WELLINGTON INTERNATIONAL CAR PARKING BUILDING

A BROWN1; M UNO1; J STRATFORD2

1 Opus International Consultants (formerly) 2 Opus International Consultants

SUMMARY Opus International Consultants (Opus) has recently designed a new 10 storey car parking building for Wellington International Airport. The design utilises a concrete moment frame with supplemental hysteric dampers in the form of buckling restrained braces to achieve reliable performance for the frame in a high seismic region. The design was carried out using a direct displacement based design approach with non-linear verification of the design solution. This paper will outline the approach used to design this dual structural system and how the suggested construction methodology influenced the design and detailing of the structural elements. It will also explain the displacement based design methods, non-linear time history and finite element analysis that underpinned the design of this combined system. INTRODUCTION The new car parking building for the Wellington International Airport Limited (WIAL) is part of the airport redevelopment plan. It includes the construction of a new hotel building, concourse and carpark. The site is located east of the existing terminal and ultimately replaces and extends the existing at grade and elevated car parking building. The terminal and new car parking building will be separated by the newly configured concourse which is also part of the redevelopment. Opus International Consultants were engaged by WIAL to provide structural engineering services for the car parking building, concourse and the hotel. The car parking building was originally conceived as a braced steel framed structure however the new building was reconfigured as a concrete moment frame building incorporating steel Buckling Resisting Braces (BRB’s). BUILDING DISCRIPTION The car parking building is 10 storeys including a mezzanine level between ground and the first floor. The design of the building considers a possible future extension as part of the master plan. It could be extended in plan over two further phases that will be structurally independent to the first phase. The overall dimension of this building is 78 m x 51 m x 28 m high with a gross floor area of the building is approximately 35,000 m2, which will provide parking for up to 1090 cars.

Gravity Structure The suspended floor system comprises of Stahlton Double Tee units with a concrete structural topping that is typically 100mm thick. The double tee units span a maximum of 13.35m between the concrete frames which run in the transverse direction. The first floor is designed for the additional load to support 85% of the axle loading specified for HN loading under the NZ Transport Agency’s Bridge Manual. The topping thickness for the first floor is increased to 175mm to allow for this additional load. The gravity floor is supported on the transverse frames running in the east-west direction (Fig. 1). The concrete frames utilise precast concrete construction with in-situ joints for speed of erection and to improve quality of the concrete finish. The beams are typically 715mm deep x 800mm wide and 900mm deep for the first floor. The ramp structure, also utilising precast concrete to reduce the amount of on-site work, is supported on concrete walls or steel beams spanning between the main concrete frames

Figure 1 Typical floor plan The secondary elements such as lift shafts and stairs are constructed in traditional steel framed structure and precast concrete respectively. Lateral Load Resisting Structure As mentioned above, the lateral load resisting system is a dual system, incorporating BRBs as supplemental hysteretic dampers in combination with the concrete moment frame. The seismic system is designed such that the percentage of the storey shear varies from approximately a 100:0 ratio taken by the frame and BRBs respectively at roof level, to a 40:60 ratio for the transverse and 20:80 for the longitudinal at the ground floor level (Fig. 2). The concrete diaphragm distributes the lateral loads into the frame and lines of BRB resistance.

BRB location Longitudinal concrete moment frame

Transverse concrete moment frame

The concrete moment frame has been designed to remain nominally ductile at the ultimate limit state (ULS), so that residual drifts, relative to other ductile braced frame systems, are minimised and that frame dilation is also controlled to limit damage to the diaphragm. Although the displacement ductility demand on the frame is limited to a range of µ = 1.25 to µ = 1.5 at the Design Basic Earthquake (DBE) (1 in 500year return period), the concrete frame has been detailed to the relevant limited ductile provisions of NZS3101 for confinement and anti-buckling reinforcement, so that well-conditioned structural performance is assured at the maximum considered event where localised rotational demands in some beams may exceed the curvature limits for a limited ductile frame. The BRBs are a proprietary system that, along with the gussets, are designed by the specialist suppliers. Properties for the brace effective stiffness and stress strain relationships used in the design were supplied by CoreBrace. Once the actual supplier was appointed, the design was checked and confirmed using actual stiffness and overstrength values for that particular manufacturer, which for this project was also CoreBrace.

Figure 2 Typical cross section of building showing the location of BRBs Foundation Structure The foundation system for the building is a 1m deep raft slab with 1.5m deep perimeter thickening. The bases of the columns where the BRBs are located were also thickened to 1.5m locally to increase shear capacity in the raft slab. The raft design was carried out by finite element modelling of the soil structure interaction in both SAP2000 and LS-DYNA, which resulted in a more efficient shallow raft slab foundation. Structural Options The building was originally proposed to be a steel braced frame structure with a 30m deep bored piled foundation. It was the client’s request that the concrete option be considered due to the close proximity to the coastline, creating a durability issue resulting in high maintenance costs for WIAL associated with exposed steelwork. The fire protection for the steelwork was also a substantial cost. With the open sight lines within the building being a key aspect of the design, a conventional shear wall system was not considered suitable.

Precast cruciform system with in-situ joints

In-situ columns with precast beams

Both concrete frame with supplemental braces and the steel braced frame systems were presented options during the preliminary design stage and cost estimate were prepared for both options. Without taking into account the on-going future maintenance of the building, the concrete option was still evaluated as cheaper to construct. Together with concrete being more durable under exposure to adverse environmental conditions compared to steel, and architectural consideration regarding sightlines, the concrete frame option was selected as the preferred building form. Proposed Construction Sequence It was considered that the precast concrete construction would provide a cost effective construction solution considering speed of erection, quality of the concrete finish, and minimising site works compared to in-situ construction. The columns between ground floor and the underside of level 2 were constructed in-situ due to the varying floor level on the first floor making precast only cost effective for beams. The upper levels utilised precast concrete elements in a form of a double cruciform in both directions (Fig. 2). This meant that the in-situ joints were placed typically near the mid span of the beam, and the often complicated beam-column joints with the BRB gusset connections were cast in a controlled environment of the precast yard. Transportable size and weight including cranage reach were also considering factors for the location of the in-situ joints.

Figure 3 Precast Double cruciform section and typical details

The ramp structure also generally detailed as precast concrete with in-situ joints, and was also optimised to allow the largest sizes that could be transportable and craned into place with cranage available in New Zealand. The precast units were to be supported on beams allowing

Typical in-situ joint detail BRB gusset connection to concrete

Typical concrete reinforcement arrangement

for the gap between precast units to act as a seismic movement joint at mid-height of the storey (Fig. 4).

Figure 4 Typical ramp cross section

SEISMIC DESIGN The site is underlain by deep flexible soils (site subsoil class D) and located approximately 8km from the Wellington fault line. The design basis earthquake (DBE) is a 1 in 500 year return period event with a zone factor, z=0.4 for Wellington. This gives an elastic seismic design coefficient that is relatively high, so some ductility was required to keep the member sizes within the overall height and plan area limits imposed by the site constraints. Designing the concrete frame as a limited ductile or fully ductile MRF was considered, but this introduced the problem of frame dilation under the DBE and beyond. Given the preference for precast concrete flooring units for their cost and time benefits, frame dilation was a problem that would need to be solved; with a 78m overall frame length parallel to direction of the span of the double tee, loss of seating due to frame dilation in the end bays was a distinct possibility. One option was to employ a slotted beam solution for the frames in the longitudinal direction. However, the open configuration of the carpark and the adverse environmental conditions introduced other problems with cover and reinforcement protection in a critical part of the structure, which would be unlikely to receive inspection and create maintenance issues. The designers decided that this was not the best option to the frame dilation problem. The preferred solution was to design the MRF such that the potential hinge rotational demands were minimised during the DBE, which would limit any frame elongation from plastic hinge formation in the beam. To achieve a low level of plastic hinge formation and keep the member sizes within the limits that the site constraints imposed, some form of additional energy dissipation was required. The carpark layout suited the use of either fluid viscous dampers (FVDs) to provided supplemental viscous damping or BRB’s to provide supplemental hysteretic damping since these types of bracing were considered for the steel frame option. The timeframe for design was challenging and the BRB’s offered a more streamlined analysis path than FVDs, so the decision was taken to pursue the BRB option in combination with the concrete moment frame to give a hybrid system. The design objective was to limit the displacement ductility demand of the MRF to between µ=1.25 and µ =1.5 to overcome the issues associated with frame dilation and loss of seating for the precast floor units. The design of the building was carried out using a direct displacement based design approach as generally described in Priestley et el., (2007) and more formally proposed by Sullivan et el., (2012). Additional publications, such as Marley et el., (2010) that expanded on the performance and design dual systems using moment frames in combination with BRB’s were also relied on. DDBD was preferred, as it allowed an intuitive means of designing this dual

system with the focus on achieving a system stiffness that maintained a nominally ductile response in the concrete frame at the DBE, with the majority of the ductility demand being confined to the BRB’s. The initial design of the frame was based on a design displacement correlating to 1.25 times the expected yield drift in the concrete frame at the critical storey. The yield drift was initially determined by simple formulae given in Priestley et el., (2007) and adjusted for the frame geometry in the critical storey. This gave a design drift of approximately 1.8% at the critical storey. The design displacement associated with this drift was then modified to account for torsion and higher mode effects as described in Sullivan et el., (2012).

To calculate the relative seismic demand on the MRF and BRB’s the proportion of the design storey shear that the concrete moment frame would resist, βi, may be calculated as:

𝛽𝑖 =𝑉𝑖,𝑀𝑅𝐹

𝑉𝑖,𝑇𝑂𝑇𝐴𝐿

The design shear profile of the moment frame and the buckling restrained brace frame could then be determined, where, 𝑉𝑖,𝑠𝑦𝑠, is the total story shear demand on the ith storey.

𝑉𝑖,𝑀𝑅𝐹 = 𝑉𝑖,𝑠𝑦𝑠 ⋅ (𝛽) and 𝑉𝑖,𝐵𝑅𝐵 = 𝑉𝑖,𝑠𝑦𝑠 ⋅ (1 − 𝛽)

Optimising the seismic system was then an iterative process, balancing the relative benefit provided by increasing the proportion of design storey shear taken between the moment resisting frame or the buckling restrained braces. βi was initially set as the value between 1.0 (no contribution to resistance provided by BRB’s) to 0.5 (50% of the storey shear resisted by the MRF). The upper stories of the MRF where BRB’s were omitted we assigned an initial value of 𝛽𝑖 = 1.0 and 𝛽𝑖 = 0.5 for the lower stories with BRB’s. The final system configuration was then arrived at through iteration as shown in Fig 5. for the transverse direction. The equivalent viscous damping at the design displacement was calculated for both the MRF and BRB’s based on the formulae given in Sullivan et el., (2012) for concrete frames and BRB’s. The equivalent viscous damping (EVD) was calculated for each of these elements at each storey based on the displacement ductility demand at that storey. For the irregular frame layout with different beam spans but with the same beam depths, the yield drifts are dependent on the each beam span, which required calculation of the average yield drift for each storey. Likewise, the same averaging was carried out for BRB’s of different lengths and arrangements in each storey. The standard 5% elastic damping was adjusted to an appropriate elastic damping value. Considering the design for a nominally ductile frame and lack of non-structural elements an elastic damping value of 3% is used for both the concrete frame and buckling restrained braces. The proportion of overturning moment resisted by each system was determined using statics with the storey force distributions from the relative split between the two systems. Using the overturning moment resisted by the respective systems, the individual system equivalent viscous damping values were weighted to obtain the dual system equivalent viscous damping, as follows:

𝜉𝑒𝑞,𝑠𝑦𝑠𝑡𝑒𝑚 = 𝑀𝑂𝑇𝑀,𝑀𝑅𝐹𝜉𝑒𝑞,𝑀𝑅𝐹 + 𝑀𝑂𝑇𝑀,𝐵𝑅𝐵𝜉𝑒𝑞,𝐵𝑅𝐵

𝑀𝑂𝑇𝑀,𝑀𝑅𝐹 + 𝑀𝑂𝑇𝑀,𝐵𝑅𝐵

This gave an equivalent viscous damping for the combined system of ξeq,sys = 12.3% in the transverse direction. This comprised EVD values at the design displacement of ξeq,MRF = 7.8% and ξeq,BRB = 19.8% for frame and BRB’s respectively, with a ratio of MRF to BRB contribution to the overall damping of approximately 1.6:1.

Figure 5: DDBD displacement profile, storey drifts, and storey shear split between components Non-Linear Static Pushover Analysis Once the initial frame design was arrived at for both directions, a 2-D non-linear static pushover of the frame was carried out to confirm the initial assumptions made about the displaced shape of the structure and the storey force distribution between the moment frame and BRBs. Firstly, a non-linear push of the concrete frame was performed to confirm the initial assumption about the yield displacement of the frame, and hence the target design displacement was correct. The final non-linear force-displacement profiles for the MRF, BRB’s and combined system are shown in Fig 6.

The storey forces for the system were then applied incrementally to the combined model in relative proportion until the design displacement was reached at the effective height to give a revised displaced shape. This revised displaced shape was then used to update the displacement based.

Figure 6: Non-Linear Force-Displacement Profile for Combined System and Components

The resulting values were compared to initial assumptions for deflected shape and strength distribution between systems and EVD. If there was an unacceptable difference between the assumption and the values resulting from finite element analysis, the values in the DDBD were revised and the DDBD process was repeated until agreement was achieved. The non-linear static pushover was carried out using 2-D analysis in ETABs. The BRB’s were modelled as link elements based on a tri-linear backbone curve adopting the stress strain properties supplied by CoreBrace. The beams and columns were modelled as linear frame elements, with lumped plasticity at the potential plastic hinge regions. The backbone curves for the concrete frame elements were determined from a moment-curvature analysis using Response2000. Boundary conditions were modelled as rotational springs at the base of the columns determined from the stiffness of the raft foundation system. After some iteration the displaced shape from the pushover analysis and the target design displacement were in close agreement, and the design of the frame and BRB elements proceeded. Concrete Frame Design Although the displacement demand of the concrete frame was limited to a ductility of less than µ=1.5, the frame was design to meet the limited ductility provisions of NZS3101. This provided a high level of resilience in the system for minimal additional cost. BRB Design The design of the BRB’s was based on units provided by CoreBrace. The braces were designed for relatively high ductility demand; typically µ=10 to µ=12. This was considered feasible as the concrete moment frame provides re-centring force to prevent ratcheting of a high ductility BRB system in close proximity to the Wellington fault line, where strong forward directivity effects are possible.

The braces were designed with pinned-welded end connections. Welding the lower end of the connection allowed the construction tolerance necessary in interfacing a steel system with ±2 mm tolerance with the concrete frame with ±10 mm to 15 mm between cast in gusset plates. Pinning the top end allowed for rotation of the brace to occur, as the system underwent lateral sway. The buckling restrained braces were designed following the approach embodied in ASCE 7/10 and AISC 360. The overstrength factors from the BRB used for capacity design of the gussets, gusset connection to frame, and frame axial actions initially followed this approach accounting for the lower bound yield to upper bound yield ratio along with factors for strain hardening and casing influence on the compression stroke. To reduce the overstrength design actions the ratio of the expected yield strength of the steel core material to the upper bound strength was used for capacity design rather than the lower bound yield. The actual yield stress was then determined from coupon testing for the core of each BRB and the core area was altered slightly to achieve the same BRB yield strength as adopted for the assumed expected yield strength values. The gussets were designed using the uniform force method as per Feeney & Clifton (1995) and AISC (2012) with the connection of the gusset to the concrete frame designed using strut and tie methods of NZS3101. Ramp Design The ramps were split at the mid height to allow for differential floor movements, however this caused large actions to resist the lateral loads of the ramp as a Part when calculated using the Parts coefficient from NZS1170.5. To mitigate this, acceleration spectra were developed for various levels of the building from the time history analysis floor acceleration results. This reduced the parts loading from the ramps to more manageable values. DESIGN VERIFICATION As noted above, the design was completed using simple static analyses based on DDBD principles, and non-linear methods were used to verify this design and confirm our understanding of the expected building response. The non-linear static pushover model described above was used to carry out 2-D non-linear time history analyses (NLTHA) for the set of ground-motion records in Table 1. The ground-motion selection includes four scaled records featuring strong near field effects (from different events worldwide) taken from the suggested earthquake record suites of both Tarbali (2014) and Oyarzo (2009). These records have been selected from the NGA database of strong ground motions from shallow crustal earthquakes with site to source distances of <25 km. All records within the set are scaled to the ULS (500 year return period) and MCE (2500 year return period) response spectra, in accordance with NZS 1170.5 (2004).

Table 1. Selected Earthquake Records

NGA # Event Station Year Mag Mechanism

1176 Kocaeli Yarimca 1999 7.51 Strike-Slip

778 Loma Prieta Hollister Diff. Array 1989 6.93 Reverse-Oblique

725 Superstition Hills Poe Road (temp) 1987 6.54 Strike-Slip

1491 Chi-Chi TCU051 1999 7.62 Reverse-Oblique

The NLTHA was evaluated in terms of storey drifts, ductility demand in the BRB’s and hinge rotations in the concrete frame at both the DBE and MCE events. Member actions in the BRB’s and frame elements were also evaluated and compared with the actions used for design of the members. The displacement response and storey drift profile for the DBE are shown in Fig. 7(a) and Fig. 7(b) respectively as well as the comparison with the displacement and storey drift values used in the DDBD. This shows the NLTHA displacements to be less than predicted by the DDBD, which may be due to a number of factors that were not considered in the DDBD, such as: Sp factor used for NLTHA; response includes modes other than the fundamental mode; rigid zones being larger at completion of design than anticipated in DDBD; and, contribution from columns about their weak axis. The NLTHA confirmed that the design actions, BRB strains and frame hinge rotations calculated from the DDBD process were within the limits of the designed solution. A comparison of the equivalent viscous damping provided by the system at the peak response from the NLTHA for the DBE was compared with that calculated for the DDBD. The values of ξeq,sys = 7% and ξeq,sys = 12% in the transverse direction for the NLTHA and DDBD respectively showed the damping for the NLTHA was compatible with that assumed for the DDBD once the smaller storey displacements were considered.

(a) (b) (c) Figure 7: NLTHA response for combined system in transverse direction showing peak

response for (a) displacement profile, (b) storey drift, and (c) residual drift. Benefits of the Dual System As stated above the member sizes and geometry of the moment frame were limited by site constraints and architectural considerations. NLTHA results for the concrete moment frame without BRBs showing displacement, store drifts, and residual drift are given in Fig. 8 for comparison with the combined system. Fig. 8(b) demonstrates that the storey drifts limit of 1170.5 are exceeded at the critical storey for the standalone MRF.

The concrete moment frame in combination with steel BRBs offered a number of benefits over either of these systems as a standalone solution. Some of the benefits and difficulties that this type of combined system offer include:

A concrete moment frame with these section sizes would have required a greater ductility demand in the frame. This would have required detailing to avoid the potential loss of seating for the system associated with frame dilation, which was avoided by the use of a combined system.

(a) (b) (c) Figure 8: NLTHA response for frame only system in transverse direction showing peak

response for (a) displacement profile, (b) storey drift, and (c) residual drift.

Larger members would have been required for a moment frame without supplementary bracing to meet the 1170.5 drift limits, as demonstrated in Fig 8(b).

A reasonable degree of hysteric damping could be provided by the BRB system utilising a ductility in the braces of upwards of µ=10, but the residual drift associated with BRB’s at this ductility was avoided by the re-centring force provided by the elastic strain energy in the frame. The residual drift of the combined system at the DBE was negligible (Fig. 7(c)) compared to the MRF alone, which was up to 1.0% (Fig. 8(c)).

The distribution of the overstrength axial loads that the foundation had to resist at specific locations could be managed more easily by the combinations of the moment frame and BRBs.

The detailing of the gusset connection to the frame for the combined system was difficult.

The concrete frame can be erected and the braces installed later with no temporary lateral bracing required as for a pinned steel BRB frame.

The analysis and design effort for this combined system is substantially greater than for either of the standalone systems.

There is added complexity in construction due to detailing at the gusset locations to transfer forces into the moment frames.

CONCLUSION Concrete moment frames with supplemental hysteretic dampers provide a cost-effective and pragmatic solution to overcome some of the deficiencies associated with either as a standalone system; the performance of the whole is better than the parts. That is not to say that the combined system is not without some challenges. The additional computation and resource in the design phase, the complexity of the gusset design and the force transfer into the moment frame, and fabrication and erection tolerances required for this system should not to be overlooked. However, on the whole, the combined system appears to have provided an effective solution for this project that, with the experience and insights the designers gained on this project, could be improved upon for future applications. REFERENCES AISC (2012) 2nd Edition Seismic Design Manual, American Institute of Steel Construction.

ASCE Standard (ASCE 7-10 second printing, 2011), Minimum Design loads for Buildings and Other Structures, American Society of Civil Engineers Feeney M.J, Clifton G.C, 1995. HERA Report R4-76, Seismic Design Procedures for Steel Structures, Hera Clifton G.C, 1994. HERA Report R4-80, Structural Steelwork, Limit State Design Guides, Volume 1, Hera Marley, T.J., Sullivan, T.J., Della Corte, G. (2010) Development of a Displacement-Based Design Method for Steel Dual Systems with Buckling-Restrained Braces and Moment-Resisting Frames, Journal of Earthquake Engineering, 14:S1, 106-140 New Zealand Standard (NZS3101:2006) Concrete Structures Standard New Zealand Standard (NZS3404:1997), Steel Structures Standard New Zealand Standard (NZS1170.5:2004), Structural Design Actions – New Zealand New Zealand Standard (NZS1170.5:Supp 1:2004), Structural Design Actions – New Zealand - Commentary Oyarzo-Vera, C., McVerry, G., and Ingham, J. M. (2011). Seismic zonation and default suite of ground motion records for time-history analysis in the North Island of New Zealand. Submitted (081510EQS105M) August 2010 to Earthquake Spectra Priestley, M.J.N., Calvi, G.M., Kowalsky, M.J. (2007) “Displacement-Based Seismic Design of Structures” IUSS Press, Pavia, Italy. Sullivan, T.J., Priestley, M.J.N., Calvi, G.M. (2012) “A Model Code for the Displacement-Based Seismic Design of Structures” IUSS Press, Pavia, Italy. Tarbali, K., Bradley, B. (2014) Ground-motion selection for scenario ruptures using the generalized conditional intensity measure (GCIM) approach and its application for several major earthquake scenarios in New Zealand. University of Canterbury. 92pp.