratory tests have been performed:• Classification tests (24x)• Conventional Oedometric tests (10xOED)• Constant Rate of Strain tests (11xCRS)• Anisotropically K0-Consolidated Undrained

triaxial tests (18xTA-ACU)• Direct Simple Shear tests (18xDSS) The subsoil conditions on the site are charac-terized by layering of natural and manmade cohesive deposits. The top layers in the dike body consist of anthropogenic clay. Three an-thropogenic layers can be distinguished, which are named Clay Dike, Clay #4 and Clay 1972. The last two layers are not included in this study since it was shown that the failure circles were only just intersecting these layers and that their influence on the overall stability is therefore ne-gligible. Below the manmade soil two natural Holocene deposits are encountered, the Holland peat underlain by the Tiel clay. Below the com-pressible and organic Holocene deposits a thick layer of sand is present.

An overview of the index properties of the stu-died layers is given in Figure 2. In the table it can be observed that within the geological deposits the density (ρ) and the water content (w) vary widely. There is a clear correlation in the vari-ability, lower density correlates to higher wa-ter contents (and higher organic content). Even though the variability within each of the layers is significant, there is still a very clear separation between the different layers.

Tabel 1 - Index properties of soil layers

description boreholelevel

(NAP)*γn

(kN/m3)γdr

(kN/m3)w

(%)ρ (kN/

m3)e

(-)Sr

(%)

Clay Dike B001

-0.1 16.2 11.0 48 25.8 1.34 91.1

-0.4 16.7 11.4 47 25.9 1.27 92.3

-3.6 17.1 11.5 48 25.9 1.24 99.9

Holland Peat

B002

-5.4 10.7 2.0 437 17.2 7.62 98.5

-5.8 13.3 5.3 149 22.7 3.25 103.9

-5.9 10.4 1.8 485 16.6 8.33 96.7

B003

-4.7 10.9 3.2 238 19.0 4.93 92.1

-5.3 13.3 5.4 144 22.7 3.18 102.9

-5.5 10.5 2.0 433 17.0 7.61 96.7

Tiel Clay

B002

-7.0 16.5 10.4 59 25.7 1.48 102.0

-9.9 15.4 8.6 79 25.1 1.92 103.2

-10.3 15.0 8.4 77 25.0 1.96 98.4

B003

-6.2 16.5 10.6 55 25.8 1.42 100.2

-7.3 16.3 10.1 61 25.6 1.54 101.6

-9.8 15.8 9.2 72 25.4 1.76 103.6

-10.7 15.0 8.3 80 25.0 1.99 100.3

-12.7 17.2 11.8 46 25.9 1.20 99.7

* γn is the unit of weight of the material at the natural saturation degree found in the field


3. Methodology of geotechnical testingIn general the methodology of the tests has been to determine the state of the soil by using CRS tests after which both the ultimate state strength parameters of the different layers could be determined by using TA and DSS tests.

3.1. CRS testingEleven CRS tests were performed to gain reli-able pre-consolidation stresses. The resulting compression parameters were not evaluated in this studies since no settlement calculati-ons were required. Traditionally, compression

parameters and pre-consolidation stresses are determined using oedometer tests (OED). In oe-dometer tests, each load step is placed instanta-neously. In order to improve the resolution of the data obtained, constant rate of strain tests are performed. Using this combination of tests, more reliable pre-consolidation stress and com-pressibility parameters are obtained. From the CRS test compression parameters and pre-con-solidation stresses were obtained. The best es-timate pre-consolidation stress (s’vy;A) is derived from normal Casagrande procedures. The upper bound pre-consolidation stress (s’vy;B) obtained where the compression curve starts to be linear at higher stresses showing true normally con-

solidated behaviour. This value is typically twice the value of s’vy;A. Using the results of the CRS tests and the calculated in-situ effective stress, the consolidation stresses for the laboratory testing have been determined. The in-situ stres-ses including the pre-consolidation stresses determined by the CRS tests have been graphi-cally presented in Figure 4. In this figure also the consolidation stresses used for TA and DSS tests have been indicated.

The initial specimen height was about 20 mm with a diameter of 66 mm, the rate of strain during the tests varied between 0.2 and 0.5 % / hour. The specimen is placed in a stainless steel ring resting on a ceramic porous stone. The specimen is loaded by means of a piston with a gear driven load frame in order to guarantee the constant rate of strain. During the test the excess pore pressure and chamber pressure are measured by means of transducers. The imposed displacement and the vertical load are also measured. Within the project the following loading schedule was followed:- Start at a low stress up to 4 kPa- Increase the load to a maximum of about 300

kPa (up to 10 times the in situ effective stress)- Series of unload, reload and relaxation.

3.2. Triaxial testAnisotropically consolidated undrained triaxial tests (ACU-TA) have been performed to deter-mine the ultimate state strength parameters for drained and undrained calculations. A number of depths were selected for testing. For each se-lected depth three different consolidation pro-cedures (levels) were used on three separate samples: • Test-I: Consolidation to the theoretical vertical

effective in-situ stress.• Test-II: Consolidation to the best estimate pre-

consolidation stress (s’vy;A) obtained from CRS and then brought back to the theoretical verti-cal in-situ stress (SHANSEP).

• Test-III: Consolidation to at least the upper bound pre-consolidation stress (s’vy;B) ob-tained from CRS in order to ensure it behaves as normally consolidated.

Six samples were tested from the Dike clay and twelve from the Tiel clay. Using the in-situ stress, pre-consolidation stress and the test methodo-logies as described before, the resulting conso-lidation stresses for the different samples are shown in Table 2.

No TA were performed on the layer of Holland peat, since TA on peat do not give an accurate

Figure 4 - In-situ effective stress and (pre-)consolidation stresses.

Figure 3 - Testing methodology (test types and resulting data).


results of the strength parameters.

The initial specimen height was about 66 mm with a diameter of 33 mm, the rate of strain during the tests varied between 0.2 and 0.5 % / hour. The tests were performed to 15% axial strain (equals 22% shear strain). In line with the research program all tests were K0-consolida-ted, where K0 has been taken as 0.5. An OCR de-pendent K0 has not been applied.

3.3. DSS testDSS with constant height (constant volume) during shearing have been performed to deter-mine the ultimate state strength parameters for drained and undrained calculations. Three sam-ples were tested of the Dike clay, Nine of the Holland peat and six of the Tiel clay. The direct simple shear tests (DSS) are consolidated in the same way as the TA-ACU, with one slight chan-

ge. In test-III the samples are first consolidated to 10% higher stress level before consolidating to s’vy;B. This way the excess pore pressures before shearing are minimized. Furthermore the tendency to develop large creep strains compensated by vertical stress relief (due to constant height test condition) is reduced. The consolidation pressures as shown in Table 3 have been used. The DSS on de Dike clay sho-wed some unrealistic behaviour and will not be discussed in this paper.

The initial height of the sample was about 27.5 mm with a diameter of 65.9 mm. The rate of strain during the tests was between 1.2 and 1.8 mm/hour. The tests were performed to 40% shear strain.

3.4. DSS simulation and interpretationDuring the DSS test only the vertical effective

stress and the shear stress can be measured. This leads to the situation where the full stress condition of the sample is unknown (the direc-tions of the principal stresses cannot be deter-mined). In order to be able to determine a reaso-nable value for the critical state parameters, some assumptions on the behaviour of the ma-terial have to be made. When back-analyzing the DSS test in Plaxis, it can be shown that the prin-cipal stresses are under an angle of 45o with the vertical (Ref. [4]). This leads to the situation in which σ’v = σ’h = s’, and Su = τ = t. When plotting Mohr-circles using this assumption, it is clear that the mobilized friction angle can be found by using sin φ’ = τ / σv’ instead of using tan φ’ = τ / σv’. This assumption is not necessarily conser-vative, but especially when Plaxis is also used in the design for a situation where the material is loaded in shear, it is considered a valid and consistent approach since the back-analysis has already shown that Plaxis will in this situation represent the actual behaviour of the material correctly.

4. Results4.1. Stress strain curvesThe stress strain curves of Holland peat and Tiel clay are presented in are presented in Figure 5 and 6. The graph of Dike clay is not added as it shows similar behaviour. The stress is norma-lised by the vertical consolidation stress. Vari-ous observations are made:• Test III with the normally consolidated sam-

ples results in the lowest stress ratio (S). Test I and II result in higher S-values due to the overconsolidation. This is in line with the criti-cal state soil mechanics (CSSM) consideration.

• Large strains are generally required to reach ultimate state, most samples show ongoing hardening.

• In particular the DSS on Holland peat (Figure 6) show large strains to reach ultimate state, even for test III.

• In particular the TA test III on Tiel clay (Figure 5) show small strains to reach peak strength and some softening at ultimate state.

In general it can be concluded that the tests results are considered to be normal and in line with experience with organic soft soils.

4.2. Stress pathSome stress path’s require careful evaluation. Two examples are presented in Figure 7 and 8. The tests are selected to address the importan-ce of critical evaluation of the strength parame-ter. Normally the ultimate state is defined as the large strain value. Here this leads to unrealistic

INTERPRETATION OF TA AND DSS TEST ON ORGANIC SOFT SOIL TO DERIVE STRENGTH PARAMETERS FOR DIKE DESIGN.

Tabel 2 - TA-ACU consolidation pressures used

soil type borehole sample testsample

level(m NAP)

s’vc (kPa)

s’hc (kPa)

Clay Dike B001

3

I -0.78 20 10

II -0.88 38 → 20 19 →10

III -1.36 98 49

8

I -3.26 32 16

II -2.38 76 → 32 38 → 16

III -3.98 234 117

Tiel clay

B002

12

I -6.38 22 11

II -6.51 44 → 22 22 → 11

III -6.61 128 64

20

I -9.6 42 21

II -9.41 84 → 42 42 → 21

III -9.51 216 108

B003

5

I -5.76 20 10

II -5.89 36 → 20 18 → 10

III -6.29 98 49

8

I -7.11 18 9

II -6.61 46 → 20 18 → 10

III -6.71 108 54


high values for Su, in particular for overcon-solidated samples (Test I and II). These sam-ples tend dilate strongly with increasing strain.

Therefore two additional criteria are defined:• The Su;ult value is defined as the maximum se-

cant value in the stress path graph (green line

in Figure 7 and 8). This is the maximum t/s’ ra-tio in TA and the maximum τ/s’v in DSS. Beyond this value the sample tend to dilate strongly where the normal stress increase significant more than the shear stress.

• The Su;ult value is defined as the first inter-section with the tensile cut-off line (Figure not added). The tensile cut-off is the state in which the σ’3 is lower than 0, i.e. that tensile stresses start to develop. In theory soil does not provide any tensile resistance , this implies that the minimum allowable effective stress (σ’3) can never be negative. Assuming that σ’3 = 0 it geometrically implies that the left limit is identified by a 1:1 line in the s’ - t plane. For DSS it is assumed the same 1:1 line can be plotted in the τ / σ’v plane.

The minimum of the ultimate state and the two additional criteria is now defined as the new condition and compared to the ultimate state only. The new condition is briefly indicated with “min” compared to “ult” for the ultimate state only.

4.3. CSSM S-OCR curvesThe undrained shear strength can be evaluated is in line with the critical state soil mechanics (CSSM) approach, where the undrained shear strength ratio (S) is plotted versus the overcon-solidationratio (OCR) and equations 1 and 2 ap-ply to derive S and mSu;nc = S∙s’v (eq. 1)Su;oc = Su;nc∙OCRm (eq. 2)

The results of both conditions are presented in Figure 9 and 10. Various observations are made:• The S-value of dike clay is considered to be

very high, in particular for a low organic mate-rial. This is not fully understood. A few aspects might be of influence. The layer is manmade

Tabel 3 - Consolidation pressures for DSS tests

soil type borehole sample testlevel

(m NAP)

s’vc;1 → s ‘vc;2(kPa)

Clay Dike B001 2

I -0.51 10

II -0.45 19.3 → 10

III -0.38 84.3 → 80

Holland peat

B002

7

I -4.14 18.5

II -4.24 35.2 → 19

III -4.28 91.5 → 86

9

I -5.03 19.3

II -4.94 31.7 → 18

III -4.90 118.5 → 112

B003 3

I -4.35 9.9

II -4.40 26.5 → 10

III -4.36 93 → 88

B002 21

I -9.78 45

II -9.81 66.9 → 43

Tiel clay III -9.98 209 → 196

B003 9

I -8.74 20.3

II -8.63 37.3 → 20

III -8.60 108.6 → 102

Figure 6 - Stress strain curves DSS on Holland peat.Figure 5 - Stress strain curves TA-ACU on Tiel clay.

0

0,1

0,2

0,3

0,4

0,5

0,6

0,7

0,8

0,9

1

0 5 10 15 20 25 30 35 40

S =

tau

/ σ

'vc (

-)

Shear strain (%)

DSS Holland peat

B3M3 test I

B2M9 test I

B2M7 test I

B3M3 test II

B2M9 test II

B2M7 test II

B3M3 test III

B2M9 test III

B2M7 test III0

0,1

0,2

0,3

0,4

0,5

0,6

0,7

0,8

0,9

1

1,1

0 5 10 15 20

S =

t /

σ'v

c (

-)

Axial strain (%)

TA-ACU Tiel clay

B3M5 test I

B3M8 test I

B2M12 test I

B2M20 test I

B3M5 test II

B3M8 test II

B2M12 test II

B2M20 test II

B3M5 test III

B3M8 test III

B2M12 test III

B2M20 test III

and is mainly unsaturated under in-situ con-ditions.

• The S-value of Tiel clay is 0.33 for TA conditi-ons and 0.27 for DSS conditions. The DSS value is about 80% of the TA value which is in line with reported experiences of other clays.

• The new condition yields generally in a better regression coefficient and less extreme va-lues. In particular the unrealistic high values are taken out. This causes in general a lower exponent (m). The intersection at the OCR=1 axis is not significantly influenced and is in all cases similar the average of only test III (va-lues between brackets in the legend).

• The determination of the exponent (m) de-pends very much on a few outliers. It is conclu-ded that these should be carefully examined

and the new conditions is an improvement. It is furthermore required to have sufficient tests and also larger ranges of overconsolidation ratio.

4.4. SHANSEPThe effect of SHANSEP can be investigated by comparing test I and II. Test-I is consolidated to the theoretical vertical effective in-situ stress. Test-II is consolidated to the pre-consolidation stress and then brought back to the theoretical vertical in-situ stress (SHANSEP procedure). Va-rious observations are made from Figure 11:• The Su value of test II is higher than test I for

all three soil types and both for TA and DSS tests. Most point are above the 1:1 line and be-low the 2:1 line.

• Figure 11 is based on the new condition. It should be noted that the scatter and ratio for the ultimate state condition would be larger.

• Test II yields in a higher exponent (m) in the CSSM model.

• Applying the SHANSEP method typically pro-vides 5 kPa higher Su values, say 20 kPa (Test II) instead of 15 kPa (Test I) at in situ stress conditions. This increase can be significant for stability calculations.

The reason of this difference should be further investigated. It might be that the actual OCR was overestimated. Another reason might be the OCR is a result from aging, and not from physical pre-loading. It should be further investigated if the effect of aging on the Su (or on the exponent


Figure 8 - Stress path example DSS on Holland peat (B003 M3 test-I).

Figure 10 - S-OCR curves from DSS.

Figure 7 - Stress path example TA-ACU on Dike clay (B001 M3 test-I).

Figure 9 - S-OCR curves from TA.



m) is less prone than for physical pre-loading.

4.5. Effective friction angleThe effective friction angle can be derived from

the failure points at the ultimate of new condi-tion. In line with the CSSM approach the cohe-sion intercept is set to zero and equations 3 and 4 apply.

sin ϕ’ = t’ult/s’ult (eq.3)sin ϕ’ = τult/s’v;ult (eq.4)

The TA results are plotted in a t-s’ diagram (see Figure 12). The DSS results are plotted in a τ-s’v diagram (see Figure 13). It should be noted that for the DSS the ultimate vertical stress (at fai-lure instead of consolidation) is used. The fol-lowing observations are made:• The ultimate condition and the new condition

yield in almost identical results. Apparently the stress path moves along the failure line.

• The best fit lines show a small cohesion inter-cept, likely to be caused by the overconsolida-ted test I and II.

• The TA show very little scatter and the effective friction angle (φ’) can be derived accurately.

• The DSS show more scatter, also for the Tiel clay at normally consolidated stress level (Test III).

• The derived ϕ’ for each layer is presented in the legend between brackets. The friction an-gle is derived by averaging each individual test III result per layer.

• The ϕ’ derived from TA on Dike clay is consi-dered to be high, as also concluded for the S-value.

• The ϕ’ derived from TA on Tiel clay is conside-red to be high, but more often seen for clays with organic content.

• The ϕ’ derived from DSS on Tiel clay yields to the same value as that from TA.

• The ϕ’ derived from DSS on Holland peat is high, as expected.

4.6. Resulting parametersThe results of the tests show high friction an-gles for both TA and DSS testing. In numerous previous researches high friction angles for or-

Tabel 4 - Overview of strength parameters

Condition Ultimate state Minimum of requirements

Layer Test S (-) m (-) t’/s’ or τ/s’v (-) φ’ (°) S (-) m (-) t’/s’ or

τ/s’v (-) φ’ (°)

Dike clay TA-ACU 0.51 0.86 0.59 36.2 0.48 0.63 0.62 38.3

Tiel clay TA-ACU 0.33 0.87 0.59 36.2 0.33 0.84 0.59 36.2

Tiel clay DSS 0.27 1.00 0.60 36.9 0.27 1.00 0.60 36.9

Holland peat DSS 0.50 0.55 0.72 46.1 0.50 0.64 0.72 46.1

With:S = Su ratio normalised by the consolidation stressm = OCR exponentt’/s’ = Stress ratio at failure of Test III samples in TAτ/s’v= Stress ratio at failure of Test III samples in DSSϕ’ = Effective friction angle.

Figure 12 - TA t-s’ diagram.

Figure 11 - Comparison Su results test I and II.

Figure 13 - DSS τ-s’v diagram.

ganic soils have been found in triaxial testing. This is especially the case for fibrous soils (peat) [8, 9], the fiber reinforcing effect has therefore been a main explanation for the high friction angles [10]. However, since similar behaviour is found in organic clays, which do not contain high amounts of fibres, it has to be concluded that at this moment the cause of this behaviour is not fully understood.

For direct simple shear testing high friction ang-les have also been reported earlier (for example [5, 11]), and recent studies on Dutch peats have shown that in the τ/σv’ plane in most cases the 1:1 line is approached [6, 7]. This leads to high friction angles for the DSS test as well, similar to the findings in this study.

The undrained shear strengths found from DSS tests is approximately 80 of the undrained shear strengths from TA-ACU. The resulting internal friction angles from the DSS tests are higher than the values found from TA-ACU. Both re-sults obtained are in accordance with literature [1; 2; 3]

The resulting parameters are presented in Table 4. It is concluded that the minimum of require-ments is an improvement for the determination of the undrained shear strength parameters (S, m). For the effective friction angle there is no difference.

5. ConclusionsSeveral laboratory tests have been performed to gain reliable soil profile and parameters in order to check the stability of an existing flood defen-ce. Table 4 provides an overview of the strength parameters obtained from the laboratory tests and used in performing stability calculations.

Based on the analyzed case study it is concluded that:

For the TA-ACU tests:• TA-ACU tests have been performed only in the

layers of clay because they are considered not reliable when performed on peat.

• The resulting Su ratio of Dike clay is about 0.50 with a power of 0.6-0.8.

• The resulting Su ratio of Tiel clay is about 0.33 with a power of 0.6-0.9.

• The resulting internal friction angles of Dike clay is about 37°.

• The resulting internal friction angles of Tiel clay is about 37°.

For the DSS tests:

• DSS tests interpreted in terms of undrained parameters provide comparable results with triaxial tests.

• The resulting Su ratio of Tiel clay is about 0.27 with a power of 1.0.

• The resulting Su ratio of Holland peat is about 0.50 with a power of 0.6.

- The resulting internal friction angles of Tiel clay is about 37°.

- The resulting internal friction angles of Hol-land peat is about 46°.

DSS tests interpreted in term of undrained Su ratio provide lower values when compared with results from triaxial test, while DSS tests inter-pretation in term of drained strength parame-ters provide similar results. These results are in accordance with literature [1; 2; 3].

In general strength parameters evidenced in this case-study are significantly higher when compared with usual and expected cohesive soil strength parameters. A good explanation of this behaviour can be related to the high organic and water content. In organic material the fibres present within the soil are able to provide high tensile strength, thus increasing the soil shear strength.

This paper presents two additional criteria to define the ultimate state, indicated with “min” compared to “ult” for the ultimate state only. It is concluded that a careful evaluation of the tests and the additional requirement are an improve-ment for the Su parameter determination. It is furthermore advised to have sufficient tests and also larger ranges of overconsolidation ratio. The determination of the effective friction angle is not effected or improved.

The comparison between two consolidation pro-cedures showed that the so called SHANSEP method yields in higher Su values, which can be significant for stability calculations. It is recom-mended to investigate this into more detail.

Finally authors would like thank Jan Tigchelaar of Hoogheemraadschap van Delfland and Henk van Hemert of STOWA for their cooperation in this research project.

References[1] Tigchelaar, J., De Feijter, J.W., Den Haan,

E.J. “Shear tests on reconstituted Oost-vaardersplassen clay”. Soft Ground Tech-nology: pp. 67-81, 2001.

[2] Hansen, L.A., and Clough, G.W. “Charac-terization of the Undrained Anisotropy of

Clays”, Application of Plasticity and Gene-ralized Stress-Strain in Geotechnical En-gineering, Ed. by R.N. Yong and E.T. Selig, ASCE, New York, pp.253-276, 1980.

[3] Jamiolkowski, M., Laad, C.C., Germaine, J.T., Lancellotta, R. “New Developments in field and laboratory testing of soils”. Pro-ceedings of the 11th International confe-rence on Soil Mechanics and Foundation Engineering, San Francisco, Aug. 12-16, 1985 Vol. A.A. Balkema, Boston, 1985.

[4] Farrell, E., Jonker, S., Knibbeler, A., and Brinkgreve, R. “The use of direct simple shear test for the design of a motorway on peat”. In 12th European Conference on Soil Mechanics and Geotechnical Engineering, Rotterdam. A.A. Balkema. 1999.

[5] Yamaguchi, H., Yamaguchi, K., and Kawa-no, K., “Simple shear properties of peat.” Proc., Int. Symp. on Geotechnical Engi-neering of Soft Soils, M. J. Mendoza and L. Montañez, eds., Sociedad Mexicana de Mecánica de Suelos, Coyoacán, Mexico, 1, 163–170, 1987.

[6] Den Haan, E.J. “Ongedraineerde sterkte van slappe Nederlandse grond, Deel 2.” Geotechniek, 1:42–51. www.vakbladgeo-techniek, 2011.

[7] Den Haan, E.J. “Modelling peat with an ani-sotropic time-dependent model for clay”, Numerical Methods in Geotechnical Engi-neering – Hicks, Brinkgreve & Rohe (Eds), Taylor & Francis Group, London, 978-1-138-00146-6, 2014

[8] Landva, A.O., “Characterization of Escu-minac peat and construction on peatland”, Characterisation and Engineering Proper-ties of Natural Soils – Tan, Phoon, Hight & Leroueil (eds), Taylor & Francis Group, London, ISBN 978-0-415-42691-6, 2007

[9] Mesri, G., Ajlouni, A.M., “Engineering Properties of Fibrous Peats”, Journal of Geotechnical and Geoenvironmental Engi-neering, Vol. 133, No. 7, 1090-0241/2007/7-850–866, 2007.

[10] Cola, S., Cortellazzo, G.,”The Shear Strength Behaviour of Two Peaty Soils”, Geotechnical and Geological Engineering, 23: 679-695, 10.1007/s10706-004-9223-9, 2005.

[11] Lengkeek, H., Bouw, R., “Triaxial, DSS, CRS tests and numerical simulations of soft soils at river dike”, Proceedings of the 15th European Conference on Soil Mecha-nics and Geotechnical Engineering, Anag-nostopoulos, A., Pachakis, M., Tsatsanifos, C.(eds), IOS Press, 10.3233/978-1-60750-801-4-427, 2011




Niall Corney Huesker Synthetic

GMBH (area manager)

This highway project required the construction of 2km of new build road, 5km of existing road upgrading to highway standard and a 0.8km long embankment to carry the highway across an existing reservoir for approximately 380m. The completion of this section of highway had been politically complex since the proposed road em-bankment location corresponded with the inter-section of three of the Federal Highways Autho-rity regions.

In addition to this It was considered that there was no economically feasible way to cross the

400m of open water as the recent sediments ac-cumulated in this part of the reservoir were up to 20m in depth and were classified as ‘very’ to ‘extremely soft’. These sediments were located under approximately 7m of water which in itself added to the complexity of the construction task.The project was tendered in June 2011 and sub-sequently awarded in September 2011 to main contractor STY Insaat with ATLASYOL Inc. ap-pointed as the specialist sub-contractor res-ponsible for the general earthworks, Geotextile Encased Columns (GECs) and basal platform construction.

The detailed design called for Huesker Ring-trac® GEC sleeves of 80cm diameter with a spacing density between 17 to 20%. Additionally three layers of high strength polyester Stabilen-ka® reinforcement were laid above the heads of the GECs as a basal platform structure. Over 10,000 GECs were installed and in excess of 80,000m2 of Stabilenka® reinforcement.

The key embankment statistics are as follows:Maximum height: 18mBase width: 90mLength: 380mLake water depth: up to 7mSediment depths: up to 20m

This embankment construction involved GECs being installed from land and advancing out-wards with the subsequent sequential installa-tion of the base reinforcement and finally the embankment fill in a progressive manner across the reservoir. A 300T crane with a boom capable of reaching some 75m was utilized to install the GEC piles and the high strength base reinforce-ment geotextile, all underwater.

It should be noted that this is the first time that GEC columns have been installed under stan-ding water. The unique installation procedure and standard equipment adaptations which were made by the contractor, ATLASYOL, have been a fusion of their previous GEC experience, common sense and the ability of the site based engineer team to design and manufacture uni-que apparatus and equipment all of which facili-tated the successful construction.

Of particular interest in this project was that the GEC heads terminated just 50cm above the bed of the lake, up to 7m below the water surface. This meant that the subsequent laying of the ba-sal reinforcement for the basal platform, located immediately above the GEC heads, had to be su-pervised by divers in a very turbid environment with limited visibility.

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Geotechnical foundation design for onshore

wind turbines

J.E. (Jurgen) Cools M.Sc. Geotechnical specialist at

Royal HaskoningDHV, the Netherlands

IntroductionIn the Netherlands, as well as internationally, many onshore large scale wind farms are being constructed or developed (Figure 1). For these wind farms the machine size and rated power continuously (Figure 2). The consequence of this development is the increase of vertical and shear loads at the tower base, along with signi-ficant overturning moments. Large foundations are needed to resist these dynamic loads. The design of the large foundations is complex com-pared to building foundations. Besides the sig-nificant loads, the design includes the selection out of numerous codes and standards, founda-tion methods, calculation methods and design models, ranging from basic to advanced. The design should also meet some very strict design criteria. For this reason the understanding of the geotechnical investigations and design are becoming more and more important to achieve a safe and economical foundation design. To provide a better understanding, some conside-rations regarding geotechnical design and soil investigation are described in this article.

Foundation typesWind turbines can be founded on a variety of foundation types. Figure 3 shows various foun-dation methods for onshore wind turbines. The selection of the best applicable foundation type depends mainly on the geotechnical conditions. The subgrade strength and stiffness of the soil or rock need to be sufficient to resist the cyclic and dynamic wind loads.

A spread footing, or gravity foundation, can be considered as the simplest foundation type. The spread footing is placed directly on the foundati-on soil or rock. The most efficient form is a circu-lar footing with a tapered cone. For construction of the cone a maximum angle of approximately 12° is often used. The weight of the concrete and optional overburden provides resistance against the overturning moments. At locations were strong bedrock is encountered near the surface, post-tensioned rock anchors

can be applied to reduce the dimensions of the cap foundation significantly. The rock anchors must be designed for fatigue.

In regions where competent soil or rock is found at shallow depth, the overlying weak or com-pressible soil can be improved. Many techniques for improvement are available and depend on the type and thickness of the soil to be improved. If the thickness is small the soil can be excava-ted and recompacted or replaced. For greater thicknesses the soil can be improved without (e.g. vibro-compaction) or with admixtures (e.g. soil mixing) and rigid inclusions (e.g. short piles). In all cases the strength and stiffness of the soil mass is improved. The foundation slab is not connected to the inclusions and should be designed as a spread foundation.

Pile and cap foundations are used in regions where the competent soil or rock is encountered at much greater depth. This foundation method is most commonly used in the Netherlands. Fi-gure 4 shows an example of the construction of a pile foundation. The overturning forces on the cap foundation are being transferred to the piles as compres-sive and tensile axial loads. The piles transmit these loads to the ground via a combination of friction and end bearing. Lateral loads are resis-ted through lateral earth pressures on the piles.

Battered piles are often required to increase the lateral foundation stiffness and to increase the pile bearing capacity.

A few decades ago wind turbines with hub heights of 20 to 30 meters were often founded on a small diameter monopile (open ended steel pile with approx. 4m diameter). Due to transpor-tation limitations of the steel piles and the in-creasing overturning forces, the monopole has long been considered as inapplicable. However, currently new techniques are available to install very large diameter (segmental) steel mono-piles (see [REF.2]) or to construct continuous bored pile circular walls that can be considered as an alternative foundation method to the more common piled foundations. An advantage of the-se more ‘innovative’ foundation methods is the limiting foundation footprint.

Design guidelines, codes and standardsThe most commonly used design codes and gui-delines for wind turbine foundation design are: • IEC-61400-1 Wind turbines - Design require-

ments [REF.3] • DNV/Risø Guidelines for Design of Wind Turbi-

nes [REF.4]• GL Guideline for the Certification of Wind Tur-

bines [REF.5]• Eurocode 7: Geotechnical design of structures

- part 1: general rules [REF.6]

Figure 1 - Construction of a large scale wind farm with 40 wind turbines in South Africa .


AbstractThe onshore wind industry continues to grow rapidly with the construction and development of many large scale wind farms using large megawatt wind turbines. This development requires a clear understanding of the geotechnical foundation behaviour in order to achieve a safe and economi-cal foundation design. The foundation design however includes the selec-tion of various types of foundations, design methodologies and mathema-

tical models, ranging from basic to advanced. The geotechnical designer needs to make a choice based on numerous codes and standards, design requirements and the anticipated geotechnical soil conditions. In some cases, standards are conflicting, or just lacking specific design guidelines for wind turbines. Based on extensive design experience some design con-siderations are described in this article.

Besides these guidelines local codes and an-nexes (e.g. the Dutch NEN-EN) are also appli-cable. These national (building) codes are more general in nature. Requirements for the foun-dation design are often specified by turbine ma-nufacturers in technical documents. Because of the differences between standards, guidelines and specifications, it is important that the desig-ner is aware of any conflicts or omissions. The most important standards are briefly described below.

NEN-EN-IEC The wind turbine design in Europe shall meet the requirements contained in the Safety Standard ‘IEC 61400-1, Ed. 3’. In the Netherlands, these standards have been incorporated into the NEN-EN-IEC 61400-series. In this standard all De-sign Load Cases (DLC) for wind turbine design are specified. The IEC standard deviates from the Eurocode NEN-EN1990 (Basis of structural design). Two major differences concern the load factors and the design life time:• In the NEN-EN-IEC61400-1 load factors are

given for wind classes IEC 1, 2 or 3 to derive design values for the wind loads. According to this standard a value of 1.35 for normal loads and 1.10 for abnormal loads. The Eurocode NEN-EN-1990 distinguishes permanent and transient destabilizing loads, with a value of 1.5 for wind loads.

• According to the NEN-EN-1990 for buildings a design life time of 50 years shall be used. The NEN-EN-IEC61400-1 specifies a design life time of 20 years for wind turbines.

These differences indicate that wind turbines should be regarded as structures, other than buildings, for which in the Netherlands speci-fic NEN-EN-IEC standards should be applied. Based on these standards a level of structural safety can be achieved, as required by the Buil-ding Act.The IEC standard is lacking specific guideli-nes for geotechnical investigation and founda-tion design, therefore additional geotechnical standards and guidelines should be applied.

Eurocode 7 (NEN-EN9997-1 including national Annex) The Eurocode 7 does not cover the specific foun-dation design of wind turbines. More general, this standard gives recommendations for the scope of the geotechnical site investigation. It also prescribes a limit state design method. The limit state design implies the application of partial factors load and material parameters or resistance. With regard to the partial factors for foundation design no distinction is made between the different consequence classes. For structures this distinction is commonly made according to Eurocode NEN-EN 1990. However, since the load factors for wind turbines are de-termined according to the IEC, a distinction in consequence classes is not relevant.

DNV GuidelinesThe Det Norske Veritas (DNV) and Risø Natio-nal Laboratory have jointly drafted guidelines for the design of wind turbines. With regard to the DLC’s the DNV guidelines are consistent with the IEC standard. One section is related to the foundation design. In this section recom-

Figure 2 - Growth in size of typical commercial wind turbines [REF.1].

Figure 3 - Foundation methods for onshore wind turbines. The grey subgrade represents competent soil or rock, the brown subgrade represents weak soil.


mendations are given for ground investigation. The guidelines also provide design methods for a spread foundation and a foundation on piles. The proposed design methods for spread foun-dations are comparable with the Eurocode 7. Regarding the design of pile supported foundati-ons the calculation methods comply with the API standard, and not with the Eurocode 7.

GL GuidelinesGermanischer Lloyd (GL) together with DNV has drafted guidelines and technical specifica-tion for the certification of wind turbines. In the ‘Guideline for the Certification of Wind Turbines’ [REF.5] requirements are given for the scope of the geotechnical site investigation. It is stated that the investigation program should comply with at least Geotechnical Category 2 according to the German DIN (DIN 4020:2010-12 and DIN-EN 1997-2). This geotechnical category is con-sistent with Geotechnical Category 2 according to part 2 of Eurocode 7 [REF.7]. The guideline also describes the methodology to be used for the foundation design. The prescribed method is consistent with part 1 of the Eurocode 7.

Limit statesIn accordance with the Eurocode 7 and the GL guidelines the design shall be checked for the ultimate limit state (ULS) and the service limit states (SLS). The limit states that must be chec-ked are listed in Table 1.

Geotechnical design criteriaSome of the most important design criteria that are specified in codes, guidelines and technical documents are briefly described below.

Rotational foundation stiffnessFor both spread foundations, as well as piled foundations, one of the main design criteria for foundation design is the rotational stiffness. In order to avoid excessive motion at the tower top and to provide the required damping, the tur-bine manufacturer always provides a minimum rotational stiffness value. The final foundation design must satisfy this minimum value. Typical minimum values of the rotational stiffness are 60 to 120 GN-m/rad.

The rotational stiffness depends on the stiffness

of the foundation structure and the subsoil. In case of a spread foundation often an infinitely rigid foundation block is assumed, so that the rotational stiffness only depends on the dynamic shear modulus G of the subsoil, and hence of the dynamic modulus of elasticity Edyn and the Poisson’s ratio v. The DNV guidelines provide a clear overview of formulas for the rotational stif-fness for various subsoil conditions and founda-tion methods. For pile foundations, the rotational stiffness de-pends very much on the cyclical spring stiffness of the piles. The cyclical spring stiffness of the piles can be determined according to the empi-rical method based on pile load tests (as descri-bed in [REF.8] and [REF.9]). A typical value for the minimum required cyclical spring stiffness is about 200 MN/m.

GappingThe GL guidelines describe ground gap limita-tions for the foundation design. The ground gap criterion requires that under specific IEC nor-mal operational design load cases, no ground gap (i.e. zero contact pressure) shall occur at the foundation-soil contact. This means that in these cases the entire foundation footprint must remain in compression. This gapping li-mit acknowledges that in the case of a spread foundation the rotational stiffness decreases non-linearly after foundation uplift (zero contact pressure). Besides this, in certain soil conditi-ons, a limit on gapping will also ensure that soils subject to cyclic degradation are prevented from experiencing multiple instances of zero pressu-re which, in the presence of water, could lead to breakdown of the in-situ soil structure and sub-sequent related serviceability problems [REF.8]. For unfactored loads (SLS) under specific ope-rational conditions it has to be proven that the eccentricity (e) of the total vertical load is less than 25% of the radius: e < R/4.

Commonly, for unfactored (SLS) extreme loads, no more than 50% of the base area may be

Table 1 - Limit states.

Spread foundation Foundation on piles

ULS SLS ULS SLS

Overturning (on rock)Overall stability (slopes)Rotational shear failureBearing resistanceSliding resistanceBuoyancy

Ground gapSettlementHeave (swelling, frost)TiltFoundation stiffness

Overall stabilityCompressive resistanceUplift or tensile resistanceLateral resistance

SettlementHeaveTiltRotational stiffnessLateral stiffness

Figure 4 - Example of the construction of a pile foundation for Wind Farm De Zuidlob in The Netherlands

Figure 5 - Relation between overturning moment and the rated power.


GEOTECHNICAL FOUNDATION DESIGN FOR ONSHORE WIND TURBINES

without compression. It has to be proven that: e < 0.59*R.

For spread foundations on competent rock with a good rock mass quality gapping can be allo-wed. The high stiffness and strength allow for some decrease due to the prolonged load alte-rations.

Horizontal stiffnessFor a pile foundation a minimum horizontal stif-fness is required by the turbine manufacturer. Often the minimum value depends on the total mass of the wind turbine and its foundation, or on the rotational stiffness. Typical minimum va-lues of the horizontal stiffness vary between 500 and 1000 MN/m.

A common procedure to calculate the lateral load resistance is by means of the horizontal subgrade reaction for the different soil layers below foundation level. The subgrade reaction can be determined according to Menard’s me-thod, based on the actual cone resistance and soil pressure. For this analysis two dimensional calculation models (e.g. DSheetPiling – single pile) can be used, in which the stiffness of a sin-gle vertical pile is calculated. More advanced programs with three dimensional models (e.g. Plaxis3D or Ensoft Group) can be used to take into account the effect of battered piles. It should be taken into consideration that the la-teral soil reaction should be reduced due to the cyclic loading. This reduction can be up to 10% for sand and 30% for clay (according to DNV gui-delines).

Overall stabilityFor spread foundations it must be verified that the overall stability is sufficient. This is parti-

cularly relevant for footings placed on or near sloping ground. For spread foundations on rock the overturning stability should be checked for the extreme loads in the ULS. According the GL guidelines and Eurocode 7 the safety against is guaranteed by verification of the bearing capa-city of the subsoil. Bearing capacityGenerally spread footings are first proportioned for bearing capacity (and overturning stability in case of rock). The bearing capacity of the ground can be calculated from the general bearing ca-pacity equations as provided in the Eurocode 7. The DNV guidelines provide additional equations for extreme eccentric load cases. In these equa-tions equations for bearing capacity and load inclination factors it is important to apply nega-tive values for the shear forces due to change in direction of the rupture.

TiltThe foundation shall be designed to minimize settlements and especially differential settle-ments. Deformation criteria are often specified by the turbine manufacturer. Generally for wind turbines the following must be kept strictly:• A maximum inclination of 3mm/m resulting

from the characteristic extreme load;• A maximum unequal settlement of the founda-

tion of 1mm/m from the characteristic opera-tional load.

In calculation of the settlements it is important to differentiate static loads, cyclic wind loads and dynamic loads.

Settlement is not governing for design in case of spread foundations on dense soil or good rock, since (average) contact pressures from vertical loads are typically quite low (e.g. 50 to 75 kPa).

Foundation loadsWind turbineGenerally wind turbine foundations are subjec-ted to vertical and shear forces along with signi-ficant overturning moments. These overturning moments are principal for foundation design. Figure 5 depicts the relation between the hub height and overturning moment at the tower base, based on technical data from various wind turbines. From the figure is can be seen that there is wide range of moments for a certain hub height due a large variation in weight of the nacelle and design wind load. Figure 6 shows the relation between the overturning moment and the rated power. Design loads of the tower are provided by spe-cialists of the wind turbine manufacturer in a Load Document (or technical specifications) that satisfies the load cases as outlined in the IEC 61400 standards. All design load combinations as indicated by the IEC 61400 standards are ana-lysed. A distinction is made between operational loads, extreme normal loads, extreme abnormal loads and fatigue. The operational loads are cy-clic loads such as low wind speeds of 5m/s or starting and stopping of the wind turbine gene-rator. The extreme loads have a low probability of occurrence, but result in high design forces at the tower base. Examples of extreme winds are 1-in-50 year 3 second gust or turbine emergen-cy stops. The extreme loads can be considered as dynamic loads. It depends on the turbine type and site specific wind conditions which DLC is governing for foundation design.

Seismic loadingThe effect of earthquake loads on the wind turbi-ne is analysed by the turbine supplier in a num-ber of prescribed load cases. In the Technical

Figure 5 - Relation between overturning moment and the rated power. Figure 6 - Relation between the overturning moment and the hub height.


Documents the assumed seismic ground acce-lerations are specified. As a foundation designer one needs to check whether the site-specific ground accelerations do not exceed the assu-med values. In case of exceeding new calcula-tions should be performed by the manufacturer. The seismic load is normally combined with operational loads and not with extreme loads. The resulting foundation loads are in many ca-ses lower than under extreme loads without ea-rthquake. However, it should be checked in the foundation design that, for example due to lique-faction, the bearing capacity is not exceeded.

BuoyancyThe foundation design shall take into account the vertical hydrostatic pressure (buoyancy). Important is how to deal with buoyancy. The up-ward water pressure can be regarded as a nega-tive stabilizing load or as a positive destabilizing load. The designer should be aware of the effect of buoyancy for various limit states.

Ground investigationBased on the level of complexity of design the Geotechnical Category 2 (GC2) is selected for spread foundations according to Eurocode 7. For pile foundations GC3 is selected due to the cyclic and dynamic pile loading. Eurocode 7 and the DNV guidelines provide recommendations for the scope of the ground investigation. Turbine manufacturers often provide additional specifi-cations for ground investigation.

Generally the soil investigations consist of the following parts: • Geological desk study• Geotechnical investigations• Geophysical survey (optional)

A geological desk study should be performed first to establish a basis for selection of methods and extent of the site investigation. A geotechnical investigation may consist of trial pits, borings, in-situ testing, soil sampling and laboratory testing. For pile foundations in

the Netherlands at least three CPTs should be performed for each wind turbine location, com-bined with one borehole in the centre point of the structure.

For spread foundations on soil or rock it is suf-ficient to perform at least two borings or CPT’s for each location, of which one is performed in the centre point of the wind turbine. The depth to be covered by the investigation is at least 5m in case of unweathered rock. For foundations on soil the depth to be covered should be at least equal to 1 to 1.5 times the largest base dimen-sion of the footing. A typical investigation depth is 20 to 30 m.

Geophysical survey can be used to extend the localized information from borings or CPTs. The results give a better understanding of the stratigraphy within the considered area. Figure 7 shows the result of a geophysical survey for wind turbines on karstic bedrock. The spatial va-riation of the top level of the rock could not have been investigated with only borings and CPTs. Geophysical survey (Ground Penetrating Radar) has also proven its success for wind farms in Finland to investigate the level of the bedrock and thickness of the moraine cover.

ConclusionsDue to the continuous developments in the wind industry it is expected that loads imposed on wind turbine foundations will also become lar-ger. This requires a clear understanding of the driving considerations for geotechnical design. Due consideration should be given to the follo-wing matters: • The extent of the geotechnical investigation

should be based on a geological desk study and comply with the Eurocode 7 and the GL guidelines. For difficult soil or rock conditions geophysical survey can is preferred to extend the localized information from borings and CPTs.

• Based on the encountered ground conditi-ons the most appropriate foundation method

should be selected, that meets the stringent design criteria. Principal design criteria for design of spread foundations are ground gap-ping and the maximum permissible settlement and tilt. For pile foundations the principal de-sign criterion often is the rotational stiffness.

• Wind turbines should be regarded as structu-res, other than buildings, for which in specific IEC 61400 standards exist. These standards are however lacking specific design guidelines for wind turbine foundations. The DNV gui-delines explicitly address the design of wind turbine foundations. For spread foundations these guidelines are well applicable, especi-ally for calculation of bearing capacity under extreme eccentricity. For pile foundations the design methods provided in the Eurocode 7 are preferred. The methods in the DNV for pile foundations comply with the API standards.

References[1] Wiser, R., et al (2011), Wind Energy. In

IPCC Special Report on Renewable Ener-gy Sources and Climate Change Mitiga-tion. Cambridge University Press, Cam-bridge, United Kingdom and New York, NY, USA.

[2] Middendrop, P. and Dorp, R. van (2012), Van Mini- tot Giga-palen. Geotechniek, Funderingsdag 2012 Special. J.16, no.5: pages 30-33.

[3] International Electrotechnical Commis-sion (2005), IEC-61400-1 Wind turbines – Part 1: Design requirements. Third edi-tion.

[4] DNV/Risø (2002), Guidelines for Design of Wind Turbines.

[5] GL Renewables Certification (2010), Gui-deline for the Certification of Wind Turbi-nes. Hamburg, Germany.

[6] Eurocode 7: Geotechnical design of structures - part 1: general rules

[7] Eurocode 7: Geotechnical design of structures – part 2: Ground investigation and testing

[8] Morgan K. and Ntambakwa, E. (2008), Wind Turbine Foundation Behaviour and Design Considerations. AWEA Windpo-wer Conference, Houston.

[9] ‘Fundamenteel’ in Cement 1995/9, pages 17-19

[10] ‘Fundamenteel’ in Cement 1995/5, pages 74-76

Figure 7 - Seismic survey (electrical resistivity) along 3 wind turbine locations down to 10 m depth for wind farm Cerfontaine in Belgium. The blue and green areas indicate the soft soil cover, which is overlying a karstic limestone (indicated by orange and red color).

Website: http://www.windfarmbop.com/wind-turbines-foundation-design/


Sander SukTerre Armee B.V.

Terre Armée structures can be built on compressi-ble subsoil. The settlements have to be accounted for in the design.Conventional Reinforced Earth walls with precast concrete facing panels can be used in areas where the anticipated differential settlements are within the tolerable limit for the precast panel system. The limit depends of the width of the joints, but ty-pically less than 1%, or 1 metre of differential set-tlement along a 100 metre wall length. At locations where greater differential settlements are expected, for example a two stage construction or an obstacle in the subsoil, special provisions like a vertical joint can be applied.

In many projects limitations to the acceptable differential settlements are set by project re-quirements, like aesthetics or a superstructure. Most often, common geotechnical solutions can be applied: preloading, soil improvement, light backfilling materials or consolidation methods. Preloading can be done before, during or after construction of the Terre Armée walls. Over compressible or poor quality soils where pre-

loading or soil improvement cannot be used, piles combined with basal reinforcement or rigid inclu-sions like Menard’s Controlled Module Columns (CMC) can be used to control both stability and settlement of the structure. Menard’s Controlled Module Columns (CMC) is a technique where grout is installed in the foundation soil to both den-sify the soil and to provide additional support. On numerous projects over the world Terre Armée structures have been built on piles with basal rein-forcement or rigid inclusions. Based on theoretical models additional soil reinforcing strips were ad-ded to account for potential additional loading due to stress concentrations caused by arching around the top of the inclusions.In 2012, in order to understand more clearly and demonstrate the behaviour of the combined sys-tem during and after construction, extended theo-retical studies and a full scale instrumentation was performed on a structure, constructed along the Garden State Parkway in Bass River (USA). Data was recorded over 5 months during construction and 2 months after construction.

Figure 3 shows the instrumentation. Stress gau-ges to measure the stress in the reinforcing strips; gauges to measure soil pressures and settlements at the bottom of the reinforced soil block and gau-ges to measure stresses in the rigid inclusions.

Based on the analysis of the recorded instrumen-tation data, it was shown that the strains within the reinforcing strips did not exceed in the very lowest

levels of the MSE wall mass the conventional de-sign values. As a result, there is no need for incre-asing the density of soil reinforcing strips in the lo-wer levels of reinforcement. The location of the top of the rigid inclusions is an important aspect of the design to prevent the soil reinforcement within the Reinforced Earth being overloaded. If the distance is too small, the stresses in the reinforcing strips may increase due to differential movement around and in between the rigid inclusions. A vertical dis-tance of 0.55 m has proven sufficient.

In some cases, additional reinforcing geosyn-thetics are used in the LTP where the loading is further spread out and the concentration of stres-ses is reduced. Geosynthetics have been found to be ineffective in the LTP for steel reinforced RE walls, since steel is relatively inextensible and will engage loading before geosynthetics can be effec-tive.

Studies and build structures have proven that the combination of two techniques: Terre Armée and rigid inclusions, is an economical and effi-cient solution that has the potential to accelerate construction and limit long term settlement. This solution gives designers, owners and contractors a valuable tool to design Reinforced earth over compressible soils, that can be used next to more conventional methods like soil improvement, pre-loading, LPTs or using light backfilling materials.

Figure 3Figure 2

Figure 1

In the early 1960’s Henri Vidal, introduced the Terre Armée (Reinforced Earth®) construction technique. Henri Vidal conceptualized this method and built the first full scale demonstration. For 50 years since, Terre Armée has set the standards for rein-forced soil structures and has been involved in more than 50,000 projects all over the world, accumulating knowledge and experience in the field of engineered backfills.

Reinforced soil walls over compressible soils

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