University of Southampton Research Repository

ePrints Soton

Copyright © and Moral Rights for this thesis are retained by the author and/or other copyright owners. A copy can be downloaded for personal non-commercial research or study, without prior permission or charge. This thesis cannot be reproduced or quoted extensively from without first obtaining permission in writing from the copyright holder/s. The content must not be changed in any way or sold commercially in any format or medium without the formal permission of the copyright holders.

When referring to this work, full bibliographic details including the author, title, awarding institution and date of the thesis must be given e.g.

AUTHOR (year of submission) "Full thesis title", University of Southampton, name of the University School or Department, PhD Thesis, pagination

http://eprints.soton.ac.uk

UNIVERSITY OF SOUTHAMPTON

Faculty of Engineering, Science and Mathematics

School of Civil Engineering and the Environment

Performance of a propped retaining wall at the

Channel Tunnel Rail Link, Ashford

Jo Clark

•

Thesis for the degree of Doctor of Philosophy

June 2006

UNIVERSITY OF SOUTHAMPTON

ABSTRACT

FACULTY OF ENGINEERING, SCIENCE AND MATHEMATICS

SCHOOL OF CIVIL ENGINEERING AND THE ENVIRONMENT

Doctor of Philosophy

PERFORMANCE OF A PROPPED RETAINING WALL AT THE CTRL, ASHFORD

Jo Clark

This thesis is based on the field monitoring of a propped bored pile retaining wall installed

in an overconsolidated clay. Pile bending moments, prop loads, pore water pressures and

lateral earth pressures were logged automatically at intervals of up to 5 minutes

throughout construction (and for 4 years afterwards) and wall deflections were measured

during construction, making this the most comprehensive instrumentation project of its

kind.

The magnitude of the over-read associated with the use of spade cells (used to measure

lateral earth and pore water pressures) in overconsolidated deposits was determined by

comparing readings from a spade cell aligned to measure vertical stress with the estimated

overburden acting on it as the overburden was excavated. This study adds significantly to

the previous data as spade cells have not previously been \lsed in the Atherfield Clay, and

the performance of spade cells under a known changing load has not previously been

measured in the field.

Analysis of the changes in lateral stress and pore water pressure during the wall

installation process showed significant reductions in horizontal stress during wall

installation, reducing the ratio of effective horizontal to effective vertical stress, K, from

about 1 to nearly the active condition. Following wall installation there was no further

change in horizontal stress over a period of about 10 months, during which time no further

construction work took place.

Analysis of the data yielded good agreement between pile bending moments estimated

from inclinometer and strain gauge measurements in the piles, and the onset of concrete

cracking was identified. The components of strain measured in the reinforced concrete

props due to shrinkage, creep and applied load were also identified, allowing prop loads to

be estimated. A simple equilibrium calculation showed that these agree with the measured

wall bending moments and total horizontal soil stresses, demonstrating the overall

consistency of the data collected.

Simple equilibrium analysis of the behaviour of the wall during construction shows that

the soil stresses measured are compatible with the measured structural loads. The long­

term horizontal soil stresses, bending moments and RC prop loads show no increase over

the 6 years since construction began.

TABLE OF CONTENTS

Index of Figures

Index of Tables

1

2

Introduction

1-1 In situ embedded retaining walls

1-2 Measurement of horizontal stress

1-3 Channel Tunnel Rail Link

1-4 Objectives

1-5 Outline of report

Literature Review

2-1

2-2

Introduction

Review of investigations into retaining wall behaviour

2-2-1

2-2-2

2-2-3

2-2-4

2-2-5

2-2-6

Background

Limiting conditions

Development of active and passive pressures

Distribution of earth pressures

Change in horizontal earth pressure due to wall installation effects

Behaviour of embedded walls in the longer term

2-3 General overview of retaining wall design standards

2-3-1 Choice of soil parameters

2-3-2 Application of safety factors

2-3-3 Calculation of design bending moments

2-3-4 Advice on wall installation effects

2-4 Use of spade cells to measure horizontal stress in overconsolidated clay

2-4-1 Background

2-5 Conclusion

3 Case study

3-1 Introduction

3-2 Geology

3-2-1 Geotechnical data and sample collection

3-2-2 Site location and history

Vll

Xll

3

4

5

5

8

8

9

9

9

11

12

15

17

18

20

21

23

23

24

24

26

31

31

31

32

33

3-2-3 Reported geology

3-2-4 Observed geology

3-3 Geotechnical properties

3-3-1 Plasticity

3-3-2 Bulk density

3-3-3 Soil strength

3-3-4 In situ horizontal stress

3-3-5 Permeability and Groundwater

3-4 Description of site and geometry of structure at the instrumented section

3-5 Construction sequence and installation of instrumentation

3-5-1 Pile installation

3-5-2

3-5-3

3-5-4

3-5-5

3-5-6

3-5-7

3-5-8

3-5-9

Sand drains

Removal of pile tops

Capping beam

Reinforced concrete props

Backfilling and material placement behind North wall capping beam

Excavation

Temporary props

Base slab

3-5-10 Installation of storm drain

3-6 Data collection and vibrating-wire instrument technology

4 Structural monitoring

4-1

4-2

Introduction

Inclinometer

4-2-1 Description and Use

34

36

38

38

38

39

40

41

42

43

45

47

48

49

49

49

50

50

51

51

51

97

97

98

98

4-2-2 Calculation of bending moments from inclinometer data 1 01

4-3 Vibrating-wire gauges 104

4-3-1 Use 104

4-3-2 Method for calculation of prop loads from strain gauge measurements 104

4-3-3 Method for calculation of bending moments from strain gauges 105

4-3-4 Correction for temperature effects

4-4 Analysis of instrumentation data

4-4-1 Temporary prop data

4-4-2 Base slab data

4-4-3

4-4-4

Reinforced concrete prop data

Pile data

107

107

107

108

109

115

11

4-4-5 Comparison between pile bending moments calculated from strain gauges and inclinometer 118

4-5 Measured changes due to construction events

4-5-1

4-5-2

4-5-3

Temporary props

Base slab data

Reinforced concrete props

4-5-4 Pile bending moments measured with strain gauges

4-5-5 Inclinometer data

4-6 Conclusions

5 Horizontal soil stress measurement

5-1

5-2

5-3

5-4

Introduction

Spade cell description

Calibration

Evaluation of spade cell over-reading

5-4-1

5-4-2

5-4-3

Introduction

Test procedure

Results and Discussion

5-5 Installation effects of a bored pile retaining wall in overconsolidated clay

5-5-1

5-5-2

5-5-3

5-5-4

5-5-5

5-5-6

5-5-7

Introduction

Calibration and installation of spade cells

Stabilization of spade cells following insertion

In situ total horizontal stresses and pore water pressures

Effective stress profile

Pore water pressure changes during and after wall installation

Total horizontal stress changes due to wall installation

5-5-8 Effective stress changes due to wall installation

5-5-9 Total and effective horizontal stress changes after wall installation

5-6 Effect of excavation in front of the wall on horizontal earth stresses

5-7 Conclusions

6 Long term monitoring

6-1 Introduction

6-2 Total horizontal stress and pore water pressure measurements

6-2-1 Installation of storm drain

6-3 Reinforced concrete prop loads

6-4

6-5

Pile bending moments

Conclusions

119

119

120

120

121

124

125

162

162

162

163

163

163

164

165

169

169

169

169

169

170

170

172

175

176

176

179

216

216

216

217

218

218

219

11l

7 Conclusions and recommendations

7-1

7-2

Conclusions

Recommendations for further work

Appendix A Wireline borehole BH1 and BH2 logs

Appendix B Piling diary

Appendix C Calculation of flexural rigidity of pile (El)

References

237

238

239

241

250

253

254

IV

LIST OF FIGURES

Figure 1-1 Map showing the route of the Channel Tunnel Rail Link 7

Figure 2-1 Coulomb's idealized earth pressure distribution 27

Figure 2-2 Fixed earth support 27

Figure 2-3 Free earth support 27

Figure 2-4 Relationship between earth pressure and wall rotation measured by Terzaghi for normally consolidated sand (modified by Simpson, 1992) 28

Figure 2-5 Relationship between earth pressure and wall rotation computed by Potts and Fourie (1986) for overconsolidated clay 28

Figure 2-6 Influence of movement type on pressure distribution 29

Figure 2-7 Stress distributions behind and in front of (a) stiff and (b) flexible embedded walls (after Rowe, 1952) 29

Figure 2-8 Over-reading v. cu. After Ryley and Carder, 1995 30

Figure 3-1 The relative locations of the areas described in this work 54

Figure 3-2 Geology of the Weald of Kent 55

Figure 3-3 Layering in the Weald Clay 56

Figure 3-4 Light brown band defining the boundary between the upper and lower Atherfield Clay (observed in the nadir sump) 57

Figure 3-5 Showing the peds and softer matrix of the upper Atherfield Clay 57

Figure 3-6 In situ block sample of the Hythe Beds 58

Figure 3-7 Photo looking into the cutting between RC props during excavation 59

Figure 3-8 Photo showing an elevation of the excavation face 59

Figure 3-9 Shear plane in the upper Atherfield Clay, 50 m from the instrumented section 60

Figure 3-10 Sketch taken of discontinuity observed in the bottom of the excavation at the instrumented section 61

Figure 3-11 Geotechnical profile including liquid and plastic limits from wireline boreholes 1 and 2 62

Figure 3-12 Comparison of geology at locations referred to in this study 63

v

Figure 3-13 Bulk density v depth 64

Figure 3-14 Effective stress paths of lower Atherfield Clay 65

Figure 3-15 Undrained shear strength v depth from laboratory samples and 66 standard penetration test results

Figure 3-16 Undrained shear strength v depth from self-boring pressuremeter test 67

Figure 3-17 Total horizontal stress measured by self-boring pressuremeter 68

Figure 3-18 Permeability data 69

Figure 3-19 Pore water pressure: measurements and assumed profile 70

Figure 3-20 Location of the instrumented section 71

Figure 3-21 Cross-section of the cutting at the instrumented section 72

Figure 3-22 Elevation illustrating the ground profile around the structure 73

Figure 3-23 Plan of instrumented section 74

Figure 3-24 Elevation of instrumented section 74

Figure 3-25 Pile installation sequence 75

Figure 3-26 Photo showing pile boring 76

Figure 3-27 Photo showing installation of casing 77

Figure 3-28 Photo showing installation of reinforcement cage 78

Figure 3-29 Elevation of instrumented section showing strain gauge arrangement 79

Figure 3-30 Strain gauges for measurement of pile bending moment 80

Figure 3-31 Photo showing sand drain and cable for spade cell 11 81

Figure 3-32 Changes in total horizontal stress and pore water pressure in front of the wall around the period of sand drain installation 82

Figure 3-33 Excavation either side of the north wall before pile top removal 83

Figure 3-34 Pile top removal 84

Figure 3-35 Wall before construction of capping beam 85

Figure 3-36 Dimensions of capping beam 86

Figure 3-37 Photo showing capping beam 86

Figure 3-38 Preparation for RC prop construction 87

Figure 3-39 Relative location of the instruemented props and piles 88

Figure 3-40 Photo showing instrumented area before excavation 89

Figure 3-41 Compacting backfill in the instrumented section 90

Figure 3-42 Elevation showing props and order of excavation 91

VI

Figure 3-43 Excavation using long-reach excavator 92

Figure 3-44 Temporary prop arrangement and strain gauge housings 92

Figure 3-45 Temporary prop gauges welded onto prop 93

Figure 3-46 Elevation showing base slab arrangement 94

Figure 3-47 Strain gauges wired onto base slab reinforcement 94

Figure 3-48 Storm drain installed behind north wall in November 2001 (Day 864) 95

Figure 3-49 Datalogger setup 96

Figure 3-50 Vibrating-wire gauge 96

Figure 4-1 Inclinometer probe 127

Figure 4-2 Analysis of inclinometer data for face errors 127

Figure 4-3 Movement of wall toe with time measured by the inclinometer 128

Figure 4-4 Correction of inclinometer data 129

Figure 4-5 Movement of wall toe and bottom 10m of inclinometer tube 130

Figure 4-6 Inclinometer measurements taken during construction 131

Figure 4-7 Movement of the top of the wall 132

Figure 4-8 Typical relationship between load and temperature (temporary prop) 133

Figure 4-9 As measured temporary prop loads 134

Figure 4-10 Temporary prop loads corrected for temperature effects 135

Figure 4-11 Base slab readings 136

Figure 4-12 Temperature v load for a typical base slab gauge 137

Figure 4-13 Strains measured in the base slab gauges 138

Figure 4-14 Strain measured in reinforced prop PI 140

Figure 4-15 Strain measured in reinforced prop P4 140

Figure 4-16 Strain measured in reinforced prop P2 141

Figure 4-17 Time/strain versus time for an RC prop gauge at chainage 89+205 143

Figure 4-18 Measured and calculated strain against time for RC prop gauges at chainage 89+205 143

Figure 4-19 Prop loads uncorrected for effects of shrinkage 144

Figure 4-20 Prop loads corrected for effects of shrinkage using Ross' equation 145

Figure 4-21 Close up showing up to after Excavation Phase 2 from Figure 4-20 146

Figure 4-22 Change in stiffness with time for RC prop 147

VB

Figure 4-23 Temperature v load for a typical RC prop gauge 147

Figure 4-24 Difference in strain measured in pile gauges over construction period 148

Figure 4-25 Strain measurements over period of construction for some gauges 149

. Figure 4-26 Deflections measured in pile Z around time of Excavation Phase 2 and Temporary Prop Removal 150

Figure 4-27 Photos showing bulge in pile concrete 151

Figure 4-28 Strains measured in Pile Y over the 30 days after excavation and 152 before Base Slab Construction

Figure 4-29 Bending moments calculated before and after Excavation Phase 2 from inclinometer and strain gauge measurements 154

Figure 4-30 Bending moments calculated before and after Temporary Prop Removal from inclinometer and strain gauge measurements 155

Figure 4-31 Individual temporary prop loads 156

Figure 4-32 Pile Y bending moments plotted against time (below base slab) 158

Figure 4-33 Pile Y bending moments plotted against time (above base slab) 159

Figure 4-34 Pile X bending moments plotted against time (below base slab) 160

Figure 4-35 Pile X bending moments plotted against time (above base slab) 161

Figure 5-1 Spade cell 181

Figure 5-2 Spade cell calibration equipment 182

Figure 5-3 Equipment for drilling horizontal borehole 183

Figure 5-4 Nadir sump plan and elevation 184

Figure 5-5 Spade cell: uncovered in excavation 185

Figure 5-6 Total vertical stress measured by spade cell in nadir sump 186

Figure 5-7 Elevations and plans detailing the excavation of material above the 187 spade cell in the nadir sump

Figure 5-8 Measured total stress versus overburden 188

Figure 5-9 Nadir sump spade cell stiffness calibration ]89

Figure 5-10 Finite element analysis mesh ]90

Figure 5-11 Measured and calculated vertical stresses against overburden 191

Figure 5-12 Over-read error: measured and calculated from limit equilibrium analysis 192

Figure 5-13 Comparison between spade cell readings corrected by 0·35 Cu and self-boring pressuremeter test results 193

Vlll

Figure 5-14 Spade cell installation 194

Figure 5-15 Spade cell measurements taken between their installation and wall installation 195

Figure 5-16 Stabilized readings of total horizontal stress and pore water pressure from all spade cells (before wall installation) 196

Figure 5-17 The profile of 0\0 with depth 197

Figure 5-18 Piezometer readings over the period of wall installation 198

Figure 5-19 Pore water pressures measured before, during and 10 months after 199 wall installation

Figure 5-20 Total horizontal stress measurements taken during the period of wall installation 200

Figure 5-21 Pile installation sequence for the elastic analysis and Mohr circle showing calculation of correction for stress change measured on spade cell 201

Figure 5-22 Measured reduction in total horizontal stress normalised with respect to the in situ total horizontal stress 202

Figure 5-23 Reduction in total horizontal stress due to installation of pile nearest to spade cell only with distance from the wall compared with the elastic prediction 203

Figure 5-24 Total horizontal stresses measured by all spade cells before wall 204 installation and by those 1·275 m from the edge of the wall after wall installation

Figure 5-25 Change in total horizontal stress 205

Figure 5-26 Effective horizontal stresses measured by all spade cells before wall 206 installation and by those 1·275 m from the edge of the wall after wall installation

Figure 5-27 Change in effective horizontal stress 207

Figure 5-28 Total horizontal stress measured before, during and 10 months after wall installation 208

Figure 5-29 Total pressure and pore water pressure measured 1·275 m behind the wall during the construction period 210

Figure 5-30 Total pressure and pore water pressure measured 2·375 m behind the wall during the construction period 211

Figure 5-31 Total pressure and pore water pressure measured 3·475 m behind the wall during the construction period 212

Figure 5-32 Total horizontal stress and pore water pressure measured in front of the wall during the construction period 213

Figure 5-33 Measured bending moments and bending moments calculated from horizontal soil stresses within 10% of those measured 214

IX

Figure 5-34 Measured and estimated total horizontal stresses 215

Figure 6-1 Long-term total horizontal stress measured 1·275 m behind the wall 220

Figure 6-2 Long-term pore water pressure measured 1·275 m behind the wall 221

Figure 6-3 Long-term total horizontal stress measured 2·375 m behind the wall 222

Figure 6-4 Long-term pore water pressure measured 2·375 m behind the wall TY' ---'

Figure 6-5 Long-term total horizontal stress measured 3·475 m behind the wall 224

Figure 6-6 Long-term pore water pressure measured 3·475 m behind the wall 225

Figure 6-7 Long-term total horizontal stress measured in front of the wall 226

Figure 6-8 Long-term pore water pressure measured in front of the wall 227

Figure 6-9 Total horizontal stress and pore water pressure measured 1·275 m 228 behind the wall over period of storm drain installation

Figure 6-10 Total horizontal stress and pore water pressure measured 3-475 m 229 behind the wall over period of storm drain installation

Figure 6-11 Immediate response of spade cells nearest to excavation of storm 230 drain trench

Figure 6-12 Long-term reinforced concrete prop loads 231

Figure 6-13 Long-term RC prop loads corrected for the effects of creep 232

Figure 6-14 Long-term bending moments in Pile Y - gauges 1-14 233

Figure 6-15 Long-term bending moments in Pile Y - gauges 15-26 234

Figure 6-16 Long-term bending moments in Pile X - gauges 1-14 235

Figure 6-17 Long-term bending moments in Pile X - gauges 15-26 236

x

LIST OF TABLES

Table 3-1: Geological succession 34

Table 3-2: Soil parameters of the Atherfield Clay and Weald Clay 40

Table 3-3: Dates of main construction events 44

Table 4-1: Example of inclinometer data analysis and correction using face errors: 41·593 mAOD 99

Table 4-2: Flexural rigidity of concrete piles 105

Table 4-3: The cracking moment of the wall, Mer, calculated by different methods 106

Table 4-4: Values of a and b calculated for all gauges at chainage 89+205 m 111

Table 4-5: amount of strain occurring due to shrinkage and equivalent load indicated by this strain for the props at the instrumented section 112

Table 4-6: Values of constants for best-fit line to stiffness/time relationship (to 5 significant figures) 114

Table 4-7: Changes in deflection, 8, (mm) (relative to the toe) and changes in curvature, K,

(x 103m-I) (from the 5th order polynomial curve fit) over periods of pile cracking.

Positions where cracks have occurred are highlighted. 1 17

Table 4-8: Changes in RC prop load due to excavation directly under individual props 121

Table 4-9: Load at specific stages of construction 121

Table 5-1: Details of nadir sump excavation 165

Table 5-2: Measured and estimated total horizontal soil stress behind the wall 177

Table 5-3: Measured and estimated total horizontal soil stress in front of the wall 178

Table 5-4: Measured and calculated bending moments 178

Table 5-5: Increase in prop load over period where no construction activities were carried out during Excavation Phase 1 179

Xl

PAPERS ARISING FROM THIS THESIS

Clark, J., Richards, D.J. and Powrie, W. (2004). Wall installation effects - preliminary

findings from a field study at the CTRL, Ashford. Proc. of the Skempton Memorial

Conference, London.

Richards, D.J., Clark, J., Powrie, W. and Heyman, G. (2005). An evaluation oftotal

horizontal stress measurements using push-in pressure cells in an overconsolidated clay

deposit. Proc. Inst. Civ. Engrs. Geotech. Engng - acceptedfor publication.

Clark, J. and Richards, D.J. (2005). Measurement of bending moments in concrete. Proc.

16th Int. Con! Soil Mech. Grnd Engng., Japan.

Richards, D.J., Clark, J. and Powrie, W. Installation effects of a bored pile wall in

overconsolidated clay. Submitted to Geotechnique.

XlI

ACKNOWLEDGEMENTS

First I must thank my supervisor David Richards for his valuable support and

encouragement throughout the fieldwork, analysis and write-up periods of this work. I

would also like to thank William Powrie and Chris Clayton for advising me and all three

for being a fantastic combination of mentors.

The work described in this thesis was carried out with the support of the Engineering and

Physical Sciences Research Council, (EPSRC), Rail Link Engineering, Union Railways

and Skanska.

I would particularly like to thank, Howard Roscoe, Gary Holmes (who's invaluable

support as a surrogate supervisor will not be forgotten), Fleur Loveridge, Hilary Shields,

Adam Chodorowski, Alex Pendleton, Vernon Pilcher, David Twine and the site staff at

Ashford Contract 430. I would like to thank Richard Wilson for collecting the bulk of the

very high quality inclinometer readings.

I would like to thank the Geotechnical Group the many other members of the School of

Civil and Environmental Engineering who have made my experience working at the

University of Southampton so enjoyable. Of particular note are Harvey Skinner, Antonis

Zervos, Michelle Theron, Martin Rust, Geoff Watson, Joel Smethurst and my office

mates.

Thank you to my parents and brother for leaving me to it when I asked, for help with proof

reading, for their constant support, and particularly for putting me up and feeding me,

often at late notice, during the fieldwork period.

If it wasn't for Nellies heroes I would have completed this work much sooner, but would

be a much poorer soul for it.

Thanks and love go. to Jon for the support, for the encouragement during the despair, for

sharing my excitement throughout the triumphs and for understanding what it is all for.

Xlll

1 INTRODUCTION

1-1 In situ embedded retaining walls

In situ embedded retaining walls are formed by installing a structure in a pre-excavated

cavity in the ground and removing the soil from one side of the structure to form a vertical

retaining wall. The term in situ is used to indicate that the wall has been constructed on

site from raw materials (fresh concrete and steel reinforcement). To provide stability, the

bottom of the wall remains embedded in the soil and sometimes props are used, for

example at the top ofthe wall, to provide extra support. These walls are typically used to

form basements and cuttings in urban areas throughout the world, where space is

restricted.

Traditionally, in situ embedded retaining walls have been designed using limit equilibrium

techniques, which are known to provide sound design solutions for walls installed in sand

and normally consolidated clays. More recently finite element techniques have been used

during design and to back analyse existing walls. Back analyses of walls constructed in

overconsolidated clays (which, due to their stress history, have a high horizontal to

vertical stress ratio compared to other natural soil deposits) have generally calculated

much higher bending moments and prop loads than both those obtained using

conventional limit equilibrium methods based on fully active pressures behind the wall

and those observed directly in the field (e.g. Potts & Fourie, 1984 & 1985; Tedd et 01.,

1984).

There are many uncertainties associated with the design of in situ embedded retaining

walls in overconsolidated deposits, for example: the effect of their installation on the stress

state of the surrounding soil; the equilibrium pore water pressures and their effect on the

overall wall stability; and the long-term horizontal stress acting on the wall. These issues

have been the subject of considerable research over the last few decades, including several

projects which used instrumentation andlor in situ testing to investigate the behaviour of

walls installed in overconsolidated clays (e.g. Garrett & Barnes, 1984; Tedd et ai., 1984;

Symons & Tedd, 1989; Ng, 1992; Symons & Carder, 1993; Carswell etai., 1993). In

order to further investigate the long-term behaviour of retaining walls, attempts have been

made to analyse walls already in service (Symons & Tedd, 1989; Carder & Symons, 1989;

Symons & Carder, 1990).

Some of these field studies have investigated the effect of in situ embedded retaining wall

installation on the stress state of the surrounding soil (e.g. Tedd et ai., 1984; Symons &

Carder, 1993) as it is generally accepted that the effect of installing such a wall influences

its subsequent behaviour in terms ofloads and bending moments (Gunn et ai., 1993;

Powrie & Kantartzi, 1996). Significant reductions in horizontal stress were observed in the

field studies. In addition, wall installation may cause significant ground movements

(which may be highly unacceptable) and its contribution to the recent stress history of the

soil will affect the subsequent stress-strain response of the soil (Powrie et at., 1998). Using

finite element analysis, Powrie & Li (1991) used a simplified technique to model wall

installation and calculated smaller wall bending moments than those obtained by Potts &

Fourie (1984 & 1985), which were closer to those measured in the field. However, the

process of wall installation in panels or piles is complex and three dimensional, so that the

magnitude, extent and longevity of the stress reduction associated with in situ wall

installation are all highly uncertain issues.

Many design standards are used for the design of in situ embedded retaining walls

(including CIRIA 104 (Padfield and Mair, 1984); CIRIA C580 (Gaba et ai., 2003);

Eurocode 7, (British Standards Institution, 1995) and BD42 (Department of Transport,

2

2000) and each contains differing advice on the method of analysis, use of factors of

safety and selection of input parameters. The overall stability of these walls and conditions

at collapse have been studied in depth (e.g. Burland et aZ., 1981; Symons, 1983; Bolton &

Powrie, 1987; 1988), however the number of standards in use illustrates clearly that there

are still many uncertainties as to the best method for design.

A better understanding of the changes in stress that occur on installation of an in sitll

retaining wall, during excavation in front of the wall and in the long-term, could result in

better estimates of prop loads, wall bending moments and ground settlements than were

possible in the past, and lead to more economical designs. In this thesis the results from a

field study carried out on a section of in situ embedded retaining wall which forms part of

the Channel Tunnel Rail Link at Ashford, Kent, are presented and discussed. Strain gauges

were used to measure permanent and temporary prop loads and wall bending moments,

and push-in spade-shaped pressure cells were used to measure total horizontal stress and

pore water pressure in the soil adjacent to the wall. Conclusions are drawn regarding the

magnitude, pattern and longevity of horizontal stress changes due to wall installation and

the overall stability of the wall. Data collection is ongoing, and data that describe the

performance of the wall and the soil stresses existing 6 years after the retaining wall was

installed and 4 years after construction was completed are included in this thesis.

Observations are made on the long-term behaviour expected from consideration of the

measurements taken in the 4 years since construction was completed.

1-2 Measurement of horizontal stress

A number of previous retaining wall fieldwork studies utilised push-in spade-shaped

pressure cells (commonly known as spade cells), to measure horizontal soil stresses as

they are relatively cheap, easy to install and give reasonably reproducible readings (Tedd

& Charles, 1981). They were originally designed for use in normally consolidated clays

but are now frequently used in stiff overconsolidated clays, although in such deposits it is

known that they overestimate the magnitude ofthe soil stress due to the complex localised

stresses created during installation (Tedd & Charles, 1983). Attempts have been made to

predict the amount by which spade cells over-read in a particular material by relating the

over-reading to soil strength (Tedd et ai., 1990; Ryley & Carder, 1995), but the data are

rather scattered.

3

Due to the considerable uncertainty regarding the magnitude of the over-reading and the

factors that may affect it, it is advisable to directly determine this over-reading wherever

possible until more data are available. As part of the work for this thesis a study was

carried out to evaluate the magnitude of the over-read given by a spade cell installed in

Atherfield Clay - a stiff, overconsolidated clay - which exists at the site of the CTRL at

Ashford. A spade cell was installed horizontally in the ground, aligned to measure vertical

stress. The soil above the spade cell was excavated in stages over a period of several

weeks, and at each stage the overburden was calculated and compared with the reading in

the spade cell. Additionally, measurements of in situ horizontal stress obtained using the

spade cells installed during the main instrumentation project have been compared with

self-boring pressuremeter readings. In line with other recent research, the over-read due to

the installation process has been expressed in terms of the undrained shear strength of the

soil.

1-3 Channel Tunnel Rail Link

The Channel Tunnel Rail Link (CTRL) is the largest construction project in the UK and

Britain's first new railway for over a century. It consists of 109 km of railway which starts

at the Channel Tunnel entrance at Folkestone and ends at St. Pancras station in Central

London (see Figure 1-1). It is being constructed in two sections: Section 1 consists of74

km of track running from the Channel Tunnel to North Kent. Construction began in

October 1998 and Section 1 opened in September 2003. There are two stations included in

Section 1, at Ashford in mid-Kent and Ebbsfleet in North Kent, just south of the River

Thames. Section 2 takes the CTRL under the River Thames and through London to St

Pancras, with a station in North London at Stratford. Construction of Section 2 began in

July 2001 and is due to be completed in early 2007.

A quarter of the total route is in tunnel, including 1·7 km of cut-and-cover tunnel and

propped cutting in Ashford, which forms the approach to Ashford International Station

from London.

4

1-4 Objectives

The objectives of this project are:

• To collect continuous field data, over the entire period of construction and into the

long-term, from an instrumented section of a propped in situ embedded retaining

wall in an overconsolidated clay.

• To investigate the changes in the horizontal stresses and pore water pressures in

the soil during installation of the bored piles that form the retaining wall.

• To investigate the wall movements and the development of prop loads and bending

moments associated with a retaining wall embedded in an overconsolidated clay,

and to relate these data to the horizontal stresses and pore water pressures

measured behind and in front of the monitored wall section.

• To analyse the observed performance.

• To investigate the performance of the spade-shaped pressure cells used in this

study.

This thesis describes work carried out towards the realization of these objectives, as well

as suggestions for future work.

1-5 Outline of report

Chapter 2 contains a review of retaining wall design, with references to previous

instrumentation projects, model testing and studies that used finite element techniques to

analyse the performance of retaining walls. This chapter also includes details of the use of

spade cells to measure horizontal stresses in overconsolidated deposits.

In Chapter 3 the case study is described. This chapter includes details of the local geology,

the geometry of the structure, the construction sequence and details of the instrumentation

used to monitor the performance of the structure and the stress changes in the adjacent

soil.

In Chapter 4 the structural monitoring is described in greater detail. The methods by which

the data have been calibrated and evaluated are described. The wall bending moment, wall

5

deflection and prop load data at various significant stages during construction are

presented.

In Chapter 5 the horizontal soil stress measurements are described in greater detail. The

instrumentation used to measure horizontal soil stresses and pore water pressures is

evaluated by means of a study in which the readings from a spade cell aligned to measure

vertical stress were compared with the estimated overburden acting on it (calculated from

bulk density measurements). The horizontal stress and pore water pressure data collected

during wall installation and excavation of the cutting are presented and analysed.

In Chapter 6 the long-term data collected from the spade cells and the RC props up to

November 2005, more than 4 years after construction was completed, are presented.

Chapter 7 contains conclusions arising from this research and recommendations for future

work.

6

5'

....J

StratfonL.···· .. -.... -(.$t Pancta's

Figure 1-1: Map showing the route of the Channel Tunnel Rail Link

2 LITERATURE REVIEW

2-1 Introduction

This chapter contains a general review of retaining wall design, including the effects of

wall installation on in situ horizontal soil stress. Work undertaken to study retaining wall

behaviour in similar field studies and using finite element analysis techniques is

summarised, and the sometimes differing advice provided by the most regularly used

design guidelines is described. This chapter also includes a background to the use of push­

in spade-shaped pressure cells (spade cells) to measure horizontal stresses in

overconsolidated deposits. Previous use of spade cells in similar monitoring projects and

the measures taken to determine the over-read that occurs when spade cells are used in

overconsolidated clays are described.

8

2-2 Review of investigations into retaining wall behaviour

2-2-1 Background

There are essentially two design stages required for embedded retaining wall design:

1. determination of the depth of embedment required to prevent overall collapse of

the soil/structure as a whole and

2. design of the wall member and ancillary elements such that they are strong enough

to support the consequential forces and bending moments induced in service.

A limit equilibrium analysis is usually used to calculate the depth of embedment of the

wall, and a factor of safety is applied to one or more of the variables involved to ensure

against global collapse. The structural elements are then designed to carry the bending

moments and loads which are expected as a result of the geometry determined by the limit

equilibrium analysis. Separate consideration is given to the requirements for the working

condition and the ultimate failure state, as deformations occurring under the ultimate state

may be well beyond those that are acceptable under working conditions.

The following sections contain a brief outline of retaining wall design in which the various

components of design and the choices that must be made are discussed.

2-2-2 Limiting conditions

Most modem codes of practice use limit state design principles, where limits are set, e.g. a

maximum soil loading or a maximum wall deflection, and the corresponding structural

forces are found by assuming the structure is on the verge of exceeding these criteria. As

previously mentioned stability and displacement are considered separately; stability is

considered to be the primary aim (the ultimate limit state, ULS) and displacements are

usually only considered as a secondary serviceability issue (serviceability limit state,

SLS). However, displacements can often become a ULS, for example, if the displacements

cause ground movements that affect nearby structures or services, or if they cause

unsightly conditions for the structure. In such cases the serviceability requirements often

dictate the construction method utilised, e.g. the construction sequence, the propping

methods and the type of wall. The serviceability requirements are different for each

9

project, and therefore the designer must have a full understanding of the overall project

and surrounding factors before the serviceability criteria can be defined.

The principle of analysing retaining wall behaviour by determining the forces present

when the structure is on the verge of collapse began with Coulomb (1776), who developed

an upper bound solution for retaining wall analysis by assuming a planar wedge failure

mechanism (Figure 2-1) and derived the limiting force as a function of depth behind the

wall. Coulomb's solution produces sufficiently accurate results despite the inaccurate

assumption that the wedge has a plane failure surface. More recent work in which the

lower part of the sliding wedge was assumed to have a curved boundary, as is observed in

model tests, produced equations that were considered overly complicated for practical use

(Terzaghi, 1943).

Rankine (1857) developed a lower bound solution by assuming that the earth pressures

acting on the wall held it at the limit of equilibrium. He derived limiting active and passive

earth pressure coefficients (Ka and Kp respectively), given by the ratio of effective

horizontal stress l, cr'h, to effective vertical stress, cr'v (Equation 2-1) and determined the

orientation of the surfaces of sliding. (The' at rest' or in situ earth pressure coefficient is

denoted by Ko.) The active state occurs when the rate of increase of horizontal earth

pressure is slower than for the vertical earth pressure (in absolute terms), in this case

behind a retaining wall, and the passive state occurs where the rate of increase of

horizontal earth pressure is faster than for the vertical earth pressure, e.g. in front of a

retaining walL Coulomb and Rankine's solutions are identical for a smooth wall with a

horizontal surface to the retained soil and a linear lateral pressure distribution.

, K = 0' h

0" v

Equation 2-1

Coulomb's method underestimates the active pressure and overestimates the passive

pressure, and hence may tend to an unsafe solution. Because Rankine's method relies on

static equilibrium it cannot take account of wall friction, unlike the Coulomb method, and

therefore overestimates the active pressure and underestimates the passive pressure,

producing an inherently safe (although unnecessarily costly) embedment depth. To

account for this, earth pressure coefficients have been developed which can be applied to

1 where effective stress is given by a' = a - u, where a is the total stress and u is the pore water pressure

10

K to take wall friction into consideration (British Standards Institution, 1995; Gaba el ai.,

2003).

Idealized stress distributions used in limit equilibrium analyses are shown in Figures 2-2

and 2-3. The fixed earth support method assumes that fixity develops close to the wall toe

so there is no lateral movement at this point. The free earth support method assumes that

the wall toe is free to move laterally. Unpropped embedded retaining walls rely entirely on

an adequate depth of embedment for their stability and so the fixed earth support method

must be used. Retaining walls propped at the crest tend to fail by rotation about the prop

level, so the free earth support method is usually used. It has been suggested that the fixed

earth support method can be used for analysis of propped walls: this is discussed in

Section 2-2-4.

2-2-3 Development of active and passive pressures

The limiting earth pressure coefficients for the active and passive states, Ka and Kp

respectively, are given in Equations 2-2 and 2-3, where ¢/ is the soil strength or angle of

friction and is given in degrees.2 (These equations are derived from consideration of the

Mohr circle of stress for a soil element on the verge offailure; see for example Powrie,

2004).

K = I-sin¢' a l+sin¢'

Equation 2-2

K =l+sin¢' p I-sin¢'

Equation 2-3

In order for active and passive pressures to be mobilized it is necessary for deformation or

displacement of the wall to take place. The amount of displacement necessary depends on

the type of soil and its overconsolidation ratio. 3

Soils such as sands and normally consolidated clays are close to their active limit at the in

situ state, i.e. the horizontal earth pressures are comparatively low (Ko < 0'7). In such

2 Tan ~ is analogous to the coefficient of friction, f..l, in the equation: F = f..l N where F is a force pushing against a normal force, N. 3 The overconsolidation ratio is given by: OCR = (J'y (max. preyious) / (J'y (current).

11

materials only a small movement of the structure is required to achieve full active

conditions behind the wall, but relatively large movements, which may be unacceptable

under working conditions, are required to achieve full passive conditions in front of the

wall. This was shown in large-scale model tests (Terzaghi, 1934) in which a wall was

rotated into and away from a bed of sand; the results are shown in Figure 2-4.

Overconsolidated clays (Ko > 1) have high horizontal earth pressures and are therefore

often closer to the passive limit of the soil, so only small movements of the structure are

needed to achieve passive conditions in front of the wall, but large, potentially

unacceptable movements are necessary to achieve active conditions behind the wall. Potts

& Fourie (1985) used finite element analysis to show that for an overconsolidated clay

with Ko = 2, a similar amount of movement is required to reach passive and active

conditions (see Figure 2-5). However bending moments measured in the field (Tedd e/ 0/.,

1984) and in centrifuge model tests (Bolton and Powrie, 1988) have not shown this to be

true in practice. In addition, Simpson (1992) argues that this study should be treated with

caution because it was carried out using a linear elastic soil model until it reached its

maximum shear strength, therefore not allowing for large stiffness at small strains (as

discussed by Burland et af., 1979).

Powrie et af. (1998) undertook triaxial tests on samples of overconsolidated kaolin to

investigate the total stress paths undertaken by elements of soil in front of and behind a

retaining wall during wall installation and excavation in front of the wall. They found that

the process of wall installation has a considerable influence in reducing the horizontal

stresses and therefore predicted that during excavation in front of a retaining wall, only

small movements of the wall are required to bring the earth pressure distribution close to

its active limit.

The factors that affect the extent to which passive or active pressures can be achieved are

still not well understood, particularly in the case of overconsolidated clay. In addition, the

distribution of stress on a retaining wall varies significantly from the idealised stress

distributions in Figures 2-2 and 2-3, as discussed in the following section.

2-2-4 Distribution of earth pressures

There has been considerable debate concerning the distribution of earth pressures on

retaining walls, both under working conditions and at collapse. Work by Terzaghi (1943,

12

1954), Tschebotarioff(1951) and Rowe (1952,1955 & 1956), studied the distribution of

pressure on retaining walls and the effect of wall flexibility on the bending moments and

prop/anchor loads.

As previously mentioned, some authors (e.g. Terzaghi, 1943) recommend the use of the

fixed earth support method for the design of propped walls, because the increased fixity at

the wall toe affects the distribution of pressure on the wall by raising the point of

application of the passive pressure, thereby decreasing the flexibility of the pile and hence

also the maximum bending moment. Terzaghi (1943) also suggested that for a flexible

retaining wall (particularly referring to steel sheet piles anchored near the top) and

assuming fixed earth support, there can be a further redistribution of earth pressure in

certain soils which causes an increase in pressure at the top of the wall around the position

of the restraint, and a decrease in pressure in the middle of the wall (see Figure 2-6). This

pressure distribution is a form of arching and is maintained solely by shear stresses.

Terzaghi stated that the bending moment calculated using the pressure distribution

indicated by the curved line in Figure 2-6(a) is less than half that calculated using the

linear pressure distribution.

Rowe (1952) undertook model tests with walls of varying flexibility in dense and loose

sand (Ko < 0·5) and compared the measured bending moments with those calculated from

a limit equilibrium calculation assuming free earth support and triangular pressure

distributions in front of and behind the wall. He showed that for the most flexible walls the

observed bending moments were much lower than the calculated values, and that as the

flexibility of the wall decreased the bending moment increased to that calculated by the

limit equilibrium analysis. This occurs because for a flexible wall the deflection at

excavation level is much larger than at toe level, and consequently more passive pressure

is mobilized at the top of the zone of soil in front of the wall than for a stiffer wall where e

the deflection at excavation level and the toe are similar (Figure 2-7). Rowe (1952) noted

that this effect also occurred with a rise in soil density (and hence soil stiffness).

Rowe (1952) studied the effect of a small anchor yield on the horizontal earth pressure

behind a retaining wall. He showed that an anchor yield of 0·1 % of the wall height led to a

breakdown in the arching that produces Terzaghi's pressure distribution and a return to the

triangular Coulomb pressure distribution. He concluded (and Terzaghi (1954) agreed) that

it was not justified to rely on the reductions indicated by the redistribution of pressure.

13

The debate on the use of the fixed earth support method for the analysis of propped or

anchored retaining walls has continued. Decreasing the flexibility of the pile decreases its

movement and therefore the ability of the soil to reach its active and passive limits. The

relatively old retaining wall design standard CIRIA 104 (Padfield and Mair, 1984-

discussed in Section 2-3) recommends against use of the fixed earth support method for

the design of propped walls embedded in clays, stating that it is not appropriate because of

the long-term deformation characteristic of the soil. However, the most recently published

guidance on embedded retaining wall design, CIRIA C580 (Gaba et at., 2003), states that

since modem support systems are somewhat stiffer it is now more likely that such a

pressure distribution could exist.

Since the development of active (and passive) pressures depends on the degree of wall

movement, there is some doubt that these pressures can be realised in overconsolidated

deposits with a stiff wall and propping system as little movement is likely to occur. The

work described above is, to some extent, of limited application to walls embedded in

overconsolidated deposits because it was based on the assumption that active conditions

would be met. In addition, the limit equilibrium methods used to calculate forces in

cantilever and singly propped walls can not be used for statically indeterminate multi­

propped walls; these must be designed using iterative computer aided techniques which

give consideration to soil-structure interaction.

Potts & Fourie (1985) used finite element methods to investigate the influence of Ko on

bending moments and prop loads. They first carried out an analysis with a similar

geometry and soil profile to Rowe's, which produced results which were in good

agreement. They then ran further analyses using the same geometry but with a Ko of 2, and

found that increasing Ko produced very high bending moments, greater than the limit

equilibrium value. They also found that the bending moments increased further as the

stiffness of the wall increased. Potts & Bond (1994) also used finite element analysi s to

demonstrate that the bending moment and prop loads increase as wall stiffness and Ko

increase, and that for Ko > 0'5 the maximum bending moment and prop load are greater

than the values given by a limit equilibrium analysis. However, this study did not take into

consideration the effect of wall installation on horizontal earth pressures around a

retaining wall, which is discussed in the next section.

Richards & Powrie (1998) used finite element techniques to show that the construction

sequence can have an effect on ground movements behind the retaining wall during

14

excavation. They found that top down construction with early installation of a permanent

prop at the top ofthe wall, a temporary prop at the excavation mid-height and use of a

permanent prop at formation causes minimum ground movements behind the wall. This

may affect the amount of stress relief behind the wall and therefore the soil's ability to

achieve active pressures.

2-2-5 Change in horizontal earth pressure due to wall installation effects

The installation of a bored pile wall (or a diaphragm wall) in overconsolidated clay causes

stress changes in the ground by the following process. On boring, the horizontal stress (ail)

near the hole is reduced. If no support is used, ah at the edge of the borehole is reduced to

zero and in the short-term the borehole is held open by shear stresses in the soil. If the soil

is soft the hole may collapse quickly, so either a casing or bentonite slurry is used as

support until the concrete is poured into place. In this case, ah at the edge of the borehole

reduces to the value of the pressure exerted by either the casing or the bentonite slurry.

When concrete is poured, ah at the edge of the borehole will increase to the hydrostatic

pressure exerted by the wet concrete. During the curing process the heat of hydration

causes the concrete to expand, exerting an increased pressure on the edge of the borehole.

The concrete then cools, causing shrinkage, and a further pressure change can occur as the

soil may swell due to the changes in pore water pressure.

It is well known that the stress state in the soil surrounding an in situ retaining wall is

affected by the process of wall installation (Gunn et at., 1993; Symons & Carder, 1993;

Gourvenec & Powrie, 1999) and that the stress state of the adjacent soil following wall

installation may have a significant effect on the behaviour as calculated in a numerical

analysis (Potts & Fourie, 1984; Fourie & Potts, 1989; Powrie & Li, 1991). Wall

installation will affect the recent stress history of the soil and hence its subsequent stress­

strain response (Atkinson et at., 1990; Powrie et at., 1998), and in soft soils may cause

significant ground movements in its own right (Stroud & Sweeney, 1977; Powrie &

Kantartzi, 1996).

Powrie (1985) discussed the change in horizontal pressure that might occur on

construction of a diaphragm-type retaining wall, and showed that a soil with Ko = 2 could

experience a reduction in horizontal stress that reduced Ko to just over 1. As previously

discussed in Section 2-2-3, Powrie et at. (1998) found that wall installation had a

significant influence on the horizontal soil stress, and that the recent stress history applied

15

by the excavation and concreting processes "move the stress state of the soil outside the

influence of its geological and pre-installation stress history." Furthermore Simpson and

Powrie (2001) noted that if the shear stresses that maintain a difference in the far-field and

near field horizontal stresses break down K will tend to no more than unity after

construction.

Tedd et al. (1984) reported results from instrumentation installed to monitor the behaviour

of a secant pile retaining wall during wall installation and excavation in front ofthe wall at

the Bell Common Tunnel on the M25. They found that ground movements during

installation of the retaining wall accounted for about 30% of the total movements that

occurred during the cut-and-cover tunnel construction, which indicates that stress relief

took place in the ground over this period. They also measured significant reductions in

horizontal earth pressure as a result of wall installation, although it was difficult to

establish the magnitude of these reductions due to difficulties arising from the construction

process.

During the field monitoring of other retaining walls in London Clay, Symons & Carder

(1993) measured changes in effective horizontal stress due to wall construction of 10% for

a bored pile wall (comprising 1·5 m diameter, 24 m long piles installed in London Clay)

and 20% for a diaphragm wall. Measurements of water pressure during excavation and

concreting showed that after an initial reduction, followed by an increase to levels

sometimes above pre-excavation levels, pore water pressures levelled out to values similar

to those measured prior to construction. This has also been observed in centrifuge model

tests of diaphragm wall installation processes (Powrie & Kantartzi, 1996).

Numerical analyses of in situ walls in which the effects of wall installation are neglected

generally give much higher bending moments and prop loads than both those obtained

using conventional limit equilibrium methods based on fully active pressures behind the

wall and those observed directly in the field (e.g. Potts & Fourie, 1984 and 1985; Tedd et

ai., 1984). Modelling wall installation by simply reducing the pre-excavation horizontal

stress across the entire mesh (Powrie & Li, 1991; Richards and Powrie, 1994) may reduce

the calculated wall bending moments. However, the process of wall installation in panels

or piles is three-dimensional and the magnitude, extent and longevity of the stress

reduction associated with in situ wall installation are all highly uncertain. Reducing the

horizontal earth pressure coefficient across the entire width of the finite element mesh is at

best a highly simplified approximation.

16

U sing a plane strain analysis to model the process of wall installation leads to over­

predictions of the magnitude of soil displacements and the degree and extent of horizontal

stress relief (Gunn & Clayton, 1992; De Moor, 1994; Ng et at., 1995). Therefore some

authors have tried to model wall installation in three-dimensions. Ng (1992) modelled

vertical and horizontal sections of a retaining wall in plane strain in order to investigate

the three-dimensional nature of wall installation effects. He found that horizontal arching

plays an important role in restricting ground movements, and that if this effect is ignored

ground movements will be over-predicted. He also noted that in clays with very high

stiffness at small strains, much stress relief can occur with only a small movement. Using

three-dimensional finite element analysis techniques Gourvenic & Powrie (1999) showed

that the installation of a diaphragm wall panel affects the horizontal earth pressures on

adjacent panels. They also found that soil movements increase markedly for diaphragm

panel shape ratios (the vertical dimension (depth) to the dimension in the direction along

the wall (width)) ofless than 3, i.e. soil arching is best achieved with a deep, narrow panel

(the extreme of which is a pile) rather that a short, wide one.

A better understanding of the changes in stress that occur during installation of a retaining

wall could result in more realistic estimates of prop loads, wall bending moments and

ground movements than have been possible in the past. In this section the data collected

during wall installation are described and discussed, and conclusions are made regarding

the magnitude, pattern and longevity of horizontal stress changes due to the installation of

walls of this type.

2-2-6 Behaviour of embedded walls in the longer term

There is some concern that the high horizontal stresses in an overconsolidated deposit may

become re-established in the long-term, despite the reductions that occur during retaining

wall installation and subsequent excavation in front of the wall. This is thought to occur

because the shear stresses maintaining the difference in the far field and near field

horizontal stresses may break down. As pointed out by Simpson and Powrie (2001) it

seems unlikely, particularly in the design lifetime of the wall, that if the horizontal shear

stresses break down then the vertical shear stresses necessary to produce a Ko of more than

1 will not exist either. Therefore the long-term value of K can be no more than 1.

In order to determine the long-term stress state of the soil around a retaining wall, and

hence the longevity of the reduction in stress due to wall installation, studies have been

17

undertaken on walls that have been in service for a number of years. Analysis of the long­

term behaviour of the Bell Common Tunnel (Symons & Tedd, 1989) showed insignificant

changes in earth pressure behind the wall and small decreases in earth pressure in front of

the wall in the four years after construction. Very small changes in pore water pressure

were measured on both sides of the wall.

Symons & Carder (1990) describe data obtained at two retaining wall sites: a cantilever

wall and a wall propped just below the carriageway, several years after construction was

completed. Measurements of total pressure and pore water pressure were taken in order to

determine whether equilibrium conditions had been reached. For the cantilever wall they

had, but for the propped wall pore water pressures were still gradually changing.

Measurements taken 150 m and 1·5 m from the cantilever wall showed that significant

stress reliefhad occurred due to construction/installation. However, for the propped wall,

measurements taken at 16 m and 1·5 m from the wall suggested no stress relief had

occurred in this case.

2-3 General overview of retaining wall design standards

A number of standards and guidelines regarding the design of embedded retaining walls

are in general use, and they sometimes provide conflicting advice on matters such as:

general approach to design, choice of soil parameters, choice of safety factors, estimation

of long-term pore water pressures, determination of wall bending moments and the

magnitude and distribution of horizontal earth pressure. Some of the guidance documents

currently used include:

• BS 8002, the British Standards Code of Practice for earth retaining structures

(British Standards Institution, 1994/2001);

• CIRIA 104 (the Construction Industry Research and Information Association

Report 104): Design of retaining walls embedded in stiff clay (Padfield & Mair,

1984);

• British Steel Piling Handbook (1997);

• The Highways Agency Design Manual for Roads and Bridges - Volume 2, Section

1, Part 2 - BD 42 Design of embedded retaining walls and bridge abutments

(Department of Transport, 2000);

18

• Eurocode 7. Geotechnical design, Part 1: General rules. (British Standards

Institution, 2004);

• elRIA C580, Embedded retaining walls: guidance for more economic design

(Gaba et al. 2003). This supersedes CIRIA 104.

CIRIA 104 has proved to be an extremely influential report which is used by 91 % of

design engineers (Gaba, 2002), usually in conjunction with one or more of the other more

recently released design guidelines. It was written strictly for the design of singly propped

or cantilever walls embedded in stiff overconsolidated clay, but its principles have been

applied to a wide range of wall types, including multi-propped embedded walls and non­

embedded walls. It provides design guidance for permanent and temporary walls. As part

ofthe development work for Eurocode 7, designers from countries across Europe were

asked to design a retaining wall according to their country's normal design guidelines for a

specific project. The UK design, based on CIRIA 104, produced a longer wall than most

of the others, leading to questions regarding the severity ofthe recommended values for

the factor of safety on soil strength (Fs - described later in Section 2-3-2) in CIRIA 104.

BS 8002 is the second most popular guideline after CIRIA 104, being used by 70% of

design engineers. It was originally only applicable to retaining structures with a retained

height of up to 8 m, but its latest amendment (2001) has increased this to 15 m. Some

elements of the design specification in BS 8002 are thought to be rather onerous,

particularly the unplanned excavation and surcharge requirements. These are being revised

for a forthcoming version.

The Eurocodes are a set of standards which cover both structural and geotechnical design

requirements, where Eurocode 7 is the geotechnical constituent. The codes have been

written with the aim of allowing the design process to proceed fluidly from geotechnical to

structural design without difficulty or confusion. The Eurocodes were written to be a

consistent, internationally agreed set of codes for use throughout Europe and the rest of

the World, so that legal disputes over choices made in design can be avoided.

The British Steel Piling Handbook was primarily written for the design of permanent and

temporary sheet pile walls.

BD 42 covers the design of earth retaining structures where the "main stability is provided

by a significant length of wall stem embedded in the ground" and adopts limit state

19

principles. It states how to "use ... BS 8002 and CIRIA 104 in a way which is compatible

with ... BS 5400 (British Standard for steel, concrete and composite bridges, 2000)."

CIRIA C580 has been compiled with the intention of forming a comprehensive state-of.­

the-art replacement for CIRIA 104, reducing the need for cross-referencing between

different design guidelines. It covers the design of temporary and permanent cantilever,

anchored, single and multi-propped walls which are embedded in stiff clay, and other tine

and coarse grained soils. Walls embedded in soft clay and rock are considered outside the

scope of the report.

2-3-1 Choice of soil parameters

Soil parameters for a particular site are determined from a combination of: previous

experience and publications; direct measurement of parameters from samples collected

during site investigation (usually expensive and therefore of limited number) and indirect

measurement of more easily and cheaply collected data for which correlations with the

required parameters have been suggested. The most commonly required parameters are the

undrained shear strength, cu, the effective angle of friction, <1/ and the effective cohesion,

cr. Determination of these parameters is a difficult process where the cost/output

relationship between the various sample and data collection techniques available must be

carefully balanced.

A designer should have a good understanding of the correlations between data collected

from difference sources, the factors that influence the values and the uncertainties and risk

associated with the often limited data. They can then use their 'engineeringjudgemenC to

select suitable parameters. The more recent design guidelines have attempted to advise on

this process and to make the designer think in terms of the probability of the parameter

having a certain value or range of values.

In CIRIA 104 two alternative design approaches are put forward; the designer can either

use 'moderately conservative' or 'worst credible' soil parameters (load and geometry are

also selected on these bases), with less conservative safety factors being applied for the

latter approach. 'Moderately conservative' is described as being a conservative best

estimate, and 'worst credible' is "the worst which a designer could realistically believe

might occur ... a value which is very unlikely to be exceeded". BD 42 requires the

20

consideration of moderately conservative and worst credible parameters in conjunction

with CIRIA 104.

BS 8002 requires consideration of the 'representative peak', which is generally considered

to be analogous with the moderately conservative and the critical state strength

parameters. The more onerous parameters are selected (after the relevant safety factors

have been applied) and only one design calculation is undertaken. Eurocode 7 uses

'characteristic' values for the geotechnical parameters, which are described as "a cautious

estimate ofthe value affecting the occurrence of the limit state".

The language varies, but in essence the emphasis is for a designer to use their experience

to balance low risk with economy.

2-3-2 Application of safety factors

Once the soil parameters have been given design values, a factor of safety, F, is applied to

one or more of the parameters in the ULS equation so that an adequate margin of safety is

supplied within the embedment depth. There are several different approaches to applying

this factor of safety: a factor on passive pressure, a factor on net pressure, a factor on

embedment depth andlor a factor on soil strength. Full details of how these different safety

factors compare with one another for different soil strengths are given by Burland el al.

(1981) - a summary is given below. They go on to suggest a revised definition of the

factor of safety for passive failure, Fr.

The factor on passive pressure method, Fp, (used in CP 2) employs a factor on Kp which is

commonly taken to be 2. This is based on the findings of Terzaghi (1934), as described in

Section 2-2-3, that (in sand) more wall movement is required to mobilize full passive

pressures than active pressures. Since overconsolidated clay is very likely to reach its

passive limit in front of the wall, it would be uneconomic to apply a factor of safety on

passive pressure in such a deposit.

The net pressure method, described in the British Steel Piling Handbook, uses the net

horizontal pressure distribution for walls propped at the crest. The depth of embedment is

chosen by equating the moment about the prop ofthe net pressure in front of the wall with

the moment of net pressure behind the wall factored by Fnp (normally 2). Burland el al.

(1981) showed that using this method a value of Fnp = 2 leads to a factor of safety on shear

strength of generally less than 1·1 for both drained and undrained conditions. CIRIA 104

21

expressly advises against the use of the net pressure method. Simpson and Powrie (2001)

state that the net pressure method is fundamentally unsound and potentially dangerous.

As indicated by the name, the factor on the depth of embedment, Fd, requires the

calculated wall depth to be increased (e.g. by a factor of 10%). CIRIA 104 advises that if

this method is used then because it is empirically based it should be checked against one

of the other design methods.

The use of a factor on soil strength, Fs, is consistent with the way in which factors of

safety are applied in structural design. A factor is applied to the main soil parameters: tan

¢l; the undrained shear strength, Cu; and the effective cohesion, c'; and in effect increases

the active pressure and reduces the passive pressure assumed in design. CIRIA 104

provides values for Fs. This method is thought to be the most appropriate approach to

design (Simpson & Powrie, 2000) and is the method recommended in CIRIA C580.

(CIRIA 104 additionally requires that an unplanned excavation is applied in front of the

wall, which should be the lesser of 10% of the retained height of the wall or 0·5 m, and

that a minimum surcharge of 10 kPa must be applied to the surface of the retained soil.

This produces a rather onerous loading condition that leads to uneconomical designs.)

BS 8002 focuses on serviceability requirements and in turn uses mobilization factors, M,

which are essentially analogous to Fs, but are designed to restrict displacements for the

serviceability limit state. The revised method devised by Burland et al. (1981) gives

results consistent with Fs, and so appears to give no real advantage over the use of Fs in

design.

Eurocode 7 uses limit state principles and partial factors with a design approach which is

expressed as:

The Effects of Actions must be Resisted.

In terms of a retaining wall, for example, the' actions' are the loads on the back of the

wall; the 'resistance' is the resistance provided by the ground in front of the wall and the

wall interface friction, etc.; and the 'effect' is the bending moment induced in the wall and

the prop loads. Eurocode 7 includes three design approaches which allow for partial

factors to be applied to different combinations ofthe Actions, Effects and/or Resistances.

It is expected that Design Approach 1 (DAl) will be recommended for use in Britain. In

DAI partial factors are generally applied to the primary variables, i.e. the ground strength,

22

however for piles and anchors the Resistances, e.g. bearing capacity, are factored.

Eurocode 7 also includes factors that allow for pile design from sensible interpretation of

load tests.

2-3-3 Calculation of design bending moments

There are two main methods in common practise for calculating the bending moment in a

wall in Britain. In the first, recommended in CIRIA 104, the wall length is calculated

without using safety factors, i.e. the wall embedment depth is found using a limit

equilibrium calculation with unfactored variables. The bending moment derived using this

depth of embedment is then multiplied by a safety factor of 1·4 or 1·5 for ULS design of

the wall section. In the second method, the wall length derived from using the factored

parameters for the stability calculation is used to calculate the bending moment, which is

then directly used for the ULS design. The second method often gives a higher bending

moment than the first, meaning that if the wall was ever called upon to use its full

embedment depth, its strength would not be sufficient to allow it, and it would fail in

bending. This inconsistency of length and strength means the walls are either longer than

they need to be, or not strong enough (Simpson & Powrie, 2000).

CIRIA C580 and Eurocode 7 use the same method for calculation of the SLS bending

moment as CIRIA 104, but Soil Structure Interaction (SSI) analyses may be used to

calculate the pressure on the retaining wall and therefore the redistribution of earth

pressures due to arching, etc. can be taken into account, in contrast to other methods that

use simple linear earth pressure distributions. In addition, CIRIA C580 requires that the

wall reinforcement must not be curtailed at the point of zero bending moment, but should

be extended to the bottom of the pile. Then, instead of applying a factor of safety to this

bending moment for the ULS design, the designer must use the greater of: the value

obtained from the most onerous soil parameters expected; a factor of 1'35 times the SLS

values and the values calculated in consideration of a progressive collapse.

2-3-4 Advice on wall installation effects

The most widely used guidelines for the design of embedded retaining walls (CIRIA 104,

BS 8002 and Eurocode 7) all agree that installation of a wall affects the stress state in the

soil and that these effects should be considered during design, although they do not give

23

specific guidance about what type of stress changes may occur. CIRIA 104 notes that the

installation of a diaphragm wall may reduce the horizontal stress in the soil from its

original value. BS 8002 does not give any specific guidance on the stress change that

might occur due to the installation process, it only mentions that the ratio of horizontal to

vertical stress for the soil at rest should not be used as it is affected by the installation

process.

CIRIA C580 goes further, giving guideline values for the reduction in the in situ

horizontal earth pressure coefficient due to wall installation of 10% for bored pile walls

and 20% for diaphragm walls installed in overconsolidated clays. These values are based

on the work by Symons & Carder (1993) described earlier. It is clear that without further

measurements of these stress changes in the field, considerable uncertainties remain and

economy of design may not be fully achieved. CIRIA C580 also recommends that for

linear elastic soil-structure interaction analyses a lateral earth pressure coefficient of 1 '"is

likely to give reasonably realistic bending moments and prop loads". This suggestion in

part relies on the fact that in such analyses the stiffness under unloading is generally

underestimated.

2-4 Use of spade cells to measure horizontal stress in overconsolidated clay

2-4-1 Background

Spade cells are increasingly used to measure horizontal soil stresses as they are relatively

cheap, easy to install and give reasonably reproducible readings. However, pushing a

spade cell into the ground generates high local stresses due to the displacement of the soil

as the cell is advanced. These stresses reduce over time as the excess pore water pressures

generated during the undrained loading dissipate (Tedd et ai., 1990). In natural and placed

soft clays, readings stabilize within a few days and there is considerable evidence to show

that realistic values of total stress can then be measured (Massarsch (1975); Massarsch et

ai. (1975); Tavenas et ai. (1975); Penman & Charles (1981)). The use of spade cells in

stiff clay was initially questioned, as it was thought that even if they could be pushed into

stiff clay, the soil disturbance might be so great that the stresses would be vastly over­

estimated (Tedd and Charles, 1981).

24

Tedd & Charles (1981) showed that it was possible to push spade cells into stiff London

Clay deposits, but that a period of up to one month was needed for excess pore water

pressures to dissipate. They reported that spade cells gave far more reproducible results

than self-boring pressuremeter tests, but that the recorded stress following dissipation of

excess pore water pressure was significantly greater than the true in situ value.

Tedd & Charles (1983) compared readings from spade cells installed to measure vertical

stress in stiff clay with the overburden acting on them calculated from the bulk density of

the soil, and concluded that the amount by which a spade cell will over-read is dependent

on the soil stiffness. Owing to the difficulties in determining the appropriate soil stiffness,

a correlation between the spade cell over-read and the undrained shear strength (ell) of the

soil was proposed. This was considered reasonable as the undrained shear strength of a

stiff clay is related to its stiffness, and it is a commonly determined soil parameter and

therefore widely available. A simple empirical correction of 0'5 ell was suggested,

although Tedd & Charles noted that where possible a direct site specific determination of

the spade cell over-reading in stiff clay, for example by comparison with self-boring

pressuremeter tests or calculation of overburden on a vertically aligned cell, is advisable.

In an experiment similar to that carried out by Tedd & Charles (1983), Ryley & Carder

(1995) pushed several pressure cells horizontally into London Clay to measure the vertical

soil stress at different depths, and compared the readings with the calculated overburden.

They found that a correction of 0·8 ell gave a closer fit for the cells in firm clay

(40 kN/m2 < ell < 75 kN/m2) and stiff clay (75 kN/m2 < ell < 150 kN/m2). In very stiff clays

(ell> 150 kN/m2), the cell overestimated the stress by more than 1·0 ell' They concluded

that for very stiff clays, the correction factor might be as high as 2·0 eu, and that the degree

of over-read might be affected by local changes in the soil structure such as high clay

content and plasticity. At the Bell Common Tunnel, Tedd et al. (1984) used a correction

factor of 0·5 ell for London Clay, which gave a close correlation with self-boring

pressuremeter results, but a correction of O' 3 ell gave better agreement with the self-boring

pressuremeter results in the Claygate Beds. The results from some previous studies into

the spade cell over-reading are plotted in Figure 2-8.

25

2-5 Conclusion

In this chapter the components of retaining wall design and the choices that must be made

by the designer have been discussed, including the distribution of pressure on the walL the

expected effect of wall installation and the long-term behaviour of the wall. Work

undertaken towards the further understanding of the importance and impact of these issues

has been detailed. The design advice given by the most widely used guidelines has been

discussed. It is evident that more information is required to further understand the pressure

distribution on the wall, wall installation effects and the general wall stability in terms of

the factor of safety.

A better understanding of the changes in stress that occur on installation of a retaining

wall could result in better estimates of prop loads, wall bending moments and ground

settlements than are currently possible and consequently lead to more economical designs.

The results of previous studies into the over-reading seen by spade cells installed in

overconsolidated clays are rather scattered and show that a direct determination of the

over-reading at any particular site is desirable.

26

Coulomb's idealized failure surfaces

Failure surfaces observed in model tests

Figure 2-1: Failure surfaces arising from Coulomb's idealized earth pressure distribution

pressure

Active earth pressure

Passive earth pressure

Active earth -----L....l-__ ~ ___ ~ pressure

Rotation

Figure 2-2: Fixed earth support

Passive earth pressure __ ---,,..c:;....j_

p

Active earth pressure

Figure 2-3: Free earth support

27

10~----------~--------------------'

~ ~-::s

5

~ 2 ~ a.

..s::::: t:: co 1·0 Q.)

-c Q.) 0·5 '0 E Q.) o ()

0·2

I

------ -_ ...

,. ,. /

/ I

...... ,. ......... ... ---­---

Loose sand

Passive 8

hr:>: ~::>:

0·1~--~~--~~----~------~----~

0·004 o 0·02 0·04 0·06 Wall rotation, OIh

Figure 2-4: Relationship between earth pressure and wall rotation measured by Terzaghi for normally consolidated sand (different scales for active and passive

pressure) (modified by Simpson, 1992)

~

~ ::s If) If) Q.) .... a.

..s::::: t:: co Q.) -0 -c Q.) '0 tE Q.) 0 ()

4r-------------------r-----------------~

3

2

1

8 Active

\ ::::::::::::: \ :::::::::::::: m····· .. ········· ......... Passive

8

hl0 ...

o~--~----~--~----~--~----~--~--~ 0·04 0·02 o 0·02 0·04

Wall rotation, 5/h

Figure 2-5: Relationship between earth pressure and wall rotation computed by Potts and Fourie (1986) for overconsolidated clay

28

(a) Coulomb distribution - rotation

near base of wall

(b) Rotation at top of wall - (c) Fixed earth support -reduction in pressure at reduction in pressure towards

bottom of retained section top of retained section

Wall ,L~te.ral ~ressure movement ~ distribution

~ \/ , , I I I I I

Wall ---movement

, I I , I I , ~ movement I

\ '" Lateral pressure I distribution

Lateral pressure distribution

Figure 2-6: Influence of movement type on pressure distribution

Prop or anchor --t~---.:~~-- Prop or anchor _ ... ~-~~--

ctive

Factored passive

(a) Stiff wall

Factored passive

Full passive

(b) Flexible wall

Figure 2-7: Stress distributions behind and in front of (a) stiff and (b) flexible embedded walls (after Rowe, 1952)

29

A Reading, SBP, flat dilatometer (Carder and Symons, 1989) + Brent, SBP, Oedometer o Bell Common Tunnel, SBP (Tedd et ai, 1984) • Balham, Overburden (Tedd and Charles, 1983) • Heathrow, Overburden (Ryley and Carder, 1995)

500 .'

..... 1.5 C u

:

: ... • .. ' .

. ...... 0.8 C u

~ ... . '

.... ....... ·········0.3 C u

..•.....

100 200 300 Undrained shear strength, c u, kPa

Figure 2-8: Over-reading v CU. After Ryley and Carder, 1995.

30

3 CASESTUDY

3-1 Introduction

This chapter describes the geology of the field study site and details of the geometry and

construction sequence of the monitored retaining wall. It includes details of the

geotechnical design parameters obtained from the original site investigation for the soils at

the field study site together with an outline of the instrumentation scheme implemented to

measure loads within the structure and the stress changes within the ground adjacent to the

retaining wall.

3-2 Geology

Two specific locations at the site are referred to in this thesis and the variation in the

geology at these locations is detailed in this section. The locations are: a section of

retaining wall that forms part of a propped cutting, which has been comprehensively

instrumented as part of this project and is termed the 'instrumented section' and a drainage

31

sump next to the cut-and-cover tunnel, approximately 600 m to the north-west of the

instrumented section, termed the 'nadir sump'. An assessment ofthe performance of the

spade cells used in this study was undertaken at the nadir sump and is detailed in Chapter

5. The location of a self-boring pressuremeter (SBPM) test, refened to within this work

because it exists in an area with similar geological boundaries to the instrumented section.

is also noted (the relative positions of these locations are shown in Figure 3-1).

3-2-1 Geotechnical data and sample collection

Preliminary information regarding the nature of the geology at the CTRL site at Ashfc)]"(.i

and its geotechnical parameters was collected from: borehole records and laboratory tests

obtained as part of the main CTRL site investigation (the locations of useful boreholes in

the vicinity of the instrumented section and nadir sump are shown in Figure 3-1); the

CTRL Central Ashford Geotechnical Design Basis Report (GDBR - Union Railways,

1997); the regional geology guide for the district (HMSO, 1978); the Geological Memoir

(Smart et al., 1966); the 1 :50 000 British Geological Survey map for the area and from

Humpage & Booth (2000).

A characterization ofthe geology at the instrumented section was carried out to gain a

preliminary assessment of the nature and extent of the geological strata. Samples of

material were taken from within the cutting during excavation. Block sampling techniques

described by Heyman & Clayton (1999) were used to obtain undisturbed samples at the

instrumented section at levels determined by the characterization. A series of steps were

cut in the material with an excavation machine and blocks of approximately 0·3 x 0·3 x

0·3 m were then cut from the steps by hand. The blocks were wrapped in Clingfilm and

foil to preserve the natural moisture content.

In order to obtain further high quality soil samples and to characterise the geological

sequence, two wireline borehole cores were taken about 25 m from the instrumented

section in July 2001 (labelled BH1 and BH2 in Figure 3-1). Photos of the wireline

borehole equipment and logs of these boreholes are included in Appendix A. Samples

taken from these cores and from the block samples were tested in the laboratory at the

University of Southampton to determine the small strain stiffness behaviour of the clay

(Xu, 2005).

32

At the nadir sump the geological profile was estimated from boreholes taken during the

main site investigation and confirmed with observations made during excavation to form

the cutting at the instrumented section. Samples were obtained during excavation of the

nadir sump to provide additional bulk density data to compliment that reported in the main

CTRL site investigation. The geological profile at the location of the SBPM test was

obtained from adjacent borehole records (borehole reference PR3593, see Figure 3-1 ).

In Sections 3-2-3 and 3-2-4 descriptions ofthe reported and observed geology determined

from these combined sources is described.

3-2-2 Site location and history

The monitoring is located in the south-east of England at Ashford, Kent. The topography

of southern Kent, Sussex and part of Surrey is controlled by an anticline which can be

seen on the early 19th century geological map shown in Figure 3-2 (Mantell, 1833). The

stratum which overlies the Weald Clay, shown in Figure 3-2 as the Shanklin Sands, is now

known as the Lower Greensand. The Channel Tunnel Rail Link (CTRL) at Ashford has

been constructed along the side of the Weald of Kent valley on the basal layers of the

Lower Greensand,just above its boundary with the Weald Clay, as shown in Figure 3-2.

Construction of the CTRL included excavations which pass through the Lower Greensand

into the Weald Clay. The Weald of Kent valley was created by the removal of the upper

layers of material by denudation caused by the weakening of the material due to the

folding process which produced the anticline. The removal of the upper strata results in

the lower existing strata being overconsolidated, i.e. they were previously subjected to

much higher vertical loading than they are at present. This process may have also caused a

reduction in the effective horizontal stress in these deposits - this is discussed in more

detail in Section 3-2-4.

Previous land use

The 1876 1:10560 map of Ashford shows that the site of the instrumented section was a

(plant) nursery. A benchmark on this map indicates that the ground level close to the site

of the instrumented wall was 40·8 m AOD (metres above ordnance datum). Before

construction of the CTRL the ground level at the site was about 44 m AOD. Several

metres of made ground have been placed at the site and were observed during excavation

33

and in nearby boreholes. The area was also used as allotments over the years and more

recently has been the site of several warehouses.

3-2-3 Reported geology

Table 3-1 details the geological succession of the relevant deposits, and their approximate

ages.

Period and Succession

Age (Millions of Stratotype Thickness, m

Epoch years)

Lower ~1l2 Hythe Beds 25-90

Early Greensand 124 Atherfield Clay 6-28

Cretaceous Wealden up to 120 in mid

135 Weald Clay Formation Kent region

Table 3-1: Geological succession

The Weald Clay is generally a freshwater deposit, although the upper region is believed to

have been laid in brackish waters. It is generally a stiff to very stiff, brown to grey clay,

and in places it is thinly bedded with silt.

The overlying Lower Greensand was deposited in marine conditions. The basal layer of

the Lower Greensand is the Atherfield Clay, a deposit generally up to 15 m thick in the

Ashford region. The boundary between the Weald Clay and Atherfield Clay is sharply

defined by an undulose erosion surface which represents a slight unconformity, however

the bottom metre of the lower Atherfield Clay is sometimes confused for the top of the

Weald Clay. This band of clay differs distinctly from the underlying Weald Clay and often

contains a number oflarge oysters (Aetostreon or Exogyra) in the lowest 0·5 m, forming a

useful marker horizon (Roberts, 2003). The boundary is also often stained light brown.

The Atherfield Clay is a stiff to very stiff, closely fissured fairly fossiliferous clay and

consists of two distinct materials. The lower Atherfield Clay includes reddish brown or

chocolaty-brown clay. The upper layer (upper Atherfield Clay) is closely to extremely

closely bedded, greyish blue to brown, sandy and about 8 m thick. Humpage and Booth

(2000) reported that the Atherfield Clay weathers to a brown mottled orange and red silty

clay.

34

The GDBR notes that the Atherfield Clay has an "occasionally brecciated I structure"

within its high plasticity zone. The boundary between the brecciated and non-brecciated

clay is difficult to determine and for the purposes of design was taken to be the high­

intermediate plasticity boundary. The GDBR also notes that there has been movement of

the Atherfield Clay: a shear zone was found in a borehole about 750 m from the

instrumented section, and cross-sections drawn through boreholes (using the high­

intermediate plasticity boundary in the Atherfield Clay as a marker) show evidence of

dislocation in the area of the instrumented section.

The Hythe Beds overlying the Atherfield Clay outcrop over much of the CTRL Ashford

site and are approximately 1·5 m thick at the instrumented section. In general the Hythe

Beds consist of alternating layers of Rags tone, a hard, greyish blue glauconitic sandy

limestone and Hassock, a grey to brownish grey glauconitic, argillaceous, calcareous sand

or soft sandstone. The junction between Hythe Beds and the Atherfield Clay is not distinct

and as a result the top of the Atherfield Clay is sandy and glauconitic (HMSO, 1978). The

Hythe Beds are generally recorded as having a basal clayey sandy layer. Springs issue at

the junction of the Hythe Beds and the Atherfield Clay, hence at the Atherfield Clay

outcrop the land is commonly wet and boggy and it is not possible to get an impression of

the in situ state of the clay from its outcrops. In addition, the springs cause instability and

there have been movements of large blocks of limestone and sand, which have slipped on

the Atherfield Clay. The construction of the CTRL has provided an opportunity for the

Atherfield and Weald Clays to be viewed in greater detail than has previously been

possible.

From a study which included the Hythe Beds outcrop in the Weald of Kent near

Sevenoaks, Skempton & Weeks (1976) noted that the boundary between the Hythe Beds

and Atherfield Clay was difficult to determine, and that there is "probably a consistent

error in plotting the base of the Hythe Beds".

I Breccia: Geo!. A composite rock consisting of angular fragments of stone, etc., cemented together by some matrix, such as lime: sometimes opposed to conglomerate, in which the fragments are rounded and waterworn. (source: Oxford English Dictionary)

35

3-2-4 Observed geology

Weald Clay

The upper Weald Clay contains many silt laminations and occasional bands of siltstone,

about 100-200 mm thick, were recorded (see Appendix A). In some cases the silt

laminations were a few millimetres thick and there was proportionally more silt than clay

in the recovered core. These laminations generally dipped at approximately 20°.

Figure 3-3 shows the Weald Clay exposed by excavation at another part of the CTRL.

Layers in the Weald Clay can be clearly seen and are similar to those observed in the

wireline borehole core taken at the instrumented section.

Atherfield Clay

The erosion surface reported to exist between the Atherfield and Weald Clay was

observed in the cores taken close to the instrumented section (see Appendix A). The

surface was stained brown and about 2 mm of crumbly organic material existed between

the Atherfield and Weald clays.

The lower Atherfield Clay was chocolate-brown and very stiff. It is delineated by a

distinctive 400-500 mm thick band oflight brown material at the top, which was observed

in the nadir sump (Figure 3-4).

In the upper Atherfield Clay zones of thin silt partings, which become more frequent

towards the bottom of the stratum, were recorded. At the top of the upper Atherfield Clay

stratum the material consists of peds2 surrounded by a softer matrix. This is evidence of

the brecciation described in the GDBR (see Section 3-2-3). It also indicates weathering

(grade III - described as "stiffer, less weathered lumps of clay (lithorelics) in a matrix of

disturbed, softer clay" by Chandler, 2000). The size of the peds reduces with depth and the

soil becomes more consistent. Figure 3-5 shows a block sample of the upper Atherfield

Clay prior to removal. On cutting the material very hard peds up to 100 mm wide, which

can be seen in Figure 3-5, fell away. These peds are very hard and impossible to cut by

hand with field instruments and the action of attempting to cut the soil caused significant

disturbance and damage to the samples. The peds are surrounded by softer crumbly

2 Ped: an individual natural soil aggregate (US. Department of Agriculture Yearbook, 1958).

36

material, to which some smearing caused by the cutting process is evident in Figure 3-5.

This observation is consistent with the description of the Atherfield Clay near Ashford

given by Smart et al. (1966) as a, "grey clay that crumbles into cubical lumps" . These

lumps are caused by the opening up of fissures on unloading of the soil- a process likely

to provide drainage paths, thereby allowing easier movement of water than the measured

in situ permeability of the material might indicate.

Hythe Beds

The Hythe Beds was approximately 1·5 m thick at the instrumented section and consists of

stiff to very stiff yellow/grey mottled sandy clay (apart from the colour, it is very similar

in appearance to the top of the upper Atherfield Clay). At the transition between the

yellow/grey mottled material and the grey material there are peds of grey material within a

yellow matrix. Samples of the Hythe Beds taken during excavation were observed to

consist of firm to stiff mottled light brown and light grey sandy clay with pottery tile,

rotten wood fragments and large dark brown rootlets (Figure 3-6). Block sampling within

the Hythe Beds was found to be relatively straight forward, as the clay cut easily due to

the presence of fine sand, and the samples remained intact.

For the purposes of this work the boundary between the Hythe Beds and the Atherfield

Clay has been taken to exist where the colour of the clay changes from mottled

yellow/grey to grey, which coincides with a small increase in the plasticity index (due to a

reduction in the quantity of sand in the deposit). On this basis the upper Atherfield Clay

layer is 8·75 m thick at the instrumented section, which is generally consistent with

measurements from boreholes across the site.

Above the Hythe Beds is about 2 m of made ground, which is firm mottled dark

brown/grey organic sandy clay with pottery tile, rootlets and rotten wood fragments.

Figures 3-7 and 3-8 show the dark brown made ground, the yellow/light grey Hythe Beds

and the dark grey upper Atherfield Clay during excavation of the cutting.

General

There is evidence to suggest that there has been historic movement of the upper regions of

the Atherfield and possibly the Weald Clay in the area of the instrumented section. Figure

3-9 shows a shear plane observed in the excavation for the cutting 50 m to the west of the

instrumented section. The shear plane dips at 45° to the south-southwest with a strike of

37

approximately 0700 (east-northeast). Other shear planes have been observed within the

Atherfield Clay and Weald Clays formations. Figure 3-10 shows a sketch of a

discontinuity (possibly a shear plane) observed in the base of the excavation for the cutting

at the instrumented section before blinding was laid for the base slab (see Sections 3-4 and

3-5 for details of the construction and instrumentation layout). There was an apparent error

of approximately 1·5 m in the logging of the two wireline boreholes taken 25 m behind the

north wall at the instrumented section. The observed shear plane suggests that rather than

this being an actual error, this discrepancy may be due to disturbance/faulting between the

two borehole locations.

3-3 Geotechnical properties

3-3-1 Plasticity

Due to the variable silt content of the Weald Clay its plasticity index ranges between 10-

30%. The upper Atherfield Clay is generally of relatively high plasticity, with a plasticity

index of approximately 50%. The lower Atherfield Clay has a plasticity index of20-30%.

Profiles of the soil found from the wireline boreholes taken 25 m from the instrumented

section, including liquid limit, plastic limit and moisture content data, are shown in Figure

3-11. The moisture content is at or close to the plastic limit and there is a clear change in

plasticity from intermediate to low plasticity within the upper Atherfield Clay stratum,

about 2·3 m above the level ofthe boundary with the underlying lower Atherfield Clay.

This is explained by the increase in silt zones in the lower part of the upper Atherfield

Clay stratum.

A comparison of the geology at the instrumented section, the nadir sump and the location

of the pressuremeter test is shown in Figure 3-12. At the nadir sump the ground level was

reduced by approximately 8·2 m before construction.

3-3-2 Bulk density

Bulk density measurements from the main CTRL site investigation and samples obtained

during excavation ofthe nadir sump are shown (plotted in m AOD) in Figure 3-13.

38

3-3-3 Soil strength

The moderately conservative (see Section 2-3-1) value for the effective angle of soil

friction, rjJ', recommended in the GDBR for the Weald Clay is 23° with the effective

cohesion, c', equal to O. Further interpretation of the presented data indicates a rjJ' of 240

with c' = 0 may give a more appropriate 'average' value.

The moderately conservative rjJ' value recommended in the GDBR for the intermediate

plasticity Atherfield Clay is 21 ° with c' = 4 kPa. However, further interpretation of the

presented data indicates a value of 24° with c' = 0 may be a more appropriate 'average'

value. In addition, triaxial tests carried out at the University of Southampton by Xu (2005)

on samples of the lower Atherfield Clay collected from the rotary drilled cores (described

in Section 3-2-1) indicated a rjJ' for the lower Atherfield Clay of 26° with c' = 10 kPa

(Figure 3-14).

For the high plasticity Atherfield Clay the GDBR generally recommends a moderately

conservative rjJ' value of 24° with c' = 10 kPa for the non-brecciated and 29° with c' = 0 for

the brecciated region in the central Ashford area. However, the report notes that the

fissuring observed in the Atherfield Clay in the localised area of the instrumented section

is indicative of land movements (described in Sections 3-2-3 and 3-2-4) and therefore a

value relating to the remoulded state of the clay was used for design: ¢Jerit = 18°. This

value was later increased to 20° on reinterpretation of the data (Union Railways Ltd, 1997

B).

In the GDBR the value taken for the Hythe Beds as the moderately conservative soil

strength, ¢J, was 30° and c' = O. The limestone and soft sandstone that the Hythe Beds

generally consist of (see Section 3-2-3) would be responsible for much of this strength, but

were not present at the instrumented section: the lowest part of the Hythe Beds consists

only of sandy clay.

These soil strength parameters are summarised in Table 3-2.

39

Effective angle of soil Effective friction, til (moderately cohesion,

conservative),O c', kPa

Weald Clay 23 0

lower Atherfield Clay and GDBR

21 4 intermediate plasticity 24 0 upper Atherfield Clay

Xu (2005) 26 10

high plasticity upper non-brecciated 24 10 Atherfield Clay brecciated 29 0

remoulded 18/20 0

Hythe Beds 30 0

Table 3-2: Soil parameters of the Atherfield Clay and Weald Clay

Figure 3-15 shows the undrained shear strength (cu) as a function of depth, determined

from unconsolidated undrained triaxial tests on 100 mm samples and estimated from

standard penetration test (SPT) results. The SPT blowcount N multiplied by 4·5 (Stroud,

1974) correlates closely with the laboratory data for the Atherfield Clay, but gives values

generally in excess of the laboratory tests for the Weald Clay. This is consistent with

known sampling effects and in particular the disturbance to the microstructure, fabric and

mean effective stress caused by thick walled tube sampling in stiff clays (e.g. Hight &

Leroueil, 2003). In the Ashford site's Geotechnical Design Basis Report (Union Railways

Ltd, 1997) a value of 4·5 was used to factor the SPT data for the Atherfield clay and 3·5

was used for the Weald Clay. The best-fit line to all the data in the Atherfield Clay is

indicated, and is given by Equation 3-1. Figure 3-16 shows the undrained shear strength

found from the SBPM test PR3593 (for the location see Figure 3-1). The SBPM data

indicate slightly higher undrained shear strengths compared with the laboratory tests and

SPTs.

Cu (kPa) = 22 + 7 z (z in m below ground level) Equation 3-1

3-3-4 In situ horizontal stress

Figure 3-17 shows the in situ horizontal stress interpreted from the SBPM test. These data

imply an in situ earth pressure coefficient of approximately 1 (although they are

measurements of total rather than effective stress). This is rather less than would be

expected on the basis of the one-dimensional stress history of the deposit. Movement of

40

the soil by faulting or a landslide (described in Section 3-2-4) may reduce the in silll

horizontal earth pressure coefficient, Ko, to below that which might otherwise be expected

for a stiff overconsolidated clay on the basis of its stress history.

As previously mentioned, the geological folding that created the anticline shown in Figure

3-2 may also have had the effect of releasing horizontal stress as the soil stretched and

weakened. The London Clay (a much younger deposit than the Atherfield and Weald

clays), also seen on the section shown in Figure 3-2, was folded inwards by this process

which may have contributed to its relatively high in situ horizontal stress.

3-3-5 Permeability and Groundwater

Data from variable head permeability tests taken during the main site investigation are

shown in Figure 3-18. Due to the paucity of the data no firm conclusions can be drawn,

however, it appears that the Atherfield Clay has a very low permeability of about

1 x 10-8 mis, the Hythe Beds have a slightly higher permeability of about 1 x 10-7 mls and

the Weald Clay has a variable permeability of between 4 x 10-9 mls and 2 x 10-6 mls (the

variation is due to the silt laminations).

Roberts et al. (in print) state that reliable results from 5 tests in the Atherfield Clay and 20

in the Weald Clay carried out in the Central Ashford area gave an average permeability for

the Atherfield Clay of3 x 10-8 mls (with a range of2 x 10-9 mls to 9 x 10-8 m/s) and

1 x 10-8 mls for the Weald Clay (with a range of 1 x 10-9 mls to 3 x 10-7 m/s). Roberts et

at. state however that borehole permeability tests are known to be unreliable, often

underestimating the true permeability by an order of magnitude or more. They go on to

report the results of a pumping tests using ejector wells which indicated the permeability

of the Weald Clay to be in the order of 1 to 2 x 10-6 m/s.

The GDBR indicates that the in situ groundwater in the area of the instrumentation is

approximately 1 m below ground level. Extension of the best-fit line to the pore water

pressure data in the strata above and including the upper Atherfield Clay (Figure 3-19)

indicates the groundwater level to be at 42·7 m AOD (ground level is at 43·7 m AOD),

concurring with the GDBR. The water pressure distribution in these strata is represented

by Equation 3-2, where u is the pore water pressure and L is level in m AOD.

41

u = 8·28 x (42·706 - L), for L < 42·706. Equation 3-2

This equation indicates under-drainage of the Atherfield Clay (if the equation represented

a hydrostatic distribution of water pressure, the multiplier value 8·28 would be 9·81, the

density of water). This is probably due to the horizontal silt laminations in the underlying

Weald Clay acting as drainage paths, and the dewatering in a section of the CTRL

approximately 100 m from the instrumented section was shown to have an influence on

the groundwater regime up to 500 m away (Roberts et ai., in print). This dewatering

occurred over the period of wall installation at the instrumented section.

The data collected at the dewatered section approximately 100 m away from the

instrumented section indicates that the lower Atherfield and Weald clays have been

subjected to a drawdown of 5 m. This is consistent with the lower pore water pressures

measured in the lower Atherfield Clay. The water pressure profile produced by this

drawdown is shown in Figure 3-19, and given in Equation 3-3. The dotted line indicates

the likely water pressure profile in the lower Atherfield Clay.

u = 9·81 x (37·709 - L), for L < 37·709. Equation 3-3

3-4 Description of site and geometry of structure at the instrumented section

The instrumented section of the Channel Tunnel Rail Link (CTRL) is approximately

350 m northwest of Ashford International Station in Kent (Figure 3-20). It consists of a

propped contiguous bored pile retaining wall formed from 21 m long, 1·05 m diameter

piles spaced at 1·35 m centres. The piles at the instrumented section were installed in

November and December 1999. Figure 3-21 shows a cross-section of the instrumented

section. The cutting width is approximately 12 m and the excavation depth 10 m. The

walls are permanently supported at crest level by 1 m square reinforced concrete props

spaced at 4·5 m centres and at formation level by a reinforced concrete base slab. A corbel

was added to the retaining wall above the base slab at a later date to prevent slab uplift.

Temporary tubular steel props were employed during excavation to support the walls until

the base slab had been constructed.

In the Ashford area the CTRL follows a contour along a valley side, with the ground

sloping down from the northeast to the southwest. The construction is therefore not

symmetrical about the centreline of the excavation. Behind the south wall lies the London

42

to Dover main railway line and other local railway lines. The ground level behind the

south wall was reduced by several extra metres during construction of the CTRL to enable

construction of another railway line, as shown in Figure 3-22. Unfortunately access to the

south wall was not available due to site constraints and so it was not possible to monitor

its movement.

3-5 Construction sequence and installation of instrumentation

Approximately one month prior to installation of the contiguous piles, sixteen Soil

Instruments vibrating-wire push-in pressure cells with integral vibrating-wire piezometers

(spade cells) were installed adjacent to the line of the retaining wall. The spade cells

measured the in situ total horizontal stresses and pore water pressures and the subsequent

changes due to bored pile installation and further construction events. Figures 3-23 and

3-24 show the general arrangement of the spade cells in relation to the retaining wall. In

determining the positions of the spade cells, careful consideration was given to the

proximity of the individual bored piles allowing for pile installation tolerances (more

relevant to the deeper instruments) and the localised effect of pile installation on the

surrounding soil (Richards et al., acceptedfor publication). The spade cells were installed

at three different distances from the line of the wall, so that any variation in the influence

of wall installation effects with distance from the wall could be determined. The array of

instruments was spread out laterally along the wall owing to the need for a separate

installation borehole for each spade cell. The spade cell cables were sheathed in brightly

coloured sturdy plastic tubing near ground level behind the wall and to the depth of

formation level in front of the wall, in order to protect the cables and to provide a visual

marker during construction and excavation activities.

The dates ofthe main construction events are shown in Table 3-3, and details of the main

construction events are outlined below. For the purposes of this thesis all graphs and data

are referenced with respect to the number of days from the beginning of the project, with

the day the first spade cell was installed, 8th October 1999, being Day 1. The last

significant construction activity at the instrumented section occurred when the last

instrumented temporary prop was removed on Day 595.

43

Stage Name Schematic Day Date

1 Spade Cell see Figures 3- 1-13 8th_20th October 1999 Installation 23 and 3-24

2 Pile Installation 47-71 23 rd November to 1 i h

December 1999

3 Sand Drain i'W'~'" 349-352 20th_23 fd September 2000

Installation

d/~" .....

4 Capping Beam .=. .=. 440-442 approximately 20_22nd

iW /

A

'" Construction December 2000

.• '1'."".

5 RC (reinforced ~ 465 Props 1 & 2: 14th January concrete) Prop I'W'A' 467 2001 Construction Prop 3: 16th January 2001

u,~ ......

6 Excavation Phase 1 483-509 1 st_2ih February 2001 t 5:4 (no work ih_21 st Feb

~~ inclusive)

.• ,1:,,,,,.

7 Temporary Prop 512 Prop 1: 2nd March 2001 Installation

i'W'A' 522 Props 2 & 3: 1 i h March 2001

.• '1'.""'.

8 Excavation Phase 2 530-537 20th_2ih March 2001

13-7

I"w~ ",~,,,,,,,

9 Base Slab 579 8th May 2001 Construction

i~ ··'f,,,,,,"

10 Temporary Prop 581 Prop 1: 1 Oth May 2001 Removal 595 Props 2 & 3: 24th May 2001

I"~ ... ,r., ......

Table 3-3: Dates of main construction events

44

An essential component of this project was the full-time presence of the researcher on site

throughout the main construction period. This allowed the researcher to compile a detailed

diary of the construction events (and any other activities which may have affected the soil

stress or structural loads) and to link these changes with the collected measurements.

3-5-1 Pile installation

The lettering on the piles in Figure 3-23 indicates the sequence in which the piles were

installed and is detailed in Table 3-4. Appendix B contains more detailed information

regarding the pile installation period.

The pile installation process is shown in Figure 3-25 and illustrated in Figure 3-26, 3-27

and 3-28. In most cases, the uppermost 8 m ofthe pile bore was excavated on one working

day following which a casing was inserted to support the sides (this process took between

20 and 80 minutes, with an average duration of27 minutes). The remainder of the bore

was excavated on the next working day without any further form of support, although on

reaching the required depth bentonite slurry was introduced up to original ground level

(this took between 20 and 50 minutes, with an average of30 minutes). The reinforcement

cage was then lowered into the borehole and concrete was tremied in from the base at the

same rate that the bentonite was removed. (This process was implemented after

observation of trial boreholes left open for several days showed that those in the Atherfield

Clay would remain stable for at least 24 hours whereas deterioration occurred more

quickly in Weald Clay; Roscoe and Twine, 2001). The exception to this installation

sequence was a single pile in period Dl (see Figure 3-23 and Table 3-4) for which

excavation and concreting were carried out within one afternoon, immediately after period

D. It is also of note that certain piles were installed on a Friday (pm) and Monday (am), so

following excavation to 8 m the hole was left cased and open over the weekend.

45

.f::.. 0\

Installation Boring started Casing depth Boring restarted Pile completed, Concreting Concreting period reached, casing bentonite added started ended

inserted

Day Time Time Day Time Time Time Time

A 47 1510-1550 1540-1620 48 1155-1255 1235-1330 1420-1550 1453-1638

B 49 1440-1720 1600-1740 50 0805-0916 0845-1001 0916-1225 1002-1311

C 53 1459-1625 1520-1646 54 0800-0934 0835-1000 0955-1227 1032-1315

D 55 1518 1542 56 0910 1000 1109 1150

Dl 56 1220 1255 56 1255 1350 1626 1724

E 57 1150-1435 1218-1500 60 0906-1310 0941-1335 1112-1540 1156-1627

F 64 1040-1108 1130-1200 67 0810-0842 0842-0905 0950-1114 1036-1201

G 67 1715 1746 68 0848 0908 1058 1145

H 69 1631-1659 1657-1721 70 0810-0841 0840-0905 0933-1040 1006-1117

I 71 1135-1203 1200-1225 74 1120-1148 1146-1210 1430-1605 1545-1705 -- -- -

Table 3-4: Sequence of pile installation. Also see Figure 3-20

In each installation period, between 1 and 4 piles were installed. The times given represent the range including all piles

within each period

An inclinometer tube was installed within pile Z (see Figure 3-23) to measure the

deflected profile of the wall and to allow the bending moments to be estimated from the

deformed wall profile. (An inclinometer tube was also installed in the pile nearest to spade

cell 12 but the tube was damaged during construction and was unusable.) The inclinometer

tubing extended 10m into the natural ground below the toe of the pile thereby establishing

a fixed point and eliminating the need to rely on surveying the top of the wall to determine

its overall movement. Bending moments were also calculated from strains measured using

vibrating-wire embedment gauges installed in piles X and Y (Figure 3-23). Pairs of gauges

were situated at 1·5 m intervals down the piles (see Figures 3-29 and 3-30).

3-5-2 Sand drains

Sand drains were installed between the cutting walls to reduce the pore water pressures in

front of the retaining wall during construction. The retaining wall design was based on the

assumption that hydrostatic pore water pressures would exist on both sides of the wall:

from ground level behind the wall and from the bottom of the excavation in front of the

wall. However, there were concerns that the flow of water through the wall (or around the

toe) could result in higher pore water pressures in front of the wall. This could reduce the

passive effective stresses and therefore overstress the wall, or, depending on the

permeability of the Atherfield Clay, cause pore water pressures to build up in the Weald

Clay underneath the Atherfield Clay (due to the relatively high horizontal to vertical

permeability of the Weald Clay) which may cause the 'plug' of Atherfield Clay above to

be pushed upwards. The sand drains were therefore designed to relieve any pore water

pressure build up in front of the wall and ensure that the hydrostatic pore water pressure

profile assumed in design was achieved (Roscoe, 2005).

The 30 m deep sand drains were installed at 3 m ± 0·5 m centres in two lines, 3 m from

each wall. The sand drains in the vicinity of the instrumented section were installed over a

period of about 3 days. Each drain comprised a 150 mm diameter hole (bored without

casing) filled with 10 mm gravel. The line of sand drains near to the north wall fell

between most of the spade cells in front of the retaining wall, shown in Figure 3-23. A

0·5 m leeway in the position of the sand drains was used to ensure that spade cell 17 and

its cable were not endangered by the sand drain installation. Figure 3-31 shows the sand

drain nearest to spade cell 11 during excavation.

47

After excavation of the cutting to fonnation level a shallow channel (less than 200 nl111

deep) was excavated and filled with 10 mm gravel to link the sand drains under the base

slab. Drain holes were installed in the base slab so that water could drain through it,

thereby preventing the build up of pore water pressure under the base slab.

Data for the period immediately following sand drain installation is limited as construction

activities had commenced within the instrumented wall section and during this period

some of the gauge wires were broken; these activities included construction of a large slab

behind the wall and excavation around the wall for the capping beam construction (see

Section 3-5-7). Total horizontal stress and pore water pressure data collected prior to

construction of the capping beam are shown in Figure 3-32 (total stress measurements are

corrected by 0·35 eu to allow for spade cell installation effects: see Chapter 5). The figure

shows that installation of these drains appears to have had little effect on pore water

pressures and total stresses in the short-tenn.

3-5-3 Removal of pile tops

A month after installation ofthe sand drains (Days 382 to 384) the soil on either side of

the piles was excavated to a depth of approximately 2 m (Figure 3-33) in order that the

piles could be broken down to competent concrete, the pile tops removed and the capping

beam constructed. The pile tops were removed using the 'Elliot' method. This involves

drilling holes through the pile at the pile cut-off point and inducing a crack across the pile

section. Removal of the concrete pile top is aided by a foam covering on the reinforcement

bars above the cut-off point which restricts bonding between the reinforcement and the

concrete (Figure 3-34). The pile tops in the vicinity ofthe instrumentation were removed

in the days around Day 413 (23rd November 2000) (see Figure 3-35).

During removal of the pile tops it was necessary to release the vibrating-wire strain gauge

cables from the concrete. The cables were installed in a duct attached to the pile

reinforcement and extending down to the level of the pile cut-off (about 1 m) before the

pile concrete was poured, so that when the concrete was 'cut-off the cables were carefully

drawn through this ducting so that they exited the pile at the cut-off level. The pile gauge

cables were then run through ducting within the blinding for the capping beam, along with

the cables from the spade cells installed in front of the wall.

48

3-5-4 Capping beam

The capping beam had the dimensions shown in Figure 3-36 and can be seen in Figure

3-37. Around the time that the capping beam was constructed the ground level between

the tunnel walls was reduced by up to 1 m in preparation for the construction of the

reinforced concrete props. A 2 m high wall was built on the capping beam approximately

45 days after Temporary Prop Removal.

3-5-5 Reinforced concrete props

The reinforced concrete (RC) props were constructed on blinding approximately 75-100

mm thick. Plastic liner was used to separate the blinding form the poured concrete.

Blackjack was pained onto the capping beam at the locations of the forthcoming RC

props, as shown in Figure 3-38. There was no steel connection between the capping beam

and the RC prop. Shuttering was then erected, the pre-fabricated steel cage installed and

the concrete poured. During excavation under the props the blinding was removed.

Vibrating-wire embedment strain gauges were installed in three of the RC props to

measure axial loads. The locations of the monitored props in the instrumented section are

shown in Figures 3-39 and 3-40. Each RC prop was fitted with four gauges, 2 m from the

northern end. The prop load is calculated from the average of the outputs of the four

gauges at each cross-section to eliminate the effects of bending. In addition, bending was

measured in prop P2 using two further sets of 4 gauges located at 4 m and approximately

6 m (at the centreline ofthe excavation) from the northern end. The gauges are situated at

0°,90°, 180° and 270° on the cross-section of the prop and numbered as shown in Figure

3-29. The pre-instrumented reinforcement cage for RC prop P3 was erroneously placed in

the position ofRC prop 4: hence RC prop 3 is not instrumented.

3-5-6 Backfilling and material placement behind North wall capping beam

On Day 476 backfill was placed behind the capping beam on the north wall up to within

0·5-1·0 m of the western extent of the instrumentation. On Day 477 backfill was placed to

within 0·5-1·0 m of the eastern extent of the instrumentation. Backfill was placed behind

the capping beam at the instrumented section on Day 480 (Figure 3-41).

49

On Days 487 and 488 a large crane platform (1200 tonnes) was constructed behind the

wall close to the instrumented section, using 400 tonnes of ballast. (The crane was used to

lift a nearby footbridge into position.) Some time between Days 489 and 495

approximately 700-800 mm of hard core was placed and compacted directly behind the

wall over an area encompassing the whole length ofthe instrumented section to provide a

stable platform for the long-reach excavator on its return.

3-5-7 Excavation

As illustrated in Table 3-3, there were two main excavation phases: Excavation Phase 1

reduced the ground level within the cutting to just below temporary prop level (36·759 m

AOD) and Excavation Phase 2 reduced the ground to formation level (33·559 m AOD).

Figure 3-42 shows an elevation of the instrumented section indicating the dates on which

particular regions were excavated. The bulk of the material was removed with a long­

reach excavator situated at ground level behind the wall (Figure 3-43). A more

manoeuvrable mini-digger located within the cutting was used to remove any remaining

material. In order to minimise instrument cable breakages the long-reach excavator

operator avoided the soil around the cables within the excavation and the mini-digger was

used to carefully remove this material at a later stage. A blinding layer was poured across

the base of the excavation within a few days of reaching final formation level.

3-5-8 Temporary props

The temporary props were hollow steel sections with an external diameter of 1016 mm, a

wall thickness of22·2 mm and a Young's modulus, E, of205 GN/m2• The load from the

wall was transferred to the temporary props by a concrete thrust block and waling beam

arrangement, as indicated in Figure 3-21 and Figure 3-44.

The load in three temporary props was measured with 12 vibrating-wire surface-mounted

strain gauges. Four gauges were fitted to each prop, 2 m from the northern end. The

gauges are welded onto the prop (see Figure 3-45) at 45°, 135°,225" and 315" on the prop

cross-section and were covered with insulated metal housings to minimize damage from

machinery and falling soil during excavation. The positions and numbering system are

shown in Figure 3-29.

50

3-5-9 Base slab

The dimensions of the base slab are shown in Figure 3-46. The base slab was constructed

in 11 m long segments. The instruments were installed such that they traversed the same

section of wall as Slab A4 (shown in Figures 3-23, 3-39 and 3-42 - note that the intension

was to monitor P3 instead ofP4: see Section 3-5-5). The base slab to the west (A3) was

poured on Day 571 and the slab to the east (A5) was poured on 590.

Eight vibrating-wire embedment strain gauges were attached between the reinforcement

steel in pairs along the centreline of the base slab, at 2 m intervals, to measure load in the

base slab. One gauge of each pair was at the top of the slab and the other at the bottom

(Figures 3-29 and 3-47). These gauges were connected to the datalogger on Day 575, four

days before the base slab was constructed, so that readings of strain and temperature could

be taken before and throughout the concrete pour.

The base slab gauges were irrecoverably lost on Day 1278 (ih April 2003) when the

tunnels were cleared out in preparation for installation of the over-head high-voltage

transmission lines.

3-5-10 Installation of storm drain

Six months after the main construction activities had been completed a storm drain was

installed approximately parallel to and behind the north wall on 9th November 2001 (see

Figure 3-48). The storm drain was about 0.5 m diameter and the distance between its

centreline and the edge of the wall was approximately 4 m. It was installed at an elevation

of approximately 41m AOD.

3-6 Data collection and vibrating-wire instrument technology

Vibrating-wire instruments were used because they are relatively easy to install and can be

monitored remotely and continuously using a datalogger. All of the instruments were

wired to a CRI0X Campbell Scientific datalogger with 2 vibrating-wire interfaces and 7

multiplexers (Figure 3-49). On collection the data was saved in both the datalogger and a

storage module to minimise data loss during data downloading. A photovoltaic cell and

inverter were connected to the datalogger battery to ensure continuous battery charging;

hence the system was effectively self-maintaining. A remote data collection system was

51

set up to collect and process the data (Richards et ai., 2003). This enabled data to be

automatically downloaded daily (via a modem) to several storage locations at the

University of Southampton, the raw data readings to be converted to AGS-M fOlmat

(CIRIA, 2002) and subsequently uploaded to a remote server where the data could be

viewed via a website within 24 hours of their collection.

For operational reasons (for example to allow any cable breakages occurring during

excavation and construction to be detected quickly) the datalogger was set up to take

readings of all instruments every 5 minutes so that at any time on accessing the datalogger

a reading obtained within the last 5 minutes could be viewed. Instrument readings were

stored in the datalogger at intervals of between 5 minutes and 2 hours, depending on the

level of activity on site.

All of the instrumentation used on site to monitor the structural loads, soil stresses and

pore water pressures used vibrating-wire transducer technology. Vibrating-wire

transducers contain a tensioned wire which is 'plucked' by a solenoid via a datalogger

instruction and the resulting frequency of vibration of the wire, which is proportional to

the strain in the gauge, is measured within the magnetic field of the solenoid (see Figure

3-50). As the length of the strain gauge (and therefore the wire within it) elongates or

contracts (depending on the stress patterns in the member in which it is installed), the

frequency at which the wire vibrates changes accordingly. The strain gauges also contain

thermistors for temperature measurement.

To obtain a strain reading from a vibrating-wire transducer the datalogger sends an electric

pulse which excites the solenoid within the instrument creating a magnetic field, causing

the wire to vibrate. The minimum and maximum frequencies of the 'sweep' are specified

by the user and depend on the type of gauge being used. The measured frequencies are

sent to the strain gauge sequentially starting with the lowest frequency and finishing with

the highest; this takes 150 ms. Although the wire vibrates at every frequency, the non­

resonant frequencies die out quickly, leaving only the resonant frequency which remains

for a relatively long time. After waiting for the non-resonant frequencies to die out (20 ms)

the datalogger then measures the time taken for a user specified number of vibrations to

occur. This is used to find a value for the average frequency of vibration. The datalogger

then records the square ofthis frequency (expressed in units of kHz2 - see Equation 3-4,

where T is the period in milliseconds).

52

F 2 = _1_ 1 •

T-Equation 3-4

For this project the datalogger was set up to sweep between 800-1300 Hz and to obtain the

average frequency reading over 80 cycles for the concrete embedded and surface mounted

strain gauges and to sweep between 1600-30000 Hz and take readings over 100 cycles for

the spade cells and piezometers. These figures were based on the instrument and

datalogger manufacturers' advice.

At the reading interval specified in the datalogger (e.g. 5 minutes) readings are taken from

all instruments in succession. When using a large number of instruments in any

datalogging system it is important to ensure that the interval is long enough for all the

instrument readings to be obtained successfully.

Some instruments experienced periods of downtime due to inevitable cable breakages

caused by the machinery used during excavation. The full-time presence of the researcher

on site, working closely with the contractor during the construction period, kept these

losses and downtime to a minimum.

53

VI .j:>.

PRESSUREMETER TEST

I ~'PR 3593 (reference from

"" CTRL Site Investigation)

"'" SA4116

"'" " "'" ~/~

" "':?O/C SA 8428 ",,?a/JI,!

BH # - Wireline boreholes undertaken by University of Southampton

SA #### - Useful boreholes from CTRL Site Investigation

NORTH WALL

BH1 ~

~:2

••• TH} ...... ~12m SOU ••

WAL'o... 12 m ~ ..... SA849~ .... d··· ..

••••• 'Instrumente •••• ... ,

••••• section ••• ••• ••• •••

SAl 813.6: ~""

"'~:~I,! NADIR SUMP "'?~:~ll'

",,~IIJ,f

•••••

SA 8~1'40'", SA ~~~~""SA 6104

",," SA 6106 "" "

SA 8315~"T ~ 002

MAIN INSTRUMENTED SECTION N

+ "~)~A8497 1 1 I

o 100 m 200 m SA 9154" SA 8316A~"-

SA 8316"

Figure 3-1: Plan showing the locations ofthe areas described in this work

Vl Vl

,f ectlio7l/ rrtnlVJ.V c wn"C&v6Tb z:n,,·

CTRL at Ashford - Instrumented section

7f-19au .riO,.,.

Figure 3-1: Geology of the Weald of Kent (Mantell, 1833)

JJlflr ru ar INNdfloe

56

Figure 3-4: Light brown band derming the boundary between the upper and lower Atherfield Clay (observed in the nadir sump)

Figure 3-5: Showing the peds and softer matrix of the upper Atherfield Clay

57

Figure 3-6: In situ block sample of the Hythe Beds showing yellow/grey mottling and dark brown root material

58

Upper Atherfield Clay

Figure 3-7: Looking into the cutting between RC props during excavation

Figure 3-8: Elevation of the excavation face

59

Figure 3-9: Shear plane in the upper Atherfield Clay, 50 metres from the instrumented section (courtesy of Mark Roberts, Geological Specialist for Skanska

Technology Ltd.)

60

North wall 0 -rN

o Instrumented 00 section of wall

Material is smoother with very: 00 hard peds visible I 0

~ I 00

South wall 0 I Material is 0 o I more crumbly o I in appearance

00 :

o ~~O Line visible in base o of excavation

o

Figure 3-10: Sketch of discontinuity (possible shear plane) observed in the bottom of the excavation at the main instrumented section (plan view)

61

0\ tv

Moisture content, %

o 20 40 60 80 100 O +I--~+-~~--~~~~--~~

5

E 10

Qj > ~

'0 15 c: :::l

e C)

3= .E 20 (1) .0 J:: Q. (1)

o 25

30

35

Made Ground

. - 0 - • Hythe Beds (firm to stiff closely •• 00 • • fissured mottled light grey and

.a • light brown sandy CLAY) • • upper Atherfield Clay (stiff

• 0 • to very stiff fissured grey :: : sandy (fine) CLAY)

.0· 0 • · Change in PI 2.3 m above ..... ---;e---.. ---~ ------------ -.----- lip~iei-il6wer Atherfield boundary

~ . - ---- - -- -- - - - --o. • lower Atherfield Clay (stiff ~. .

to very stiff closely a • 1 c;: • fissured brown CLAY) a • o. •

~~ . F • •

---CJI •

o • •

Weald Clay (stiff to very stiff closely bedded grey CLAY with frequent sections containing sand (fine) laminations)

~ ---. - c ·-E Q) E .- ....... . -uc~ ._ 0 ._

- U :::l en CJ ~~= 0... ::::l -en

'0 E

Moisture content, %

o 20 40 60 80 100 o +I __ ~~-J __ +--L~~J--+ __ ~~

5

E 10

~ ~ '0 15 c: :::l

e C)

3= .E 20 (1) .0 J:: .... Q. (1)

o 25

30

35

o o

Made Ground

Hythe Beds o q

t · . } upper boundary to Atherfield CLAY

0: . . upper Atherfield Clay (stiff E • to very stiff fissured grey

• 0 • sandy (fine) CLAY) -------- --.:> --- . --- ----- ---- ------ --- --- -- --

.a. Change in PI 2.3 m above _ _ _ !k:(jI •• _____ ~ee.e!~~~t:r Atherfield boundary

c. • lower Atherfield Clay (stiff CJI • to very stiff closely fissured o. •

o. • brown CLAY)

o. • CJI •

o. • 0_ o. • ~ . o • •

o

Weald Clay (stiff to very stiff closely bedded grey CLAY with frequent sections contain ing sand (fine) laminations)

~

~(]} ....... E ..... C = .3 2 ~ '6 § ~ E U

0:::

Figure 3-11: Geotechnical profiles including liquid and plastic limit data from wireline boreholes BHl and BID (see Figure 3-1)

60

50

40

c ~ 30 E

20

10

Hythe Beds

upper Atherfield

Clay

lower Ath. Clay

.. -'" --...... -_ .. ------.. -----.. .. - .. ------- ... - ... -... --_ ..... -----------_ ... _--- .... -------------.. "" .. -----------.. ---.. ---.... -.. ------.... -----. ------------ ... -------- -------------~~~~~~:~ -:-:-: -:-:-: .. :-: -:-:-:-:------- -_ ... _ .... ----- ---------------------.. _---------­--------------------------------------

Original ground level material removed i··· ···· ·· ··· ·· ····i before nadir sump

i Hythe Beds ! construction . i ~ -----------------i I

upper Atherfield

Clay lower Ath.

Clay ------... -.. -.. -.. -... -.. -.. ------~: .. :-:-: -: ... : .. :-:-:-:-: -:-.. -- ... ------------- ... ----_ .. _- --- ... ---- .. - .. _----- - ... --------- .. _-----------------------------------------.. -.. -.. ---.. -----.. -.. -------------- --- .. ------------- ... ---------~~~~!~~~i~-------------------- .. ----------- .. ------------------... _---------_ .. _ ... _-------_ ... -------------_ .. _--------­"' .. '" --...... --... -----........... -" ... -... --------------------------------------------------------------------------------------------------_ .. _-

Hythe Beds

upper Atherfield

Clay

lower Ath . Clay

...... _ .. _- .. - ... -­------ ... --.. _----- ----­------------------_ .. ----. .... -.. -... -.. -.. -.. -.. -.. -... - ..... ------------- .. --------.. -... -... -... -... ---... ---... - -----

::wealEf:Gia~-:: _________ '1...

--------- .. -----_ ... _------..... __ ._-----_. ----.. -_ .. _-... ---.--------- .. .. -_ .. _-----..... _--- ---------,

Northwest 250 m 600 m South

o +-------------~------------~------------. self-boring

pressuremeter PR3593

nadir sump SA8313A

instrumented section

Figure 3-12: Comparison of geology at locations referred to in this study (see Figure 3-1 for locations)

63

Bulk density, Mg/m3

1.5 1.6 1.7 1.S 1.9 2.0 2.1 2.2 2.3 2.4 2.5 50 +t-------= .. -=--_-...... --J.'-.... -.-.... -1. ...... _ ..... _ .... --'----L----JL..---'---...L...--'----'

Ground level at/ + * Gr?und !evel and soil 45 nadir sump x -ttt profile at Instrumented

_______ .. _.. _:T" x /! section

40

35

30

c ~ 25 E

20

15

10

5

n

x :- Made ground

================{~==========B¥frliLBeds • x ~ • x W' ... ~. upper Atherfield

x ......... ~ Clay

x # -------------------------+--------_. x • lower ...

...... Atherfield Clay -----------------------.----------_.

...... +. Weald Clay

+ • ... +

+ ...

+ ... ...

... SAS315 - 71 m from instrumented section • SA9154 - 90 m from instrumented section x SA4116 - 590 m from instrumented section (nr. nadir sump) + SAS313A - 60S m from instrumented section (nr. nadir sump) • samples taken durin...9 excavation of the nadir sump

Figure 3-13: Bulk density with depth (see Figure 3-1 for relative borehole locations)

64

c;;-c... 6 (/) (/)

~ 1ii ro Q) .r: CJ)

400 ,---------~--------~----------~--------~--------~--------_,

, , , 300

, ,.. --- --- -- - --- ------ --:- ----------- -- ------ --------------------:--------------------:--------------------;--------------------

........ : " ...... :

200 ------------- ---- -- -1----- ----- - --------- --------------------r--- -- ---------------~------;?- --- -i----- - ____ a_awe_a.

, , , , , , , ,

:,.....

100

O +-------~_r---------+--~----~r_----~--+_--------_r--------__4

o 100 200 300 400 500 600

Mean effective stress s' (kPa) mid

Figure 3-14: Effective stress paths oflower Atherfield Clay specimens during monotonic shearing (Xu, 2005)

65

Undrained shear strength (CU)' kPa

o 100 200 300 400 500 600 700 O+----'---t--'---t------'--+-~_+-'--_t_-'--I__--'--____I

5

10

15

E

:S g.20 c

25

30

35

40

x x .... ~

• x

•

• I

x

x x ,......... 1013 kPa

~--------------------------~

• Cu found from 100 mm samples x SPT 'N' x 4.5

- best-fit line used for upper soil ..... best-fit line to Weald Clay SPT data

• • 38 mm tests from SA 8313A ... 38 mm tests from SA 8315

xx

approximate boundary between Atherfield Clay and Weald Clay

x

x

x x

• x x x ....

x

.... • x •

x

x x

x

"~

x

x

x

x

Figure 3-15: Undrained shear strength with depth found from 100 mm diameter laboratory samples and SPT results (38 mm test data are included for comparison)

66

5

10

15

E

:5 20 c. CI)

C

25

30

35

40

Undrained shear strength (c u), kPa

o 100 200 300 400 500 600 700

•

• • •

•

best-fit line found from laboratory samples (Figure 3-15)

•

• +--928

•

Figure 3-16: Undrained shear strength with depth measured with the self-boring pressuremeter

67

5

10

E

:515 c. CD C

20

25

30

Total horizontal stress, kPa o 100 200 300 400 500 600 700 800 900

, , , , , , , , , , , , . , , , , . " , , , , , ." ., , , ~ , . , ,

•

•

• •

'. , , , , , " . . \ , \-;--- overburden

\

• \ \

\ \ ,

\ , , ,

Figure 3-17: Total horizontal stress measured by self-boring pressuremeter

68

1.00E-09 1.00E-08

Permeability, m/s

1.00E-07 1.00E-06 1.00E-05 0+---~~-k~~4---~~~~~4----k~~~~~--~~~~~~

5 E

~ 10 ~ '"C

§ 15 e C)

~ 20 a; .c :5 25 Co CI)

c 30

35

•

Atherfield Clay •

•

•

•

•

Hythe Beds • • •

•

•

Weald Clay

\ trend line for Weald Clay data

y = -2.5364Ln(x) - 24.016

Figure 3-18: Permeability data

69

44

42

40

38 \

36

34

32

30

28

0 26

~24 E~22 ~20 CI)

...J 18

16

14

12

10

8

6

4

2

0

0

Ground level

Intercept = 42.7 m AOD Made Ground

_-========================8y~~=~~~~== \

\ \

\ \

./" Best-fit line through the measured pore water pressures (above 32 m AOD) u = 8.28 x (42.706 - Level (m AOD»

x\\x lower Atherfield Clay -------~-----------------------------

\ , Wall toe

Hydrostatic

, , , \ , , , ,

\ , , ,

/\\, groundwater profile based on 5 m drawdown from original ground

, \ , ,

Weald Clay

x pwp measurements 1.275 m from wall

, , \

\

water level \ u = 9.81 x (37.709 - Level (m AOD)\ , , , , ,

50 100 150 200 250 300 350 400 450 Pore water pressure, kPa

Figure 3-19: Pore water pressures: measurements and assumed profile

70

N

Instrumented section

@ CreoNn Copyri!~ht Ordnanee Survey. An EDI~tA. D igirnal','JISC supplied service .

Figure 3-20: Map of central Ashford showing the location of the instrumented section (Grid reference: 60081424)

71

South wall

mAOD

43.709 -.SZ.

37.259 _~ . -

/ I

I

I

North wall

Capping beam

Reinforced concrete prop I I

Thrust block ~

Temporary prop £&I --------- Waling beams ~

33.559 ...sz:.... r Base slab 1

Contiguous ~ bored pile wall

23.459

12.1 m

~ -

Figure 3-21: Cross-section of the cutting at the instrumented section

72

-.....) W

~

"""""" LMJ.. 1 I'ROI'05Il' ~

~. tDC£ _ _ "".."..,....,.. -i ~ Of. owe. Nroa. 4.JO-t)US- 08'H~-.£0920 6:21

~ ~.

II ~

if ;;

'" ~

&9,

~ 1000 -tJ~

11_ ft$)[ -pAPJ.PO ~AU.S

SOUTH WAll _J~_ 1.Q!l!t,

~.~_:.:-n .. _ ... -.... _. "'.

J 000.:'000 RD!'EDRCED COIIC>I£l£ __

PR(~ CONCReTE ORAlIMGE ~ NID I)I!I.INACE '!WET wnL rOf! l'II'ICJol. DEfNLS WU! '0 D¥I'G 'ios. OO-OUB- 088SO-409JO to 12

c.au- 00C1

2200

" 14~

t~ ..... I

lISlTy CCIHr.9Dt 111& SUoII

75- ntC. Sf ......,..; I.<VDI

4"!'O

~ $"'-...,M

..... ..... ~ .... -..

300 • 600 toHt1NUOuS RC COI!8EL (IYP.)

noo

=

+.~~

:l :. ~

~,

1'-- ~._. __ _

9" ~ .

§

1 ' J

n ~ ~ ~

l! ~

PR<lPOS(I) ...... It> Ll\IEl.

I~ CIA COHfQJOU5 ~ PU

Figure 3-22: Elevation illustrating the ground profile around the structure

I . ~IINOST_LOCP

~,~ '<7 •

I :r-'--______ L _________ _

000 IWJ.AST lNO 8'1"" C4J.4 COI'ITftltCTOft

6 7 8 9 10 -1100 ll5 - W 1100 1 - 2 3-_ _ _ ! i

- - - - -1800 Z X Y

-cr- @®®©@@@)®®®®© ®C;>CD©®®CD®® 1800 11 12 '

Dimensions in mm

N

~

- - : -11000

Extent of base slab A4

• Spade cell

(j) Pile and installation period

o Approx. sand drain positions

Z Inclinometer

X & Y Instrumented piles

Figure 3-23: Plan of the north wall in the instrumented section showing locations of instrumented piles, spade cells, sand drains and sequence of pile installation (see

Table 3-4)

Ground level

~ SZ 3.3 m bgl

~ sz 5.3 m bgl

Line denoting change Approx. ~J'. 316 8 in plasticity ~ casing depth ~~I/' ~ t ~ sz 8.3 m bgl

------------ -- ------ --- -- ---------------- ------------ --- ----- '. --- i-- ~-- -~ ------ - - - --- - -------13 11 ~ i ~SZ11.3mbgl

--~------~------------~--~~:~

Made ground

Hythe Beds

upper Atherfield Clay

lower 1:0 Atherfield Clay t sz 15.3 m bgl

20.3 m bgl sz ___ ____ ____ _

• Spade cell Horizontal dimensions in mm bgl- below ground level

18001800

:2200:

Weald Clay

: 2~00 : Dimensions in mm

Figure 3-24: Elevation of instrumented section showing spade cell locations

74

Day 1 - Bored Day 2(a) - Bored Day 2(b) - Filled Day 2(c) -open to 8 m then to 21 m with no with bentonite Concrete tremied casing inserted support support in from the base

CaSing!

Excavate to !

21m 1

Figure 3-25: Pile installation sequence

75

Figure 3-26: Pile boring

76

Figure 3-27: Installation of casing

77

Figure 3-28: Installation of reinforcement cage

78

North Centreline of wall excavation

mAOD These gauges only in RC prop P2 (see Figure 3-39)

43.709.5Z 1 Gauge reference

4EJ2 1 mAOD numbers

/ 3 B

Prop cross-sections .5Z 41 ·5 25 & 26 ~

showing positions of strain Strain gauges in gauges (facing north) the props

5L 40·0 23 & 24

~ ~ ~

5L 38·5 21 & 22

401

Temporary prop 5L 36·9 19 & 20

3 2 I

5L 35-4 17 & 18 ~

33.5595L Base slab 5L 33.9 15 & 16 ~

5L 32.5 13 & 14 ~

A pair of strain 5L 31 .0 11 & 12

gauges in the ~

base slab

5L 29.5 9 & 10 ~

5L 28.0 7&8

~ ~

Pairs of strain 5L 26.5 5&6 gauges in a pile ~

5L 25.0 3&4 ~

5L 24.0 1 & 2 23.459 'V ~

- -- ------- - --------------------------

Figure 3-29: Elevation of instrumented section showing strain gauge arrangement

79

00 o

Strain gauges ~~~~:

Figure 3-30: Strain gauges for measurement of pile bending moment

00 ........

Figure 3-31: Photo showing sand drain and cable for spade cell 11 (taken on 13th February 2001 - Day 495)

450 period of pile installation

1----.1

Beginning of main construction period

400 85-15.3~ :

~350 ~

t300 ~ .... (1)250 S c 2200 '': o .c 150 S ~100

50

1

812 "\r- l-/'

84 -~l ~1

O+OTT+r~~~"O+TT~"rhno~TT+r,,~~no~TT~~rhno~TT+r~

~150 ~

~100 ... Q. ... CI) 50 ;

o 28 56 84 112 140 168 196 224 252 280 308 336 364 392 420 448 Time in days

period of wall period of sand installation drain installation :~ -::-- P13-11.6m

P13 1 1 II 1 1 II I P11~~~ "~?-15.3m~ Pi? 1 : ~ P11-mm P12 : --~-v- ~-1"[3m

1 II 1 II

~ O+o"+OoT+>+r+r"r.rrrrrrhnrrl"""~"o+,,o+,,o+~.rrr.rrrrhno~ o c.. 0 28 56 84 112 140 168 196 224 252 280 308 336 364 392 420 448

Time in days

Figure 3-32: Changes in total horizontal stress and pore water pressure in front of the wall around the period of sand drain installation

82

00 w

Figure 3-33: Excavation either side of the north wall before break down to level of capping beam (taken on 1st November 2000 - Day 391)

Chains attached to pile with bolts

Figure 3-34: Pile top removal (south wall) (taken on 31st August 2000 - Day 329)

84

00 VI

Spade cell cables protected in plastic tubing

Figure 3-35: Wall before construction of capping beam (taken on 4th December 2000 - Day 424)

Wall was constructed here at a later date

t 850 500

Capping g Beam L!) T"""

150 Pile

I" • I 1050

Figure 3-36: Dimensions of capping beam (in mm)

Datalogger

Inclinometer • • tube

Figure 3-37: Capping beam (taken on 16th January 2001 - Day 467)

86

00 -....l

r·~aWlJl.;W;; V'ltJ:--·::·:---...!._' ''d ' ,.fI;~.~ """ , { .m!!iI·"t.l· .. ·l!I!!)!i!iii',·" .. :" ,. .* --------

Shuttering

Figure 3-38: Preparation for RC prop construction (taken on 16th January 2001- Day 467)

Direction of construction

Props P# - RC prop T# - Temporary prop

Shaded props are instrumented

Extent of base slab A4, 11 m

N

'x Pile

Z Inclinometer

X & Y Instrumented piles

Figure 3-39: Location of the instrumented props in relation to the instrumented piles

88

00 \0

Figure 3-40: Photo showing instrumented area before excavation (taken on 220d January 2001 - Day 473)

\0 o

Figure 3-41: Compacting backfill in the instrumented section (taken on 29 th January 2001 - Day 480)

\0 ......

Level, mAOD

43.709

42.209

Piles RC props

/ ~

Shaded props and piles are instrumented

Dates show when shaded areas of soil were removed

Piles X and Yare instrumented with strain gauges

Pile Z is instrumented with an inclinometer Capping

+-~~~L-______ ~-L ________ ~~ ______ -li~ ________ ~-L ______ ~~~ ______ -L~ ______ ~beam(see

:;:::;: Figure 3-36)

483-484 (Thu/Fri) ;;;:;:; I II I: I

!~m:'~i1rlU : : . Ir Tlcn) II Sli (Tue) 37.25~_\Z _ _ -_·_,·--·FJ-II III I,d ~ II ~ I !~:\ l i1.:>1 1 ~I ~::C 11 ® II]JJ~r~-J[ ~-n~ () Waling

beam

33.5~

I II II II II I nq TU ,'-,',), ...... .

by 524 "flied) .... ,', ,.,' .. , I-- \VY ... • ..

I II S~~b A~ ~ ins,~lIed : D~~~1 ! ~Iab A4 - installed .~ay 5!9 ~I~~e

If

wwwwwwwwwww~~ww~wwwwwwwww~

Figure 3-42: Elevation showing props and order of excavation. The shaded areas were excavated on the dates indicated, and the props and base slab were constructed on the dates indicated (excavation took place in 2001).

Figure 3-43: Excavation using long-reach excavator

Figure 3-44: Temporary prop with strain gauges

92

Figure 3-45: Temporary prop gauges welded onto prop

93

Corbel

2.5 % slope

~ I r---_____________ -+-____ B_a_s_e_s_la_b ______ -l

1600 4450

Figure 3-46: Elevation showing base slab arrangement (dimensions in mm)

Figure 3-47: Strain gauges wired onto base slab reinforcement (photo was taken looking down through base slab)

94

Figure 3-48: Storm drain installed behind north wall in November 2001 (Day 864)

95

Vibrating-wire interface

Storage module

" -,

Multiplexer

Modem

Battery

Figure 3-49: Datalogger set-up

~/ 2 Core Cable ~ ____ s~r

r--~~ (/ /"

Electro ~ " Magnet ,...

f "" .... _, t 1 \ \ , , \ , I , , \ •

INPUT PLUCK PULSE

(12 to 60 volts peak)

~ , , , ,

• f I \ , ___ ; I I 1 _ _

, • ______________ ~'~\~)~--~l ~I~I~-----------4 • • (" ., I I \

- , , , .-,~, ... - ..... -/-,- .- .... \' ; I , . ' ; " ,

" Magnetic Fleld---' , - - - - Piano Wire

Figure 3-50: Vibrating-wire gauge (Gage Technique)

96

4 STRUCTURAL MONITORING

4-1 Introduction

In this chapter, details of the instrumentation and the methods used to determine the

retaining wall displacement profile and the structural loads are presented. Data

representing the wall profile, wall bending moments and the prop loads at significant

stages of construction are presented and discussed. The instrumentation used to monitor

soil stresses and pore water pressures in the soil around the retaining wall (spade cells) is

described separately in Chapter 5.

97

4-2 Inclinometer

4-2-1 Description and Use

An inclinometer is a device for measuring inclination, or tilt. Inclinometers have been

used in numerous construction projects over at least the past 30 years to monitor ground

and structural movements, often as part of observational methods of construction (PecL

1969), e.g. The World Trade Centre (Saxena, 1975); Limehouse Link (Glass &

Powderham, 1994); Channel Tunnel Rail Link (Loveridge, 2000) and the Heathrow

Airside Road Tunnel (Hitchcock, 2003).

To find the retaining wall's deflected profile an inclinometer probe (Figure 4-1) is

manually lowered to the bottom of a tube installed within a pile. As the probe is pulled

back up, readings of the tilt angle are taken (using accelerometers) at 0'5 m intervals in

directions perpendicular and parallel to the wall. The inclinometer probe follows

longitudinal grooves in the tube to ensure that it does not rotate as it moves upwards. Base

readings are taken before the construction process being monitored begins (e.g.

excavation) and subsequent readings reveal changes that occur due to construction

activities.

To measure the absolute movement of the structure or ground mass being monitored,

either the tube must be extended below the structural component into ground known to be

stable (i.e. not affected by the construction process), or the top of the tube must be

surveyed at each reading. At the instrumented section the inclinometer tube extends 10m

into the ground beneath the pile toe to ensure fixity so that the absolute movements of the

wall can be determined. Further information regarding the use of inclinometers can be

found in Dunnicliff (1993).

Mikkelsen (2003) describes the types of errors that may occur when using an inclinometer

probe. In summary, these errors can be minimised by ensuring that:

• the same probe is used for all readings in a borehole;

• the same user takes the readings every time;

• the user is technically competent;

• the person who collects the data processes it; and

• the inclinometer probe is regularly calibrated.

98

These measures were taken by the monitoring team on the CTRL, who took inclinometer

readings across the site at Ashford (the data were mainly used for implementing the

Observational Method during construction).

In order to improve the accuracy of the inclinometer measurements two sets of data arc

collected for each profile: for the second set the inclinometer probe is rotated through

1800

• The difference between individual pairs of readings at a given level is called the I:lce

error or checksum. This is equal to twice the zero offset, or bias. The face errors are

displayed on the readout unit during data collection so that the user can check that these

errors are approximately constant throughout the survey. If the face errors are constant

there is no overall effect on the final inclinometer measurements. However, operator

technique, instrument performance and casing conditions (such as variability at joints or

dirt) can affect the face errors. For example, because the wheels on the inclinometer are

angled (Figure 4-1: they are sprung to ensure that the inclinometer runs in the tube

grooves), at tube joints the wheels may be resting on different sections of tubing and

therefore larger face errors may occur. In this case, larger face errors will exist in all

profile readings and can therefore be identified.

Ifthe face errors are randomly inconsistent, analysis ofthe individual readings can reveal

where the errors occur and the data can be altered accordingly. Figure 4-2 shows the face

errors from an inclinometer profile taken on Day 581 (after Base Slab Construction and

before Temporary Prop Removal). Some unusually large face errors are recorded at

various locations in the inclinometer tube. Table 4-1 shows the Face A and Face B

inclination readings at 41·593 m AOD and the calculated face errors for the data sets taken

on Day 581. At this reading location only the Face A reading was inconsistent. Having

identified the locations of these large face errors, comparison of the readings recorded at

these locations with those taken on the preceding and following days allows the rogue

readings to be identified and adjusted (amendments are shown in brackets).

Day 523 537 538 542 549

Face A 4·1 4·1 4·6 (4'1) 4·0 4·2

FaceB -2'4 -2,4 -2·5 -2·4 -2·3

Face error = (A+B)/2 0·85 0·85 1·05 (0'8) 0·80 0·95

Table 4-1: Example of inclinometer data analysis and correction using face errors: 41-593 m AOD

99

The original and amended profiles for readings taken on Day 581 are plotted in Figure -+-2.

The other data points highlighted in this plot were similarly analyzed and altered. For this

profile adjusting the data in this way 'moved' the top of the wall by 1·5 mm. This is about

8% of the overall movement at the top of the wall. All the inclinometer data were adjusted

in this way.

A further check on the accuracy of the data can be made by studying the wall toe

movement. Figure 4-3 shows the deflection of the wall toe with time (from the

inclinometer measurements). Although the data are scattered, those recorded before the

base slab was installed appear to show a general movement of the wall toe into the

excavation. The equation of the best-fit line to the readings taken before Base Slab

Construction is shown on the graph. (NB: The movement profile is not really linear as

different construction activities would produce more movement than others, however

since the total movement is so small (1'5 mm), the assumption of a linear movement

profile is justified.) Figure 4-4 shows (a) the as read profiles of the entire inclinometer

tube (adjusted for face errors as described above); (b) the corrected wall profiles assuming

a fixed toe; and (c) the corrected wall profiles assuming the toe moves according to the

equation found in Figure 4-3, for a small selection of the data taken during Excavation

Phase 1. Figure 4-4 shows that the adjustment described above produces much more

consistent data which is more easily viewed and analysed. The scale on which the

deflection data are plotted covers a small range of deflections, showing that in general

these errors are very small.

Less frequent inclinometer readings were taken after Base Slab Construction so it is more

difficult to infer the wall toe movement after this time. Figure 4-3 suggests that after Base

Slab Construction the wall returned to its original position and then slowly over the

following year moved back out into the excavation by a similar magnitude as during

excavation. This movement might seem reasonable; however Figure 4-5, which shows

inclinometer measurements taken within the entire inclinometer tube (i.e. including those

taken in the ground below the pile toe) following Base Slab Construction, shows that the

whole tube below the wall toe appears to be moving. This is highly unlikely because

movement of the wall toe would only be expected to cause movement of a metre or so of

the tube beneath. Therefore after this time the wall toe has assumed to have been

stationary.

100

Figure 4-6 shows all wall profile data found from the inclinometer data (adjusted for errors

as described above). Figure 4-6(d) shows the inclinometer profiles taken after Base Slab

Construction over this period corrected for zero movement of the toe. The profiles show

the top of the wall gradually moving into the excavation with time, with the exception of

the data taken in June and July where the wall moves very little. This behaviour would be

expected during and following Temporary Prop Removal. Therefore, the assumption that

the toe has remained in the same position as it was at the end of construction appears to be

valid. The inclinometer measurements show that the wall toe is in a similar position on

Days 585 Gust after Base Slab Construction) and 718 (over 4 months later).

The large deflections measured at the top of the wall are likely to be due to the

unsymmetrical geometry of the structure. The readings indicate that the structure has

swayed towards the south wall, where the ground level is lower than that behind the north

wall (Figure 3-22). This movement is described in more detail in Section 4-5-5.

The data taken below the toe of the pile can be used to determine the random errors

associated with the particular instrument and survey point. The maximum reading taken

over this section of inclinometer tube is 1·7 mm, indicating a maximum random error of

1·7 mm per 10m of inclinometer tubing.

Figure 4-7 shows the movement of the top of the wall with time. It shows gradual

movement into the excavation until Base Slab Construction after which the movement

reduces significantly. (It must be noted that this data is however based on the readings that

have been corrected for zero toe movement after Base Slab Construction. Therefore at a

minimum, the graph shows that there is little relative movement of the top of the wall to

its toe).

4-2-2 Calculation of bending moments from inclinometer data

Bending moments are frequently calculated from the bending strains measured using

vibrating-wire strain gauges embedded in concrete retaining walls (e.g. Tedd et al., 1984)

and piles (e.g. pile stabilized slopes: Smethurst, 2003). In order to estimate pile bending

moments from inclinometer readings, it is necessary to differentiate twice an equation

representing the deflected profile of the pile (measured with the inclinometer). To find an

equation for the wall's deflected profile, a curve (such as a polynomial) must be fitted to

the deflection data.

101

Ooi & Ramsey (2003) compared twelve methods for calculating bending moments from

inclinometer data including piecewise fitting of quadratic and cubic curves, and global

polynomial curve fitting. They noted that if using global polynomial curve fitting, the

lowest possible degree of polynomial that will "adequately" describe the data should

ideally be used. This will allow the main changes in wall shape to be modelled without

including small local errors in the inclinometer data. In general Ooi & Ramsey concluded

that piecewise fitting of cubic curves to windows of five points yielded "middle" values of

curvature and that when maximum curvature values found using this method were

compared to strain gauge data collected from laterally loaded piles the values were in good

agreement. For inclinometer data collected at the CTRL site global polynomial fitting has

been found to give good correlation with bending moments calculated from strain gauge

readings.

The second derivative of the equation representing the deflected wall profile yields an

expression for its curvature, K, of the wall, where K = liRe and Re is the radius of curvature

of the pile. The bending moment, M, is calculated from the product of the curvature and

the flexural rigidity of the pile, EI (Equation 4-1), where E is Young's modulus and I is

the second moment of area. For a circular cross-section with constant stiffness the second

moment of area is given by Equation 4-2, where r is the radius of the pile. The stiffness of

steel is approximately eight times that of concrete, therefore the flexural rigidity must be

calculated to include the effect ofthe steel in a reinforced section (see Section 4-3-3).

Equation 4-1

Equation 4-2

The pile concrete had 20 mm granite aggregate and a characteristic strength of 40 N/mm 2.

The mean 56 day cube strength of 61 cubes taken from the piles cast in the period the piles

in the instrumented section were cast was 73'0 N/mm2• It should be noted that this is

higher than might be expected, however it does refer to the 56 day cube strength (it is

more common practice to quote the 28 day strength for concrete).

The bending moment is defined as positive when the excavated side of the wall is in

tension.

102

It is important to carefully consider the degree of polynomial used to fit to the profile data.

A polynomial with a similar number of inflections as the shape of the profile should be

chosen, otherwise the bending moment plot produced will have too few/too many

inflections and indicate an inaccurate bending moment profile. A 5th order polynomial is

most suitable for representing the deflected shape of a wall which has a linear horizontal

pressure distribution acting on it (the pressure distribution is given by the 4th derivative of

the chosen polynomial). In reality the pressure distribution is unlikely to be perfectly

linear, and therefore both 5th and 6th order polynomial equations were fitted to the

deflected wall profile (determined by the inclinometer readings) before and after two

construction stages at which cracking appears to have occurred (described later). The

correlation coefficients for the curves were all greater than 0'997, which superficially I

indicates a high level of accuracy for the curve fit. Comparisons between the bending

moments calculated from the inclinometer and strain gauge measurements are made in

Section 4-4-5.

Factors that affect the stiffness of the pile and in tum affect the suitability of the curve

fitting include:

• changes in cross-sectional area of the element;

• changes in reinforcement stiffness; and

• concrete cracking.

These factors affect the continuity of the pile element and hence the mathematical

accuracy of the curve fit. It may be better to fit several curves to sections of the pile with

continuous properties. The reinforcement in the piles in the instrumented section reduced

from 50 mm to 40 mm diameter bars about 6 m above the pile toe (described in Section 4-

3-3). In addition, there was a bulge in the piles just below the level of the casing,

(described in Section 4-4-4). In this case however a better curve fit and bending moment

distribution was given by the curves fitted to the entire pile.

Cracking reduces the pile's cross-sectional area; therefore the value of I must be modified

to take account of this or the calculated bending moments will tend to be overestimated.

Further comments on fitting curves to inclinometer data are made in Section 4-4-5.

1 It is important to note that the correlation coefficient can be misleading: the higher the order of polynomial fitted to the data, the higher the correlation coefficient, but this does not necessarily mean the curve fit will give the best equation for the bending moment profile for the reasons already explained.

103

4-3 Vibrating-wire gauges

4-3-1 Use

Precautions that must be taken when using vibrating-wire gauges include the avoidance of

magnetic fields and magnetic particles. Since the railway was opened there has been no

obvious influence from passing trains on the data. It is also important to ensure that the

elasticity of the gauge approximates to the elasticity of the concrete for the period over

which the measurements are to be taken (British Standards Institution, 1986).

4-3-2 Method/or calculation o/prop loads/rom strain gauge measurements

In order to calculate prop loads and bending moments from strain gauges the strain in the

wire (and therefore the prop or pile), 8, is calculated from Equation 4-3 (where F\ is the

datum frequency of the wire equivalent to zero strain in the prop, F2 is a subsequent

reading and GF is the gauge or calibration factor, equal to 3·025 x 10-3 for the gauges used

in this project). The prop load, P, can then be found from Equation 4-4 (where A is the

cross-sectional area of the prop and E is the stiffness of concrete (for the RC props) or

steel (for the temporary props). The stiffness of steel has been taken as 205 MPa and a

value of25 MPa has been used for the stiffness of concrete (Mosley and Bungey, 1976).

This is a conservative value and the actual value may be higher (the concrete cube strength

data collected for the pile concrete is relatively high, however once load is applied creep

has a significant effect, effectively reducing the concrete stiffness). The calculated prop

loads and bending moments might therefore be underestimated.

Equation 4-3

P=& E A. Equation 4-4

Surface-mounted strain gauges are commonly used to measure temporary prop loads (e.g.

Richards et al., 1999; Batten & Powrie, 2000; Loveridge, 2000) as part of construction

monitoring (for example using the observational method: see Section 4-2). Many of the

issues concerning their use have been discussed by Batten et at. (1999), including

temperature and end effects and the influence of the instrument's location on the prop.

104

4-3-3 Method/or calculation o/bending moments/rom strain gauges

Bending moments in concrete retaining walls are calculated from strain measurements

obtained using embedment strain gauges placed in pairs at intervals up the concrete

section, with one at the front and one at the back of the wall at the same elevation (see

Figure 3-29). For an uncracked section of wall, the bending moment in the wall, lvI, is

calculated from the longitudinal bending strains &\ and &2 measured by the strain gauges at

the back and front of the wall respectively using Equation 4-5, (derived from Equations 4-

1 and 4-6) where y is the distance from the gauge to the neutral axis.

Equation 4-5

Y Rc =-. Equation 4-6 &

In the concrete piles used at Ashford there are 16 longitudinal reinforcement bars making

up 3·6% of the total cross-sectional area of the pile. The flexural rigidity (Elvalue) of the

pile has therefore been calculated to take the effect ofthe steel into account (see Appendix

C). The lower section ofthe pile (incorporating gauges 1-10: see Figure 3-29) has smaller

reinforcement bars and therefore a lower flexural rigidity. The flexural rigidity of the two

pile sections are given in Table 4-2.

Top of pile

Bottom of pile

Gauges

11-26

1-10

Flexural rigidity:

per pile, kNm2

2015300

1826600

per m run of wall, kNm2/m

1492800

1353000

Table 4-2: Flexural rigidity of concrete piles

For a circular section the flexural rigidity is proportional to the square of the cross­

sectional area, A. (Equation 4-7: given that I = A 12, where k is the radius of gyration (for a

circle 12 = Y4 ? and A = 1t r2)). Therefore when the area decreases (for example when

cracking occurs) the flexural rigidity of the pile decreases rapidly.

EI= EA2 41Z"

Equation 4-7

105

Uncracked behaviour is usually assumed to exist in the calculation of bending moments

from strain gauge data (Tedd et al., 1984; Wood & Perrin, 1984; Hayward, 2000).

However, it is important to determine whether cracking has occurred because if so and the

value of the E1 is not adjusted accordingly, bending moments may be significantly over­

estimated from gauge measurements. Details of how to calculate the cracked flexural

rigidity can be found in Branson (1977).

The cracking moment of the wall, Men is the moment corresponding to the maximum

tensile stress that the wall can accommodate (the modulus of rupture of concrete,.!,'.). The

most commonly used relationship between the modulus of rupture and cracking moment is

shown in Equation 4-8, wheref,. is given by Equation 4-9 (American Concrete Institute,

1992), Ig is the gross second moment of area of the pile (0·059666 m 4) and y, is the

distance from the centroid to the edge of the section. In Equation 4-9,!c' is the cylinder

compressive strength of concrete, and is taken to be 0·8!cu, where!cu is the 100 mm cube

compressive strength (Eurocode 2, 1992). On the basis of the analysis of the results from

over 12000 tensile strength tests, Raphael (1984) proposed Equation 4-10 for the

calculation off,..

Equation 4-8

Ir = O· 623.[l (in MPa). Equation 4-9

Ir =2.3/;2/3 (in psi). Equation 4-10

As stated in Section 4-2-2, cube tests of samples taken on site showed that the pile

concrete has a cube compressive strength of 73 MPa, giving the values for Mer listed in

Table 4-3. For these calculations Yt was taken as half the diameter of the pile.

Equationfr is calculated from

Equation 4-9

Equation 4-10

541

749

Table 4-3: The cracking moment of the wall, Mcr~ calculated by different methods

106

4-3-4 Correction for temperature effects

Structural elements expand and contract with temperature which causes changes in load

which are not induced by excavation or other structural loading. Temporary props are

particularly susceptible to changes in load due to temperature, as they are made of steel

which has a high thermal conductivity and therefore they absorb a great deal of heat

particularly when in direct sunlight. The relationship between temperature induced load

and temperature for temporary props is approximately linear (see Figure 4-8) and a

correction between temperature and load can be made based on this relationship.

However, it is important to remember that the readings represent actual loads in the props

which must be considered in design, and that the fluctuation of load with temperature can

be very large. During construction of the Canada Water underground station in London the

axial load varied by up to 50% during summer due to temperature effects (Batten e/ al.,

1999). The temperature versus load correction is therefore only made so that a comparison

can be made between design loads and loads calculated from back analysis of field

measurements.

The temperature corrected load, NT, is given by Equation 4-11, where N is the measured

load, To is the temperature of the gauge at installation (before any loading in the prop), TN

is the temperature at the time of reading Nand m is the gradient of the load versus

temperature relationship (for example m = 44·439 kN/oC in Figure 4-8).

Equation 4-11

4-4 Analysis of instrumentation data

4-4-1 Temporary prop data

The temporary prop loads uncorrected for temperature effects and the average temperature

measured by all gauges are shown in Figure 4-9. Before excavation there is very little

fluctuation in the readings because the gauge wire and body are composed of material

having the same coefficient of expansion as the prop and therefore there is no overall

change in tension in the wire as the prop and gauge expand and contract with temperature.

During excavation the prop loads are significantly affected by changes in temperature as

the ends of the props are restrained by the retaining walls.

107

The temporary prop loads corrected for temperature effects are shown in Figure 4-10. The

temperature correction has significantly reduced the variation in the readings with

temperature when load is recorded. However, before excavation, when there is no end

restraint to the props, the correction distorts the data so that it appears load is being

recorded. Hence, temperature correction is only appropriate when the props are restrained

to some degree.

The base readings for the temporary prop data (and the temperature correction) were taken

as the average of the readings from installation of the prop until the day before Excavation

Phase 2 began, over which time the readings were reasonably stable (this was 7 days for

TI and 6 days for T2 and T3). One ofthe gauges on TI (number 4) was broken during

excavation. Thereafter the load at the position of gauge 4 was taken to equal that measured

in gauge 3, as the loads had been similar up to this point, and the loads measured in gauges

3 and 4 of the other props are also similar.

4-4-2 Base slab data

Data from the base slab for the period from just before installation to approximately two

years later are shown in Figure 4-II(a). Set I relates to the pair of gauges to the north-west

of the instrumented section (closest to slab A3 which was constructed previously) and the

pairs are then equally spaced along centreline of the slab A4. Data collection began before

the concrete was poured so the fluctuation at the start of the graph (shown in detail in

Figure 4-II(b)) relates to the gauges expanding and contracting at a different rate to the

reinforcement to which they are attached. Zero readings for the strain calculation were the

average of the readings taken over the 72 hour period prior to concreting.

A large increase in load is indicated after the concrete was poured. The majority of this is

not load carried by the slab but a change in strain induced by the rise in temperature

caused by hydration of the cement as the concrete cures. The temperature increased to just

over 40°C, causing expansion ofthe gauge which is resisted by the setting concrete. There

is no similar decrease in load as the concrete cools which indicates that the gauge and the

concrete are cooling and contracting at the same rate. (It will be shown later in Section 4-

5-4 that the bending moments measured in the piles at the level of the base slab reduce at

the time of Base Slab Construction.)

108

Figure 4-12 shows the typical relationship between temperature and 'load' indicted by the

measured strain for a base slab gauge over the period of the concrete pour and after

temperature stabilization. There is no obvious relationship between load and temperature

for the period after temperature stabilization, and it is not possible to determine whether

the compressive strain indicated by the gauge is due to thermal effects or load being taken

up by the slab.

The temperature fluctuations experienced by the gauges in the base slab are small because

the slab is in contact with the ground at the bottom of the excavation which is at a

reasonably constant temperature (being generally out of direct sunlight).

Figure 4-13 shows the strain measured in the separate gauges for sets 1 to 4, where

compressive strain is positive. The strain behaviour is different in all sets and it is difficult

to distinguish a general pattern. Figure 4-13 shows that after ballast was placed on the base

slab between Days 998-1028 (July 2002), the daily fluctuations in temperature recorded

by the strain gauges became negligible.

4-4-3 Reinforced concrete prop data

When concrete cures it undergoes shrinkage due to the hydration process, and when load

is applied it is also subject to creep2. Analysis of reinforced concrete (RC) prop strain data

is therefore complicated because the strain measured is a combination of the strain due to

shrinkage, strain due to the applied load and strain due to the effects of creep. To illustrate

the possible magnitude of strain occurring due to creep effects, Neville (1970) states that a

typical creep deformation after a year under load is three times the deformation under that

load.

Concrete undergoes continual shrinkage as it ages, with the majority of this shrinkage

occurring in the first 28 days after the concrete is poured. It is possible to estimate the

amount of strain occurring due to shrinkage if measurements of strain are taken in the

immediate aftermath of the RC prop concrete pour and before any load is applied to the

prop. Any strain occurring during this period is due to shrinkage alone and therefore the

relationship between shrinkage and time can easily be established. A curve fitted to even

2 Creep is deformation occurring under, and induced by, a constant sustained stress (or applied load).

109

only a few weeks' of such data can provide a sound basis for estimating the strain due to

shrinkage that has occurred after any time period, even after load is applied to the prop.

The shrinkage due to creep is more difficult to quantify. Creep occurs when the strain

continues to increase after the application of load. This indicates that the stiffness, Ecollc • is

decreasing. It is possible to estimate the amount of creep that has OCCUlTed through

analysis of the data collected in the long-term, when the applied load is considered to be

relatively constant.

Calculation of strain due to shrinkage

The reinforced concrete prop strain data was not collected until 9 days after the last RC

prop concrete was poured, and it was then only another 7 days before Excavation Phase 1

commenced. It is therefore difficult to determine from the strain gauge data the rate of

shrinkage due to concrete curing and how much of the final shrinkage had occurred before

load was applied to the props. The strains measured in the concrete props over the period

of excavation (using for the zero strain reading the average of the readings taken over the

24 hours immediately before excavation under the props began) are shown in Figures

4-14,4-15 and 4-16 for props PI, P4 and P2 respectively. It appears from these figures

that the strain is increasing at a reasonably regular rate and the changes due to construction

events are difficult to identify.

The strains plotted in Figures 4-14 to 4-16 are a combination of strain due to shrinkage

and loading. To establish the effects of shrinkage it was proposed to cast a short section of

'prop' using the same concrete mix design and reinforcement detail, and to embed a

number of vibrating-wire strain gauges in it. Owing to site constraints this was not

possible, however it was possible to analyse strain data collected from RC props (with a

similar concrete mix and dimensions) cast elsewhere on site to estimate the concrete

shrinkage. These props were at chainage 89+205, approximately 700 m away from the

instrumented section.

Ross (1937) proposed the hyperbolic equation for concrete shrinkage given in Equation 4-

12, where Esh(t) is the shrinkage at time t and a and b are constants found from

experimental data, being the y-intercept and slope respectively of the linear relationship

between t I ESh(t) and time. This relationship was deduced for three instrumented props at

chainage 89+205. Strains in these props were monitored (using identical vibrating-wire

embedment gauges to those used at the instrumented section) throughout the concrete pour

110

and the period before excavation; a period of at least 28 days. It should be noted that at the

instrumented section the RC prop concrete was poured in January when temperatures were

3-8 DC, while at chainage 89+205 the concrete was poured in July when daily

temperatures ranged between 6-28 DC (measured by the thermistors in the RC prop

vibrating-wire gauges). The rate of curing and shrinkage is likely to be affected by the

ambient temperature at the time of the concrete pour and during curing, however the

resultant total shrinkage is considered to be independent of temperature.

t &s,,(t)=--·

a + bt Equation 4-12

Figure 4-17 shows a graph of t / &sh(t) against time for a typical strain response gauge at

chainage 89+205. The values for a and b calculated for all the gauges at chainage 89+205

are presented in Table 4-4, together with the maximum strain due to shrinkage indicated

by the data from each gauge using Equation 4-12.

Prop gauge a b r t 1 (. . ) Im--=- mlcrostram HOC! a+bt b

1 23170·8 4235·3 236·1

2 21346·9 4756·33 210·2 1 (Ch 89+205)

3 15583·5 5238·6 190·9

4 28178·6 4626·79 216·1

1 26562·5 4676·08 213·9

2 19539·1 4624·83 216·2 2 (Ch 89+205)

3 18549·1 4685·94 213·4

4 22791·9 4389-48 227·8

1 17151·9 4683·19 213·5

2 17198·8 4474·94 223·5 3 (Ch 89+205)

3 13626·8 5113·68 195·6

4 15783·0 4551·24 219·7

average 215 /-!E

Table 4-4: Values of a and b calculated for all gauges at chainage 89+205 m

The strain predicted by Equation 4-12 is compared with the measured strains (for the same

data plotted in Figure 4-17) in Figure 4-18. Using Equation 4-12, the strain measurements

from the RC prop gauges at chainage 89+205 predict that 215 /-!E of strain (with a range of

111

191-236 Ilf:) would have occurred in the long-term due to shrinkage, with 95% of this

strain occurring within 85 days of the props being cast. At the instrumented section this

was Day 552 (11 th April 2001); 15 days after Excavation Phase 2 and 27 days before Base

Slab Construction.

In Table 4-5 the following details are listed:

1. the length of time between construction and excavation directly underneath for the

props in the instrumented section,

2. the amount of shrinkage strain that would have occurred during excavation under

the props (calculated from the analysis on the props at chainage 89+205) and

3. the equivalent load that this strain would indicate if a correction was not made.

Prop PI P2 P4

Day constructed 465 465 467

Day Excavation Phase 1 began directly 483 487 505

underneath prop (see Figure 3-42)

Number of days between prop construction 18 22 38

and excavation directly underneath

Average strain at chainage 89+205 after this 160 166 202

length of time, Ilf:

Load indicated by this strain, kN 3989 4149 5049

Amount of shrinkage strain occurring in prop after excavation began directly 55 49 13 underneath, Ilf:

Equivalent load, kN 1379 1218 318

Table 4-5: amount of strain occurring due to shrinkage and equivalent load indicated by this strain for the props at the instrumented section

The calculations in Table 4-5 clearly indicate that if a correction is not made for concrete

shrinkage in cases where load is applied soon after prop construction, prop loads may be

significantly overestimated. Figure 4-19 shows the RC prop loads calculated using the

strains in Figures 4-14 to 4-16 and Figure 4-20 shows the loads calculated by reducing the

measured strains according to the analysis carried out on the readings from chainage

89+205. There are several points to make about these graphs regarding the shrinkage

correction:

112

1. The correction for shrinkage reduces the prop load by approximately 800 kN by

Day 552 (after which 95% of the shrinkage strain will have occurred).

2. The correction appears to have reduced the amount by which the readings increase

during periods when load is (apparently) not being applied, however there are still

periods where the load increases more than expected: during Excavation Phase 1

(particularly in PI and P2); before Excavation Phase 2; and after the temporary

props are removed.

Figure 4-21 shows the RC prop load data only up to the end of Excavation Phase 2. On

this graph the periods during Excavation Phase 1 when material directly under the RC

props was removed are indicated, and the corresponding increases in load are clear.

Another point of note is that after Tl was installed, the RC prop loads appear to decrease

when the temperature increases, and visa versa. This effect is particularly prominent at the

positions indicated by IT] and [2]. This may be because as the temporary props expand

with an increase in temperature the walls are pushed into the soil and the load in the RC

props is therefore reduced. It is also likely that because load is now being applied to the

RC props, creep is beginning to have an effect.

Calculation of strain due to creep

Equations representing the change in stiffness with time (due to the effects of creep) for

each prop have been estimated by assuming that the prop loads do not change after Day

607 (other than for seasonal and daily temperature variation). Day 607 was chosen

because it is 28 days after the base slab was poured and after the temporary props were

removed (on days 581 (Tl), 595 (T2 and T3) and 600 (T4)), and therefore it is assumed

there is no further change in load due to the preceding construction events. To remove the

effect of seasonal and daily variations, the load in the props has been assumed to be the

same as that recorded at 9 am on Day 607 as at the same time on the same day every year

following, i.e. on Days 972, 1337, 1702 and 2067. The variation in temperature recorded

at 9 am on these days by the thermistors in the gauges was only 3°C.

The Young's modulus, Econc, was found on each of these days using Equation 4-4,

assuming that the initial stiffness was 25 MPa. Equations with the form given in Equation

4-13, where N is the number of days and A, B, and C are constants, were then found (using

the least squares fitting method) to these data for each prop - see Figure 4-22. The

calculated values of A, Band C which provide the best-fit curves to the prop stiffness/time

113

relationship are given in Table 4-6. (Exponential expressions have commonly been used to

model creep behaviour in the past: see Neville, 1970.)

-N -

Econc = Ae B + C . Equation 4-13

PI P2 P4

A 12·756 12·971 14'741

B 12·222 11·981 10·222

C 558·12 816·98 606·20

Table 4-6: Values of constants for best-fit line to stiffness/time relationship (to 5 significant figures)

These equations were then used to calculate the stiffness used in the calculation of the RC

prop load from Day 607 onwards. As data collection is ongoing, in the future it will be

possible to get a better estimate of the creep behaviour.

Because the correction for creep has only been applied to the measurements taken after

Day 607, it only affects the long-term data, presented in Chapter 6. An adjustment for

creep should really be made as soon as load is applied, and therefore the reported RC prop

loads may be over-estimated (however the value of the initial Young's modulus is also an

estimate - see Section 4-3-2). In the first year after Day 607, 39 /-lE can be attributed to

creep. This is equivalent to 974 kN ofload, calculated using a Young's modulus of25

MPa. It was 4 months (or a third of this time) between the time excavation began under

the props (i.e. load began to be applied to the props) and Day 607. Although the strain due

to creep occurs at its fastest rate when load is applied, the load was initially small.

Therefore the creep occurring before the correction was made can be estimated to be a

maximum of 500 kN.

Effect of changes in temperature

The temperature versus load data for a typical RC prop strain gauge during periods when

no construction activities are taking place is shown in Figure 4-23. This figure shows that

there is no clear relationship between prop load and temperature which can be used for

correcting the daily and seasonal temperature induced fluctuation in load. However, the

RC props are not affected by temperature to the same degree as the temporary props and

are therefore more straightforward to analyse.

114

Further comments

There are a number of further comments that can be made regarding the reinforced

concrete prop data.

1. The load measured in prop P4 is significantly lower than that measured in props PI

and P2. The reason for this is uncertain, but may be due to anomalies in the

construction detailing such as a lack of fit at the prop wall interface.

2. The daily fluctuations in prop load in P4 are smaller than in props PI and P1. This

may be caused by variations in the concrete cover to the gauges, making them

more thermally insulated. This might have occurred if either P4 is larger than the

other props (which is unlikely but would also explain the smaller load recorded in

this prop as a larger surface area would indicate a larger load) or the gauges are

embedded further into the prop. Smaller strains would be expected to be measured

in either ofthese cases, as the gauges are proportionally closer to the neutral axis

of the prop.

3. A reduction in load was recorded in all 3 RC props on the removal of temporary

prop Tl, whereas an increase load in props PI and P2 was recorded when props T2

and T3 were removed, and similarly an increase was recorded in prop P4 when

prop T4 was removed. The base slab concrete reached its maximum strain and

temperature on the same day as Tl was removed, resulting in the decrease in load

measured by the other RC props, and masking the effect of the temporary prop

removal.

4-4-4 Pile data

Theoretically, if the neutral axis is in the middle of a pile section and there is no axial load

the bending strains at the front and back of the pile at any elevation will be equal and

opposite. In reality, as well as the self-weight of the pile and the capping beam, axial load

in the pile will be caused by wall friction on the front and back faces of the piled wall.

Figure 4-24 shows the difference in strain measured between the front and back of the pile

at each gauge pair elevation against time for piles X and Y (see Figure 3-29 for gauge

elevations). Figure 4-24 shows that for most gauge pairs the difference in strain between

the front and back ofthe pile is small. However, the values increase suddenly for gauge

pairs 15&16 and 17&18 in pile X at the start of Excavation Phase 2, and for 21&22 in pile

115

Y after Temporary Prop Removal. Figure 4-25 shows the strain measurements for these

individual gauges. Figure 4-25(a) shows that during the Excavation Phase 2 much larger

strains are measured by gauges X16 and X18 (even numbered gauges are at the front of

the pile) than by their partners, gauges X15 and X17 respectively. Figure 4-25(b) shows

that during Temporary Prop Removal a larger strain is measured by gauge Y22 than Y21.

These changes in strain are likely to be caused by to cracking in the piles. Figure 4-25(a)

also shows that following cracking of the pile during Excavation Phase 2, the response to

Temporary Prop Removal is much larger in the gauges at the front of the wall than the

gauges at the back.

Pile deflections measured with the inclinometer in Pile Z around the time of Excavation

Phase 2 and Temporary Prop Removal are shown in Figure 4-26. The elevations of gauges

15& 16, 17& 18 and 21 &22 are indicated. The figure shows that the points at which the

strains have increased to a greater degree at the front of the pile than the back of the pile

correspond with the points where the maximum deflections over the period have been

measured.

Table 4-7 lists the wall deflections and curvatures measured by the inclinometer for the

gauges at which the largest values were recorded before and after the two construction

activities that appeared to cause cracking (temporary prop removal and excavation phase

2). The changes in the values over these periods are included. The values relating to the

positions at which cracks occurred have been highlighted. As expected, the data show that

cracks occurred at positions where the largest changes in deflection and curvature

occurred, and indicate that the change may be more important than the degree of

deflection and curvature in initiating cracks. However, cracks do not appear to have

formed at the position of gauges 13&14 during Excavation Phase 2 and at gauges 23&24

during Temporary Prop Removal, where the changes in curvature that occurred were at

least as high as values at other locations where cracks did form. It is therefore likely that a

combination of the changes in deflection and curvature affects the propensity for cracking.

As previously noted, the bending moment in the pile may be affected by local changes in

the pile's cross-sectional area, which affects the flexural rigidity, EI, of the pile. Figure

4-27 shows a bulge in the pile just underneath the bottom of the casing used during

installation, indicating that overbreak has occurred at this level during excavation for the

piles, and the resulting cavity was consequently filled with concrete. This level

corresponds approximately with the level of gauges 17& 18 and 19&20, which indicate a

116

smaller bending moment than those calculated at the locations of gauges 15& 16 and

21&22 in the stages before cracking occurred, shown in Figure 4-29 (see next section).

This excess concrete was removed after construction was completed, some time between

August and October 2001.

Gauges 13&14 15&16 17&18 19&20 21&22 23&24

Elevation, m AOD 32·5 34 35·5 37 38·5 40

8, pre Exc. Phase 2 (Day 529) 7·5 8·7 9·7 10·3 10·6 10·8

8, post Exc. Phase 2 (Day 540) 16·9 18·6 19·1 18·8 18·1 17·3

Change in 8 9·4 9·9 9·4 8·5 7·5 6·5

8, pre TP Removal (Day 564) 18·3 20·2 20·7 20-4 19·5 18·6

8, post TP Removal (Day 617) 21·7 24·3 25·9 26·5 26·1 25·1

Change in 8 3·4 4·1 5·2 6·1 6·6 6·5

K, pre Exc. Phase 2 -0·065 -0·099 -0·125 -0·139 -0·139 -0·119

K, post Exc. Phase 2 -0·306 -0·344 -0·348 -0·312 -0·232 -0·102

Change in K -0·241 -0·245 -0·223 -0·173 -0·093 0·017

K, pre TP Removal -0·3 -0·343 -0·353 -0·321 -0·24 -0·101

K, post TP Removal -0·289 -0·402 -0·459 -0-446 -0·366 -0·231

Change in K 0·011 -0·059 -0·106 -0·125 -0·126 -0·13

Table 4-7: Changes in deflection, 8, (mm) (relative to the toe) and changes in curvature, K, ( X103m-l) (from the 5th order polynomial curve fit) over periods of pile

cracking. Positions where cracks have occurred are highlighted.

Figure 4-28 shows strains measured in Pile Y (see Figure 3-29 for the gauge elevations)

after excavation and before Base Slab Construction; (Days 542-571) therefore the ground

was at the level of the bottom of the base slab, just below gauges 15& 16. Over this period

the average temperatures measured in the temporary props ranged from 1 to 21°C, with an

average of 10°C. The figure shows that there is a linear relationship between temperature

and strain for the gauges at the front of the pile (odd numbers). The temperatures in these

gauges vary by 6°C; this was 30% ofthe temperature variation measured by the

temporary prop thermistors over the same period. This caused the strain to vary by

approximately 30-40 j.l8, which would indicate 66-89 kNm of bending moment. There is

no clear relationship with temperature for the gauges at the back of the pile, although the

temperature range experienced by these is much smaller due to the insulation provided by

the concrete and soil. Below excavation level the gauges at the front and back of the pile

are equally affected by temperature. Gauges 13&14 are affected slightly, gauges 11 & 12

117

even less so, and below this there is no relationship between strain and temperature. From

this, ambient changes in temperature can be assumed to only affect gauges within 2·5 m of

the excavation or ground level (excavation level minus the level of gauges 11 & 12).

4-4-5 Comparison between pile bending moments calculated/rom strain gauges and

inclinometer

Figures 4-29 and Figure 4-30 show bending moments calculated from the strain gauge and

inclinometer measurements at four instances; before and after Excavation Phase 2 (when

cracking occurred in Pile X: Figure 4-29) and before and after Temporary Prop Removal

(when cracking occurred in Pile Y: Figure 4-30). Data at points where cracking has been

shown to have occurred (in Figure 4-24 and Figure 4-2S) are highlighted on Figures 4-

29(b) and 4-30(b). Bending moments calculated from the Sth and 6th order polynomial

curve fits to the inclinometer data are shown. The upper and lower bounds for the cracking

moments, Mer (calculated in Section 4-3-3) are indicated on the figures. E1 values are all

based on uncracked concrete properties.

In Figure 4-29(a) and (b) there is close agreement between the bending moment data

derived from the wall profile and the strain gauges. In Figure 4-29(a) the measured

bending moments are all smaller than Men and there is little difference between the

bending moments calculated from the Sth and 6th order fits to the inclinometer data.

Figures 4-29(b) and 4-30(a) show that at some strain gauge locations the calculated

bending moments have exceeded Mer. The two highest of these are those where cracking

has been observed. At the other locations the section may be cracked but not specifically

at the strain gauge (the strain gauges are 140 mm long and at l'S m intervals). In general

there is close agreement between bending moments calculated from inclinometer readings

and strain gauge measurements for the middle section of the pile. However, at the pile toe

and the pile head the curves bear away from the strain gauge measurements. This is

obviously a mathematical problem with the curve fitting process, and is caused by

attempting to fit a polynomial curve to a wall profile which contains straight sections (the

bottom of the wall is straight below approximately 27·S m AOD - this was observed by

noting that the gradient of the pile is constant below 31· S m AOD, and that the strain

gauges measure zero bending moments below about 27·0 m AOD). Further analysis has

shown that polynomials fitted only to the curved parts of the pile (i.e. leaving out the

lowest few metres) produce very similar bending moment profiles but with the top and

118

bottom parts absent. It follows that the bending moments calculated for known straight

sections of the pile can simply be ignored.

Figure 4-30(b) shows the bending moment profiles found after Base Slab Construction and

Temporary Prop Removal. Again, at the pile toe and top the inclinometer derived bending

moments bear away from the strain gauge measurements. In Figure 4-30(b), unlike Figure

4-30(a), the inclinometer derived bending moments are larger than those measured using

the strain gauges around the position of the maximum bending moment.

4-5 Measured changes due to construction events

4-5-1 Temporary props

Figure 4-31 shows the individual temporary prop loads. It can be seen that approximately

800 kN ofload was recorded in temporary props T1 and T3 as a result of Excavation

Phase 2, and 1000 kN in T2. When the base slab in bay A3 (see Figure 3-42) was

constructed the load in the nearest temporary prop to it, T1, reduced by approximately

200 kN over the following 3 days. This is because the base slab concrete applied a lateral

load on the wall as a result of the weight of the wet concrete and the initial expansion of

the concrete due to curing. The combined load and expansion of the slab forced the wall

back into the soil for a short period, inducing a temporary decrease in the temporary prop

loads (in the piles a reduction in bending moment is measured in the gauges around the

level of the base slab when the base slab concrete is poured against them, see Section 4-5-

4). After the initial expansion of wet concrete it contracts while cooling and setting, and

therefore the load in the temporary props recovered to close to the original value over the

following 5 days, until the next base slab, at the instrumented section, was poured. This

caused the load in prop T1 to reduce by nearly 100 kN over the following 2 days, until it

was removed.

Temporary props T2 and T3 exhibited no change in load due to the installation of base

slab A3. Following construction of base slab A4, the base slab in the instrumented section

(i.e. directly below props T2 and T3), a reduction of approximately 200 kN was recorded

in both T2 and T3 over the following 2 days. Then only 2 days after casting slab A4, prop

T1 was removed and the load in prop T2 rose rapidly by approximately 200 kN. The load

in prop T2 continued to rise by approximately another 200 kN over the following 14 days,

119

until it was removed. There was no change in load in prop T2 when the next base slab, A5,

was constructed.

After base slab A4 was poured the load in prop T3 increased slowly over the following 9

days, by a total of approximately 80 kN, until the next base slab, A5, was installed. At this

point the load reduced by approximately 40 kN. The load in prop T3 was then reasonably

steady over the following 5 days until it was removed along with prop T2.

It is important to note that the changes in loads and bending moments due to the removal

of prop T1 are masked by the loads applied by the expansion and contraction of the base

slab concrete poured only 2 days earlier. Therefore the removal of props T2 and T3 might

be expected to have different consequences.

4-5-2 Base slab data

Although the temperatures indicated by the base slab gauges during the concrete pour

were very similar, Figures 4-11 and 4-13 shows that the strain rates were different. The

strain in set 1 rose much more quickly than the strains in set 4, and the strains in sets 2 and

3 rose at a very similar rate. Despite their different rates of strain during the concrete pour

and curing, sets 1-3 indicated a similar 'load' by Day 586 and set 4 by Day 596. The

difference in strain may be affected by the different frictional qualities of the boundaries to

the base slab: the previous concrete slab acts as a boundary near set 1 and there is

shuttering near set 4.

4-5-3 Reinforced concrete props

The changes in the reinforced concrete prop loads that occurred as a result of excavation

directly under the individual instrumented props are listed in Table 4-8 (also see Figure 4-

21). The loads measured at various construction stages are presented in Table 4-9.

120

Change due to excavation under:

PI P2 P4

PI 550 200 0

P2 400 500 0

P4 150 150 150

Table 4-8: Changes in RC prop load due to excavation directly under individual props

Load after excavation

Load after Load after construction was Temporary completed

Prop I year 2 years 3 years Removal

PI 1600 2700 3800 4500 5200

P2 1800 3100 4200 4900 6600

P4 900 1900 2700 3300 3900

Table 4-9: Load at specific stages of construction

4-5-4 Pile bending moments measured with strain gauges

Figures 4-32 and 4-33 show bending moments measured in Pile Y against time for the

gauges below and above the base slab respectively. For clarity only the data taken at 0900

each day is plotted to remove the fluctuations due to daily temperature changes. Figures

4-34 and 4-35 show similar data for Pile X.

Table 4-10 shows the different changes in bending moment that were experienced by

different parts of the pile as a result of specific construction activities. The information in

the table is representative of both piles, as they behaved similarly (although the

magnitudes of some of the changes differ).

Analysis ofthe data shows that the largest changes in bending moments occurred during

Excavation Phase 2, at the level of the base slab. This corresponds with the largest

changes in deflection and curvature shown in Table 4-6. Reductions in bending moment

were recorded during this construction event in the gauges near the top of the wall; this is

because rotation occurs around the temporary prop, the point of maximum bending moves

down the wall and therefore the top of the wall is under less stress (this rotation can be

seen in Figure 4-6 (a) and (b)). Before Excavation Phase 2, the maximum bending moment

was approximately at excavated ground level, i.e. about the location of the temporary prop

121

(gauges 21 &22). After Excavation Phase 2, the point of maximum bending moment was

approximately at formation (gauges 15& 16). The reduction in bending moments at the top

of the wall could also be partly due to redistribution of stresses in the wall resulting from

the cracking induced at this stage of construction.

The expansion and contraction of the base slab during its construction and the effect this

had on the wall induced a temporary decrease in the bending moments close to the level of

the base slab. The gauges near the top of the wall were not affected, probably because of

the presence of the temporary props, which restricted movement of the wall and therefore

load redistribution above its position.

As the heat generated in the base slab during the curing process dissipated the concrete

contracted, therefore the load being applied to the wall reduced and the bending moments

near the level of the base slab increased, as shown in Figures 4-32, 4-33, 4-34 and 4-35.

However, temporary prop T1 was removed only 2 days after the base slab was cast, and

therefore the change in bending moment is a combination of the loads occurring due to the

concrete setting and temporary prop removal. On removal of props T2 and T3, the bending

moment near the level of the base slab decreased. Therefore it is reasonable to assume that

the removal oftemporary prop T1 would also initiate a decrease in bending moment near

the level of the base slab, and if prop T1 had not been removed so soon following Base

Slab Construction the increase in bending moment in the following days may have been

substantially higher.

Temporary prop removal resulted in an increase in the bending moments in the top of the

wall and a decrease in those near the level of the base slab. The inclinometer

measurements in Figure 4-6 (d) show that the point of maximum deflection moved up the

wall after temporary prop removal (the last temporary props, T2 and T3, were removed on

Day 595).

122

....... tv w

Stage Top section of wall Middle section of wall Bottom of wall Max change Notes:

Exc. Phase 1 11&12-25&26: not applicable 1 &2-9& 10: decrease Increase at 21 &22 Approx. the same change increase recorded in 15& 16-19&20 and

23&24 due to second part of Exc. Phase 1.

Exc. Phase 2 21&22-25&26: 5&6-19&20: increase 1&2-3&4: no significant Increase at 15 & 16 Trend of changes generally decrease change continues for about 2 days after

Exc. Phase 2 has finished.

Base Slab 21&22-25&26: no 11&12-19&20: 1&2-9&10: small Decrease at 15&16 Construction significant change reduction increase/no significant

change

Removal of 23&24-25&26: small 13&14-21&22: slow 1&2-11&12: slow Increase at 15& 16 Tl increases over period increase over period of decrease over about 10

of 10 days/no about 10 days days. No significant significant change change in 1&2 and 3&4

Removal of 19&20-25&26: 17&18: no change; 1&2-7&8: no significant Increase at 21 &22 T2 and T3 increase (no change in 15&16-9&10: change

25&26 pile X) decrease -- -- --- - -- -

Table 4-10: Changes in bending moment for piles X and Y due to particular construction stages. Ranges are inclusive of gauge numbers quoted

4-5-5 Inclinometer data

The inclinometer readings illustrated in Figure 4-6 show that the wall moved consistently

into the excavation as construction progressed and that the majority of the movement

occurred during excavation phases 1 and 2. Further movement occurred at temporary prop

removal. The readings indicate that after temporary prop removal the wall continued to

move, with the top of the wall moving into the excavation by 4 mm over the following 1 1

months. However, Figure 4-7 shows that the movement was not in one direction and that

few readings were taken over a relatively long period of time, so the accuracy of these

measurements is questionable. It is possible that the wall movements may be affected by

yearly temperature changes, however the paucity of the readings prevents a definitive

conclusion.

As previously mentioned, the relatively high deflections measured at the top of the wall

lead to the conclusion that the cutting has swayed due to the difference in ground level

behind the north and south walls (see Figure 3-22). (N.B: A compression of 1 mm in the

RC prop would produce 25 000 kN of load; therefore the movement could not be due to

compressive loading.) Unfortunately it was not possible to monitor the south wall due to

access restrictions and very little is known about the progress of the construction 0 f the

'Maidstone loop', the line which was constructed behind the south wall, and the

corresponding changes in ground level. However from photographic evidence (Figure 3-

34 - taken 110 days before excavation activities began) the existing railway lines can be

seen and preparations for construction are being made, therefore the ground level must

have been reduced by this time.

In the early part of Excavation Phase 1 (Days 483-488: before work temporarily stopped)

the wall appears to be well propped at the top with excavation causing the wall to bend in

the middle. Figures 4-32, 4-33, 4-34 and 4-35 show that the bending moments induced at

this time were small. Later during Excavation Phase 1, the changes in deflection measured

at all points in the section of wall above temporary prop level were similar, indicating that

the wall was rotating and that the RC prop was not providing significant restraint. Figures

4-33 and 4-35 show that bending moments did not increase in either pile until the end of

Excavation Phase 1. The top of the wall moved approximately 7 mm during Excavation

Phase 1 and a further 3 mm before the start of Excavation Phase 2.

124

Between Days 530 and 533, at the beginning of Excavation Phase 2, the top ofthe wall

moved very little while lower down the wall the measured change in deflection was

approximately 5 mm. Days 534/535 were a weekend, after which excavation appears to

have produced similar changes in deflection in the top section of the wall, indicating little

restraint in the prop. The readings taken during Temporary Prop Removal also show the

whole top section of the wall moving.

4-6 Conclusions

At the CTRL site at Ashford, UK, wall deflection profiles have been found from

inclinometer measurements, inclinometer measurements and strain gauges have been used

to derive bending moments in a propped contiguous bored pile retaining wall, and strains

gauge measurements have also been used to measure temporary and reinforced concrete

prop loads. The following points can be made:

1. Use of vibrating-wire instruments and a datalogging system capable of taking

readings at up to 5 minute intervals has allowed (effectively) continuous data to be

collected on site at Ashford.

2. The implementation of rigorous procedures during data collection, thorough

analysis and error checking has produced good wall displacement profiles from the

inclinometer data.

3. It was not possible to infer load measurements from strain data collected in the

base slab, because the strains caused by changes in load could not be distinguished

from the strains caused by thermal effects arising from the concrete curing process.

4. The temporary prop data proved to be more straight-forward to analyse and was

corrected for temperature effects so that the average prop loads occurring due to

construction events could be determined.

5. Analysis of concrete shrinkage occurring in reinforced concrete props constructed

elsewhere at the Channel Tunnel Rail Link site has allowed prop loads at the

instrumented section to be estimated, despite the fact that excavation under the

props occurred very soon after the concrete pour while considerable shrinkage was

occurrmg.

125

6. Careful consideration ofthe mathematical restraints to curve fitting have allowed

wall bending moment profiles to be determined from inclinometer readings, and

these show good agreement with bending moment profiles found from vibrating­

wire strain gauges. Identification of concreting cracking and an analysis of the

effect this has had on the bending moment profiles calculated from the

inclinometer measurements has explained the differences in bending moment plots

found by the two different methods after cracking had occurred.

126

E E C) c

= ::s CJ 0 ... . 5 ... c (1)

E (1)

> 0 E (1)

.s cu 3:

E E Cl c

= ::s CJ 0 ... c ... c (1)

E (1)

> 0 E (1)

.s cu 3:

2

Toe deflection = 0.0149 x Time in days - 7.0649

1.5 :-- Previous base slab construction (A3) .. -- Base slab construction (A4) j-- Next base slab construction (A5)

1

0.5

0 -.::t N 0 0 00 co -.::t N 0 00 co -.::t N ...... -.::t I'-- ...... (\') co 0) N LO I'-- 0 (\') co LO LO LO I'-- I'-- I'-- I'-- 00 00 00 0) 0) m

-0.5 Time in days

-1

(a)

2 y = 0.0149x - 7.0649

Exc. Ph. 1A Exc. Ph. 1B 1.5 ~:

::5-;

1

0.5

0 0) co (\') 0 I'-- -.::t ...... 00 LO N 0) co (\') 0 0 ...... N (\') (\') -.::t LO LO co I'-- I'-- 00 0) 0 LO LO LO LO LO LO LO LO LO LO LO LO LO co

-0.5 Time in days

-1

(b)

Figure 4-3: Movement of wall toe with time measured by the inclinometer (a) construction period and the following year (b) construction period only

128

N -0

c o « E CI)

> CI)

...J

-1

Readings taken on days 488,489,491

45

40

35

30

1

o

Readings taken on days 496,497,498,

501,502,503

1 2 3 Deflection, mm

(a)

4

c o « E

~ CI)

...J

-1

Readings taken on days

488,489,491

45

40

35

30

25

20

15

o

Readings taken on days 496,497,498,

501,502,503

Wall toe

1 2 3 Deflection, mm

(b)

4

c o « E

~ CI) ...J

-1

Readings taken on days

488,489,491

45 T r.

40

35

30

25

20

15

Readings taken on days 496,497,498,

501,502,503

~

Wall toe

o 1 2 3 Deflection, mm

(c)

Figure 4-4: Correction of inclinometer data: (a) As read, (b) adjusted for zero movement at the toe and (c) adjusted for movement at toe indicated by equation given in Figure 4-3

4

,..... w o

45

40

35 +

c 0 30 « E CI)

> 25 CI) ..J

20

15

-5 o

586

"M\~617 .

Exc. Phase "1" "Ievei

Exc. Phase "2 "Ievei 718

Wall toe

5 10 15 20 25 30 Deflection, mm

c 0 « E

CD > CI)

..J

28

26

634~~592

617

-5

2

2 JU !~ 585

14

o 5 Deflection, mm

(a) (b)

Wall toe

10 15

Figure 4-5: (a) Measured deflected profile of the wall and the bottom 10 m of tubing installed in the underlying soil. (b) Excerpt of (a) showing deflection ofthe bottom 10 m of inclinometer tubing

Days 496, Days 501, 497,498 502,503 Days 516,

45 / ~519,522 Days

488'40 489,

W-~~~--Day504

491

g 35 « E Q)

~ 30 ..J

25

,~-!H---Day 508

Days 510, ~511,512

Day 523

Wall toe

-5 0 5 10 15 20 25 30 Deflection, mm

45

40

(a)

Days 545,550,552, 558,560,564,566

~

Phase 1 o excavatiori"ievel o 35 Phase 2 « E excavation'Ievel

~ ..J 30

25 Wall toe

Days 580,581

-5 0 5 10 15 20 25 30 Deflection, mm

(c)

Days 524,525, 526,529,530

45 I Days 536,

/537,538, " 539,540

40

g 35 « E Q)

~ 30 ..J

25

Days 532 533

Wall toe

Phase 1 "exe, level

Phase 2 '''exe, level

-5 0 5 10 15 20 25 30

45

40

Deflection, mm

(b)

Days 718,719,747, 777, 862, 945

Days 585 -

Days 592--

Phase 1 excavatiori"ievel

00 35 « Phase 2 E excavation 'level

~ ..J 30

25 Wall toe

-5 0 5 10 15 20 25 30 Deflection, mm

(d)

Figure 4-6: Error adjusted inclinometer measurements taken during construction (2001): (a) around Excavation Phase 1 (Days 483-488), (b) around Excavation Phase

2 (Days 530-537), (c) around Base Slab Construction (Day 579) and (d) after Base Slab Construction

131

E E

25

C) c E 20 ~ (.)

.s c

E E C) C

E ~ (.)

0 .... c

(ij ~ -0 e. 0 .... -0 .... c CI)

E CI)

> 0

::E

0

25

20

15

10

5

0

-5

1.() ('f) ...... ...... oq- I'-oq- oq- oq-

0) I'-0) N oq- 1.()

t ro -00

...... ~ 0... ci x W

c: o • t5 ~ ..... -00 c: o ()

.0 JQ 00 ,: (1) " 00 :: ro " co ::

1.() ('f) 1.() ex) 1.() 1.()

a. 0 -00

...... ; .'

.!:: 0...: ci: x' w;

Removal of temporary prop T1

Removal of temporary props T2 and T3

...... 0) I'- 1.() ('f) ...... 0) I'- 1.() ('f) ...... 0) I'- 1.() ...... ('f) <0 0) N 1.() I'- 0 ('f) <0 0) ...... oq- I'-<0 <0 <0 <0 I'- I'- I'- ex) ex) ex) ex) 0) 0) 0)

Time in days

(a)

t ro t - "0 "0 00 ro ~

c: - c: "0 (1) 00 (1) (1)

N; N > ...... ...... 0

~ .' ~ E .!: .!: : "0:

0... 0... 0...: 0... (1): (1) ..... ..... ,

ci ci ci: ~, ...... x x x' 0: I-w w w; a.: a.

.0: 0 ro: ..... w; a. (1)' a. 00: E ro: (1)

co: l-

<0 oq- N 0 0 ('f) <0 0) 1.() 1.() 1.() 1.()

Time in days

(b)

Figure 4-7: Movement of the top of the wall (a) all measurements (b) excerpt of (a) showing construction phases only

132

2500

2000

z 1500 JII:

" CIS o

...J 1000

500

y = 44.439x + 1522.5

O+--r--.-~-'--+--r--.-~-'--+--r--.-~-'--+--r--.-~-'~

o 5 10 Temperature, °c

15 20

Figure 4-8: Typical relationship between load and temperature for temporary prop data taken over the 30 days between Excavation Phases 1 and 2, when no

construction activities were taking place.

133

....... w +:>.

2400

2200

2000

1800

1600

1400 +

i 1200 I .3 1000

800

600 ~

400

200

0 5

-200

t: co. en: ('\I: Q). en· co·

.£: • a...: 0: X· w:

T1

Temperature

-0; c· Q):

('\I.

Q): en. co·

.£:. a...:

535 542 549 556 563

Time in days

.0: co· Cii: Q)' en' co '-0 .0 :~ en':J :J'O

.Q :0. >'''-''' Q) '('1') ...... « a... :.......-

570

Figure 4-9: As measured temporary prop loads

40

30

20

10

0

-10 0 0

-20 ~ :::s -ca ~

-30 C1l Q.

E C1l

-40 ~

-50 .0'

'<to co: «: :-0 Cii'"O

T3 + -60 .0:

.Q) Q) : ~

. > en':J co . . 0 CO ' o Cii '"0 • E .0:0. + -70 Q) : ~ : Q) xfo en ' :J . ..... co ' 0 . T""" Q).« ca :o.:1- z.....-

I I I I I I I I I I I ~~8-80 577 584 591

-90

....... w VI

2400

2200

2000

1800

1600

1400

~ 1200 '0 co .3 1000

800

600

400

200

0 5

-200

t'

~ ~ N ' w: en, cu'

..c ' 0; 0 : x , w '

Temperature

-0 ' c ' w: N: w' en' cu '

..c ' 0.:

~T3 535 542 549 556 563

Time in days

.0 cu 00 : w ' en' cu ' -0 .0 : ~ en ' ::l ::l : 0 0 , Q. .;;' ''--'' w ,(V) L- ' « a.. : __

570

~: ' -0 « : 'w .0 ,-0 ' > cu , w : o oo:~:E w,o,w en ' Q. ' L­cu ' , T-

en : :1-

577

.0 cu 00 '

'-0 ~ : ~ cu ' ::l .0 ' 0

T2 x:;::

\

W ,L()

z ::5.

584 591

Figure 4-10: Temporary prop loads corrected for temperature effects

40

30

20

10

0

-1 0 (J 0

Q) -20 5 -co ... -30 ~

E w

-40 I-

-50

-60

-70

-80

5~8 -90

10000

8000

N 6000 .§ z ~

€4000 ::::I en en e

D.. 2000

2003

Probable time b9"ast

Set4 was laid on the s ab

O+-~.-.-~r-~~-.-+~-+-.-r-r-r,-.-.-+-.-.-r-~,,-.-+~

5 0 652 764 876 988 1100 1212 13

-2000 Time in days

(a)

10000 Base slab (A4) concrete poured

50

40

30

20 u

0

10 eli ... ::::I

0 ... C1l ... Q) c..

-10 E Q) I-

-20

-30

-40

-50

50

40

8000 : Next base slab (AS) concrete poured 30 ~

N 6000 E Z ~

€ 4000 ::::I en e

D.. 2000

I

Set 1 Pressures

o +-~~~AH~'-TO~-r,-r.-r.-ro-r~~-''-TO-+-r'-rT-r~ 5 4 588 595 602

-2000 Time in days

(b)

20 u

o 10 Q) ...

::::I

o 1U ... Q) c..

-10 E Q)

I-

-20

-30

-40

-50

Figure 4-11: (a) First 2 years of base slab readings. (b) Excerpt from (a), showing period including concrete pour

136

z ~

6000

5000

4000

3000

-g- 2000 increasing o ..J

1000 )()(xx

O+o-.~-r~~,,~~~-.~nrtw~x"~~-.~~r.'-rT~~~~~~~

-1000 x xJ5xrif20

concrete pour . \ dUring pour

25 30 35 40 45

-2000 Temperature, °c

Figure 4-12: Temperature versus load for a typical base slab gauge from before concreting to after stabilization of concrete temperature

137

138

0.0003

0.00025

0.0002

0.00015

c 'cu 0.0001 ... ... en

0.00005

0

-0.00005

-0.0001

0.0003

0.00025

0.0002

0.00015

c 'cu 0.0001 ... ... en

0.00005

0

-0.00005

-0.0001

N CD CD

N CD CD

"" t--t--

CD ex:> ex:>

(c)

CD ex:> ex:>

(d)

ex:> 0') 0')

Time in days

ex:> 0') 0')

Time in days

a ..--..--..--

-gauge 3A I

-gauge 3B I

N N N ..--

-gauge4A

-gauge4B

N N N ..--

Figure 4-13: Strains measured in the base slab gauges: (a) set 1, (b) set 2, (c) set 3 and (d) set 4

"" ('i') ('i') ..--

139

0.0003

0.0002

c: 'cu 0.0001 ... -en

-0.0001

0.0003

0.0002

(j)

c: 'cu 0.0001 ... -en

0.0000

-O.OOO j

Excavation Excavation phase 1 phase 2 1/\ CI I I 14+1 _ 0 I 1++1

I :';:;1 I I CUI I I =1 I

21 I ~I I

.-1 I oJ I 0 1 I

~ : · 1 I

112 I I I Q) I

:: 1-:-1

I

o N LO

CD I'­LO

.;;t o CD

west east c.. blue e c.. black W purple

red

The zero reading is the average over 24 hours 2 weeks after concrete pour and the day before excavation

N ("I)

CD

o CD CD

co co CD

Time in days

Figure 4-14: Strains measured in PI

c.. blue west east Excavation Excavation C 0

0 ..... black M purple phase 1 C phase 2

:.;:; c.. u >-

1 /~llg ::::J ..... red l::; I I cu_ I~ 1++11 CUI 00 II ..... cu I I 1=1 C II 8.> I I 21 011 0 I I 001 U II l EE I I CI .a II----C: ~ Q) I I .~ CUll I ..... I I Cii II I

I el Q) II I I Cl.J 00 II I I ~ CU ll I I (!) I I ~I I I 0 1 I I Cl.J I EI I Q)I I f-I I I I The zero reading is the average over 24 I I I hours 2 weeks after concrete pour and I I I

I I the day before excavation I I

N 0 co CD .;;t N 0 co CD .;;t Q) N .;;t I'- 0 ("I) CD co ..- .;;t .;;t LO LO LO CD CD CD CD I'- I'-

Time in days

Figure 4-15: Strains measured in P4

140

CI.)

t:

0.0003

0.0002

'iii 0.0001 .... -en

-0.0001

0.0003

0.0002

t: 'iii 0.0001 .... -en

-0.0001

Excavation § phase 1 § I ;/" i\. II rol 1401 1_ I enl

I I I Cl I 11'-1 I II Oi I II el I I I o!

: :: b I I I COl I 1101

: :: ~ I II ~I : :: 1-1 I I I I I I I

o N L()

c Excavation 0

phase 1 g I ;/" i\. 11's1 1-..1 I_I (/)1 I I I .!: I

I % : el I Q..j

: ~ I ~I I 01 I Q..j I EI I Q) I I I I I I I I

o N L()

Excavation phase 2

I I I+-+f I I I I I I I I I I I I I I I

00 '>t L()

Excavation phase 2

I I 1+01 I I I I I I I I I I I I I I I I I I I I

a. blue west east c e o a. t5 ~ 21 1 ~ _

en l l 0 co C II a. > 01 1 I E °E () I I I .0 II-----C: ~ Q) co I I I ....

black W purple

red

Cii II I Q) I I I (/) I: :

The zero reading is the average over 24 hours 2 weeks after concrete pour and the day before excavation

'>t o CD

Time in days

(a) Gauge set A

ro ~~

II ~ E II 8. ~

.0 II I E a. ~:: :~ Q) e Q) II-r- I- a. (/) II I co I I I

OJ I I

N ('i) CD

o CD CD

00 00 CD

blue west east . black W purple

re ~ ~

The zero reading is the average over 24 hours 2 weeks after concrete pour and the day before excavation

'>t o CD

Time in days

N ('i) CD

o CD CD

00 00 CD

(b) Gauge set B

141

0.0003

0.0002

c 'IV 0.0001 ... -en

-0.0001

Excavation § Excavation phase 1 ~ phase 2

c o blue west east

1 ;/ \. I I I.. 1+01 1+->1

I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I

o N lC)

:g ro 2 ~~ en

ll ~ E

§ I I 8. ~ 011 E Q. .0 II I Q) e ~ ::~f- Q. C/) II I

black E3 purple' ,

red ~

Q) II I ~ II

zero reading is the average over 24 hours 2 weeks after concrete pour and the day before excavation

<0 I'­lC)

~ o <0

Time in days

(c) Gauge set C

N ('f)

<0

o <0 <0

00 00 <0

Figure 4-16: Strains measured in P2

142

400000

350000

300000 II) '0 I:

8 250000 Q) II)

.~ 200000 cu ... -~ 150000 E i=

100000

50000

0

0

y = 4624.83x + 19539.09 R2 = 0.99

Excavation began underneath props

7 14 21 28 35 42 49 56 63 70 77 84

Time in days (concrete pour on day 1)

Figure 4-17: Time/strain versus time for an RC prop gauge at chain age 89+205

400

350

300

II)

:::I. 250 I: cu ... 200 -II) 0 ... .~ 150 :!i

100

50

0

0

Excavation began underneath props

'. -' .. ' - ' .-'

Measured strain

.... -... .. .. -- --- _ ... . -I" -' _ . . • 0. ···· _.-I Strain predicted by : Equation 4-12 I I I Temperature

i \

7 14 21 28 35 42 49 56 63 70 77 84

Time in days (concrete pour on day 1)

150

130

110

u 90 0

Q) ... :::I

70 -cu ... Q) Q.

50 E Q) I-

30

10

Figure 4-18: Measured and calculated strain against time for an RC prop gauge at chain age 89+205

143

5000

4500

4000

3500

3000

-z 2500 ~ -"0 nJ .3 2000

1500

1000

500

0

-500

..........

.j:>.

.j:>.

Direction of construction ~Piles

ooooooo®®-oowoouoooo Props ~ ~ fj ~ " .,~ T \ ,' ,' " " " 'I~ P# - RC prop .;·i :;: . ;.;:;.;.:'.;.; . . : T# - Temp. prop ~ Z Inclinometer Temperature . ,:-: :.:".:. . :.;~?: "

X & Y Instrumented piles ,. "" " '. ; .. ': •. , ;.' . '---_ ________ -:---:--__ ----,-__ -----:-__ -,--'--_-\ .,. . ~ I I, ' ,, . , It ., ..

. ," I, I, '. II,," " ,, ',", I

I I I ,'r'· I . : I "

"~ .,, »1 I 'E I I ~ I I 8,1 I E

ClJI 2 tl 191

CIJ a. o ClJ

I ~I ClJI ~I 0..1 01 XI WI

I I I I I I I I I

..... CIJ

..... Q) CIJ co ..c 0..

I 1 I I ,... ........... .' . ' ,.; . ',' , . .••. 'I I I I :·· .... 1:··,·.:· ... I :'. I . ';',' ', :.,' ; " ':,:': '; ':'. ,.; ,' .,.; . ... I I

...... 1'" I I ' I ': '. I " "'1:' .... ' :'~' I I ... ," "{j; '1'," 1' , "0 I ':,, : I I I

-g' 1 "," ":' ~ I I I P2 AnJ1 ~ n l Q) I I 19 I I I

..... 1 I ~ I I I Q) I 1"0 .- I I I CIJ Q) ('i)

...... col l=ro f-I I Q)1..c1 I ..... ~I

~I ~I I .~ NI ..cl ~I I ..... f-0.. I W I I

~ co ..... CIJ

1

~

I

"0 Q) ..... CIJ' P4 u "01 ::J "0 a3 1

... Q) ..... CIJ >

N I c 0 0 "0 E Q)I u Q) Q)

ClJI .0 > ... ~ I co 0 ('i)

0..1 Ui E f-Q) Q)

~ CIJ

... co ..... N III f- f-

Il d llll ldllllld

l "\l U} I.U lTJ U r-- '<t ..... co ..... '<t '<t LO CD ,...... ,...... co CJ) CJ) LO LO LO LO LO LO LO LO LO LO

Time in days

Figure 4-19: Reinforced concrete prop loads uncorrected for effects of shrinkage

LO 0 CD

N ..... CD

30

25

20

15

10

5

o -5

5000

4500

4000

3500

3000

-z 2500 .lie: -'0 C'tI

.3 2000

1500

1000

500

0

-500

..... +:>. Vl

Direction of construction ~Piles

ooooooo~®ooooaoooooo

~o~~c prop ~ ~ rj ~ . ' . ,"' " ,,. • "" 1 • • ,' "

," II It, • I I:'~ .::', " .,,-. \',

T# - Temp. prop , Temperature Z Inclinometer I '

,It I " " " ". ,' 11 , I. -,, " " "", I

j I X & Y Instrumented piles I:',;, "'. ;'. '. :.;';': +J-+---+-------+---+--+-----+�---I�~-->'I '----I. I I " .. " • " I,. " " ~ " , I I ·1· I I .. .. ." .'~ •. .. ~. " ,' .' ' . • - ' ... ··1

.. .. .. , " I I ~ , \ I " I " '. II I' l ' II '" ' , ' . " .. .. ' _- r . I I I .... r.· .. ·· I ". I . . .. ' . .' , '. .- •.• • :.: I I , : I " . , " ' " ' ' . ,~ ' -. I I ' I ': '·. 1 ",( ., ,: ', :.. I

'l" I ,,' I ····· '1 ·:·\. . I .'1': I I ' - ' '' '.'. 00

I >,1 I 00 I I .. '0 I 001 '01 I I 'E I 00 I '0 I I ~ I 151 CI I I ~ I 15 I a5 I I J§ I til ~I I I ~I ti I ...... I I 00 I NI Q)I I I E I cD I Q) I I '0 .S: I Q) I 00 I I

el Q) I L.. I 00 I I ~ f'" I ~I ~I P2 I 21:;1 ...... I ~ I 00 I 0..1 Q) I 0.. I

...... 0 00 . <

Q)I ti I ~ I ~ I 1'- L...: I 01 XI Kl' ...... 1 0.. I 11,1 £1 Q)I u 0..1 00 I X

ul ~I W X I 0..1 WI ul

I X I I WI I I I I

I I I I

'0 I I .0211

I P4 .£Qg lI '0 00 L.. I I Q) I

I I

N ~ L{)

0> ~ L{)

to L{) L{)

Time in days

C") to L{)

Q)- > OO~II 0

~8 11 ...... ~ ,I,I.Ir,I, 'l,

0 I'- ~ I'- I'- co L{) L{) L{)

",/ , .' ,: ",'. ·"t· ".

...... co 0> 0> L{) L{)

-/ \ ,,',. ",0" " , '.:': '

L{) 0 to

Figure 4-20: Reinforced concrete prop loads corrected for effects of shrinkage using Equation 4-12

N ...... to

30

25

20

15

10

5

o -5

(,) -10 0

~ -15 .a

~ -20 Qj

Q.

E -25 Qj

I-

-30

-35

-40

-45

-50

-55

-60

-65

......... +:-0\

5000

4500

4000

3500

3000

~ 2500 -'0 III

.3 2000

1500

1000

500

(/)

t .s (/)

..-

>­~

..-~ ..-o ..-

0)0. ~ 0) "0

KJE cu(/) KJ ~ .c2 .c t .c .s

0) a... (/) a... .s a... (/) (/) • 0. U (/) . ' (/) 'c ro 0 0 x' , 0,-0 I ._ " , .. .c X - wi 0) x C 1"- . a... W(/)I L.. WI 0) I-

.'1 I " ". I I U , r . , , '., • , r ' . , 'I . xI

" 'j""'" ," _,' ' r" " "" " wI I ' I I . " "I' " , , . " ", ' I I I I 111

' Temperature: I I ~ Exc. directly under P1

Exc. directly under P2

° r'~ . 4: .. 4

_5004 9 >" , . , 486 493 500 507 514

Time in days

"0 0)

m -(/) .~ C'? I-~ N l-

'. ·L,:. " ,

521

,.. ... ~

I ,I . . . , ,

• J • ~, :- .:~ ~ I L.::J en l NI 0)1

KJI .c l 0...1 UI ~ I

I I ~ , ' , ,

-'I:" " ,: (/)1 "gl 0)1

NI 0)1

KJI .cl 0... 1 UI x Wi

! I

:l

30

25

20

15

10

5

o -5

o -10 0

~ -15 .a

~ -20 Q)

Co E -25 Q) I-

-30

-35

-40

-45

-50

-55

528 535 5t 2-60

-65

Figure 4-21: Close up of Figure 4-20 showing reinforced concrete prop loads up to just after Excavation Phase 2

26 -- &- - P1 calcu lated stiffness

24 -- 0 - - P2 calculated stiffness

22 cu --~ --P4 calcu lated stiffness

Q.. 20 :iE - 18 " c:: 0

UJ" 16 iii 14 tfj CI) c 12 :=

P4 P1

:0:; 10 tfj Best-fit lines CI)

8 > :0:; cu

6 Qj 0:::

4

2 2001 2002 2003 2004 2005

0 452 817 1182 1547 1912 2277

Time in days

Figure 4-22: Stiffness calculated by assuming that there is no change in prop load after construction is completed and the calculated best-fit lines to these data

4000

3500

3000

2500

z 2000 ..=.::

~ 1500 o

..J 1000

500

-2 -500

-1000

No-work stage with in Excavation Phase 1

~D"B 10

/

0 aD a o DO

12 14 16 18

Before Excavation Phase 1 T t 0c empera ure,

readings taken in month after construction was completed

20 22 24 26 28 30 32

Figure 4-23: Temperature versus load for a typical RC prop gauge during periods when no construction activities are occurring

147

0.0003

0.0002

0.0001 Period of Excavation Phase 2

1 1

: Removal of T1: 1 -I 1 1

: Removal of T2:T3 1 • I I' --r---l

1 1

Difference in strain between other pairs of gauges

1 1

.~ ~~~~~~ ... 0 II)

.5 4 5 443 471 499 5211 555 ~83 639 ~ 1 g-0.0001 Time in days : ~ ~ 0-0.0002

-0.0003

-0.0004

-0.0005

0.0003

0.0002

0.0001 c

(a)

1 1 1 1 1

: X15&X16

Difference in strain between other pairs of gauges

~ 0 r~~~~~~~_~~!liiii~~ .5 4 5 443 471 499 527 1 555 639 ~ 1 1 u-0.0001 Time in days 1 1

C : : ~ 1 1

~-0.0002 Period of : : is excavation phase : :

2 in front of w=-=a""'II-';' 1:---;1_4

-

1 1 1 1 -0.0003 : : Removal of T1 1

1 1 :;:::.-L 1 1 1 Removal of T2,II 3 1 1 1 -'-l-..-I

-0.0004 1 1

-0.0005

(b)

'\ Difference in strain between gauges Y21 & Y22

Figure 4-24: Difference in strain measured for the pairs of strain gauges in (a) pile X and (b) pile Y over the period of construction

148

0.0003

0.0002

0.0001

Odd numbered gauges: back of pile (retained side)

0+-~~~~~~~~~~-+~-.~-+~~-r~~.-.-41~-'~'-+-'-'-~ 4 5 443 471 : 555 583:

-0.0001

-0.0002

-0.0003

-0.0004

-0.0005

0.0003

0.0002

0.0001

Even numbered gauges: front of pile (excavated side)

1 1 1 1 1 1 1 1 1 1

Period of excavation : in front of wall -:

Odd numbered gauges: back of pile (retained side)

(a)

1 1 1 1

: Removal of T1: : 1 • 1 1 : Removal of T2,T3 : I I _I

1 1 1 1 1 1

1 1 1 1

1

611 639

Time in days

O~~--~~~~~~~~~+T~~~~~~~~~~~

4 5

-0.0001

-0.0002

-0.0003

-0.0004

-0.0005

443

Even numbered gauges: front of pile (excavated side)

1 1 1 1 1 1 1 1 1 1 1

Period of excavation ~ front of wall :

(b)

1 1 1 1

1 1 1 1 1 1

Y22

1 1 1 1

1 1 1

Removal of T1 : -I

Removal of T2, T3 1 I .1 - 1 1 1 1 1 1 1 1 1 1 1 1 1 1

611 639

Time in days

Figure 4-25: Strain measurements over period of construction for (a) gauges X1S, X16, Xl7 and Xl8 and (b) gauges Y21 and Y22

149

44

42

40

38

36

c 0 34 c( E (j)

32 > CI)

...J

30

28

26

24

-5 o

DAY:

Wall toe

ro > o E ~ c.. e a.. ~ ~ o c.. E ~ ..... Q)

~

Exc. Phase 1 ...... ievel

Exc. Phase 2 ...... Iewel

Cracking occurred over the periods in which these movements took place

5 10 15 20 25 30 35 40 Deflection, mm

Figure 4-26: Deflections measured in pile Z with the inclinometer around the time of Excavation Phase 2 and Temporary Prop Removal. Readings taken in 2001.

150

......... VI .........

.. Instrumented piles

'---- ...

Figure 4-27: Photos showing bulge in concrete between the levels of the temporary prop and base slab. The concrete above the bulge was poured within the casing and is therefore very smooth .

1&2 20 3&4 20

18 18

16 16

.P 14 .P 14

~- 12 i 12 .a .a

10 l! 10 l! '" '" 8 0. 8 0. E E 6 .. 6 '" l-

I-4 4

2 2

-0 .0002 -0.0001 0 0.0001 0.0002 -0.0002 -0.0001 0 0.0001 0.0002 Strain (s) Strain (6)

5&6 20 7&8 20

18 18

16 ...l6 .P 14 .P 14

i 12 i 12 .a .a l! 10 l! 10

'" '" 0. 0. 8 E 8 E

'" '" 6 I- 6 I-

4 4

2 2

-0.0002 -0.0001 0 0.0001 0.0002 -0.0002 -0.0001 0 0.0001 0.0002 Strain (s) Strain (6)

9&1 0 20 11&12 20

18 18

16 16

.P - .P 14 14 i 12 i 12 :::J .a E 10 E 10 '" .. 0. 8 0. 8 E E ..

6 .. 6 l- I-

4 4

2 2

-0.0002 -0.0001 0 0.0001 0.0002 -0 .0002 -0.0001 0 0.0001 0.0002 Strain (s) Strain (6)

13&14 20 15&16 20

18 18

16 16 u { 14 .P 14 0 , i i .a 12 .a 12

E 10 I! 10 .. .. 0. 8

0. 8 E E .. ..

I- 6 I- 6

4 4

2 2

-0.0002 -0.0001 0 0.0001 0.0002 -0 .0002 -0.0001 0 0.0001 0.0002 Strain (s) Strain (e)

17&18 20 19&20 20

18 18

16 16

.P 14 u 14 0

i

I 12 i

I 12 .a .a

E 10 E 10

'" .. 0. 8 0. 8 E E

'" 6 .. 6 l- I-

4 4

2 2

I I

-0 .0002 -0 .0001 0 0.0001 0.0002 -0.0002 -0.0001 0 0.0001 0.0002 Strain (s) Stra in (e)

152

21&22 20 23&24 20

18 18

16 16

14

I 12

10 '%-8

6 " - 2

10

-- 8

6

4 4

2 2

-0.0002 -0 .0001 0 0.0001 0.0002 -0 .0002 -0.0001 0 0.0001 0.0002 Strain (s) Strain (s)

25&26 20

18

16

14-

j

6

4

2

-0.0002 -0 .0001 0 0.0001 0.0002 Strain (s)

Figure 4-28: Strains measured in Pile Y over the 30 days after excavation and before Base Slab Construction

153

VI +:>-

Top 9fwfill _44T Strain gauge readings I I o Pile Y b. Pile X I I I I I I

25&26! 4of\ I I I I

23&24: I I

111 21&22i 38 + t

19&20: 36

-I

17&181

34 15&16:

I _I

13&14: 32 I I

11&12: ~ Bending moment

9&10 I 11' : calculated from 5th

: : and 6th order fits I

7&8 fl I

c I

0 I 101

5&6 « 0) I "-1 ..,tf'1- E I I I I

'<t I '<t I CI CI a; _I I

2l

.QI.QI

3&4 I I > .... I +-' I I I C1I ro I ro I

-I I ...J ::J I ::J I 1&2 I I 0"1 0"1

I I WI WI

:Wall toe °L.I "L.I 01 UI ~I ~I

-1000 -500 0 500 1000 1500 2000

Top Qfw4i1l 44 ... I I

'l"" I I ,I I 1"1-, , I I "',42 I I

-I I ""-25&26 '

23&24 40

21&22 38

19&20 36

17&18

-I 34 15&16

13&14 32

11&12

9&10

7&8

5&6 I /

I " /

3&~: I "" 1/ I ) -I

1&2 I -I I I I

'¥Val! toe

-1000 -500

,

0

Strain gauge readings I

: 0 Pile Y b. Pile X I I I I I I

~ : Bending moment :~ calculated from 6th

c o « E

~ C1I

...J

: order fit I I I I I I I I

14,!\\ :

0+ I I

+1

+:T ° +1

f tl1i ~

I I

I ~I

Cracking has , occurred

~

.. I I / I I Bending moment

/ I I )~ calculated from 5th

/ I I order fit I

.0 0)1"­

I I I '<t.'<t

I °L.I "L.I

01 0, ~I ~I

500 1000 1500 2000 Bending moment, kNm Bending moment, kNm

Figure 4-29: Bending moments calculated from inclinometer and strain gauge measurements before and after Excavation Phase 2

(a) Before Excavation Phase 2: Day 530 (20/03/01) (b) After Excavation Phase 2: Day 540 (30/03/01)

Top qfwCflIl 44 1 Strain gauge readings 1 Top ofw9" 44 1

Strain gauge readings 1 , , 1 1 1 1 o Pile Y ;:,. Pile X 1 1 1 o Pile Y ;:,. Pile X "'f 1 1 1 .- ....... 1 1 1

, 1

~ 1-_ 1 1

'""I 1 -.... -- 1

1 "-t..., 42 1 1 1 1 -""4Z 1

D. I: Bending moment 1 1 1 Bending moment 'f., -I 1 1

25&26 ,

: 1 calculated from 6th 25&26 : 1 ~ 1 calculated from 5th 1 1 1

1 order fit 1 1 o "'J:/ order fit 40 1 _I 1

23&24 1 23&24 : 1 4 to, 1 , , I 'i<,1

21&22 38 + o \J -I

~ "" 21&22 ! 38 1 1 , : + Cracking has

I c 0 C IZ1 : occurred 19&20 1 0 19&20 o 1

1 36 « 36 « 1 1 E E o : ,"'T 1 D. D.

17&18 1 Qj 17&18 Qj

I, 1 1+ 1 > > If} 1 34 Q) D. 34 Q) D.

15&16 1 -I 15&16 -I 1 v.(. : 1 1 Q'" ...: 13&14 1 13&14 1 32 32 1 1 1 1 1 1 1

11&12 1 D.I

11&12 1

1 1 1 1

30 1 Bending moment 1 1 1 1 1 1 calculated from 5th 1

9&10 1 1 9&10 1 1 1 order fit 1 1 1 1 1 28 1 1

7&8- 1 1 7&8- 1 1 1 1 Bending moment 1 1 1 1 01 ~: calculated from 6th

5&6-: 1 CJ) ..- 1 5&6

CJ) 1 I , 1 I ...t! order fit 1 '>t '>t, '>t 1 1

3&4 1 1 .f. 3&4 1 1 I _I 1 /24 1&2-1&2 1 1

1 1 1

I W911toe • ~I "L.-I }/Vall toe 1 'f;: 'f;: 0) 01

:2: 1 :2: 1 I , ,:, ,: I , ,2~ I I , ~I: ,:2:, :

-1000 -500 0 500 1000 1500 2000 -1000 -500 0 500 1000 1500 2000 Bending moment, kNm Bending moment, kNm

- Figure 4-30: Bending moments calculated from inclinometer and strain gauge measurements before and after Temporary Prop Removal VI VI (a) Before Temporary Prop Removal: Day 551 (10/04/01) (b) After Temporary Prop Removal: Day 617 (15/06/01)

2400 40

2200 Temperature 30

2000 20

1800 10

1600 N' '<t ' , 0 I

~ ~:-o -0 1 () Q)I Q) I 1400 (/)1 .0 I Q) --. 1 -1 0° ~

rol -0 ro I> :J I Z 00 01 0 0 1 .J:: I C ~ ':'::~1200 0..1 Q) Q) ol E c.. -20 ,a

·1 (/) 0-1 Q) ..-.I 't:l lOl C'CI U I ro I 1--' ~1000 Jj l I 1"- ~ I -30 ~ I II- '-1 Co ...J I .0 1

-40 ~ 800 I ro l 00 1 Q)I I-

600 (/)1 -50 ro l

400 .0 1

-60 I Xl

200 Q) I

-70 z: I

0 -80

-2005 535 542 549 556 563 570 577 584 591 5 8-90

Time in days

(a) Tl

2400 40

2200 Temperature 30

2000 20

1800 10

1600 0 I ()

1400 1ij : -1 0° z 00 : ~ ':'::J200 N I -20,a 't:l Q) I ... C'CI

C'CI (/) 1 01 000 ro I -30 Q) ...J .J:: I Co

0.. 1 -0 E 800 . : Q) -40 Q)

~ I ~ :; I-600 L!J II ro I 0 -50 .o l e.

I (/) I ~

400 II :J I~ -60 O l ~ '>:;; 200 ~ I ro -70 0.. 100

0+rt~~+n~~~+n~~~~~~+n~~~~~~~Tn+M~~-80

-2005 535 542 549 556 563 570 577 5 8-90

Time in days

(b) T2

156

2400

2200

2000

1800

Temperature

40

30

20

10

a 1600 1 ~ ~ ;~ ~; _I '" 10) 1 0» 00 1 U

1400 CUi ~ I:J I 0 0) I ~ -1 0° _ z 00: 0) :8.: E ~: :J ~ ~J 200 Nt ~ I I ~ ~: 8. -20 .a - 0)1 ~ ~II~ _ ... 001 IoU ....... X I --- ~ evo 1 000 CUi I I- 0) Il() -30 Q)

...J £1 I Zl~ C-800 0..: ,..-y-vvflVlt1I"""'""",,..r.nNll

l -40 !ii 01 I-XI

600 w: -50 I

400 : -60 I

200 -70

0+rtMrt~+n.rtTh~+n~Th~.rt~~+n~oH~.rtTh+n+n~+h~~-80

-2005 1 535 542 549 556 563 570 577 584 591 5 8-90

Time in days

(c) T3

Figure 4-31: Individual temporary prop loads

157

V1 00

1800

1600

1400

1200

E ~ 1000 ..;-c:: CD E

800 0 E C) c::

" 600 c:: CD m

400

200

0

4~6 -200

RC prop

Pile Y - Gauges below level of formation/base

slab Temp prop

Base slab

c 0

1:5 ::J .... -(/) c· -g I I -g 01 tl cil °1 1:5 I I 1:5 -EI .81 EI 2 II 2 (/)1 (/)1

ffil - - ............ (/) I I (/) 0) I 0)1

IlJI § II § 001 001 0>1 c..> I I c..> ~I ~I ·~I (\I I I -.::t 0...1 0...1 8::1 0... I I 0..._ c.31 c.31 col ...... I I ('I) XI XI °1 0... I I 0... wI wI

444 472

• 13 & 14 • 11 & 12 • 9 & 10 • 7 & 8 "0

c 0) • 5 & 6 (\I 0) (/)

co • 3 &4 l! 1 & 2

L: 0...

(/)

t. ti -ffi I "0 I I c I

0) I c I I .Q I -EI .... IO)llcrol

(/)

............ 0= (\II 0)10) II:;::::;-E I 0)1 (/)1 00 II~ 00 I 001 ~1~II]iEI ~I 0...10... II ~~ I 0...1 c.31 c.3 11- _I c.31 XI X 11"""(\11 Jjl WIWIII-I- I I

500 556 . 3&4 1&2

Time in days

Figure 4-32: Pile Y bending moments below the base slab plotted against time

c:: 0 15 ::J .... -(/) c 0 ° ro .0 > co 0

CI) E 0) 0::

co > 0 E 0) 0:: ('I)

l-N· 1-1

I

~

5&6

~3&14

9&10

7&8

612

Vl \0

1800

1600

1400

1200

E ~ 1000 ...;' s:: III E

800 0 E 0'1 s:: '0 600 s:: III m

400

200

0

4~6 -200

RC prop

• 25 & 26 Pile Y - Gauges above level of formation/base

slab Temp prop • 23 & 24

__________ • 21 & 22

C o U 2

+oJ I/) C o o E co ~I 0>1 .~ I §:I COl 0 1

I

C C o 0

U liu 2 112 ti I I ti § II § 0110 N I I '<t a.. I I a.. ~ I I C'? a.. I I a..

I I

-----------. 19 & 20 Base slab • 17 & 18 ___ --l. 15 & 16

t . ci.. ~1.81 1/)11/)1 ..... 1 ..... 1

~I~I ~I~I 0..10..1 cjlcjl XIXI wlw i

I I

~ co· til ~I

..... 1 0)1 Kli .!:I 0..1 cjl Lijl

I

c l 01

c ]11 o eo I

:;:; ti I -oIIJgCI C II co-0) I I ti ("') I

..... IIE ~I ~ II ..... ~I ~ III- I 0..11~

1-

444 472 500 528

Time in days

"C C 0) N 0) I/) co .!: a.. u x

W

556

Figure 4-33: Pile Y bending moments above the base slab plotted against time

I I I I

C I I 0 I I t5 I I ::s I I "-

+oJ

I -I I/) C I ~I 0 0 1 0 leo

§I .0 I> co 1 0 0:::1 en I§ ("')1 0) 10::: 1-1 I/)

co I ..... NI co .1- 1-1

584 612

0\ o

1800

1600

1400

1200

E ~ 1000 ..;-c: (II

E 800 0

E C) c: :c 600 c: (II

m

400

200

0

416

-200

Pile X - Gauges below level of formation/base

slab

C 0 t5 :J .... -en c· 01 01 EI rol ~I 0)1 .~I

8:1 rol 01

444

RC prop

Temp prop

Base slab

• 13 & 14

• 11 & 12

• 9 & 10

• 7&8

• 5&6

I: 3&4 1&2

en· c l I c I : I 1:1 I I CI .QII.Q 1: I ci.l 21 "0 I I 21 ul I u 21 SI ~I CII c ~ I 2112 en I en I .... 1 ~1I2 ~ I enl I en "'-1 "'-1 "'-1 <DIIJg~1 511 5 511 511 511 enl12 -I 0110 ro I ro I ~I ~II ~ ~ I NII~ it I it I a.. I a..11- _I a.. II a.. ·1 ·1 0 1 011 ~ ~ I ~I I C')- ~I ~I ~I ~II I a.. a.. ill ill ~--.~, 0 x

ill

~,~ I:?E:":;:':~ IUIIII ~ : A XVi!

472 500 till' .. • 556

Time in days

Figure 4-34: Pile X bending moments below the base slab plotted against time

::J .t. !:;I ~ II 011 o 11m .gll~ en II E <DII<D en I 10::

~ II~ I I

584

13&14

11&12

612

1800

1600

1400

1200

E ~ 1000 ..,.;-c: Q)

E 0 E

800

en c: "C 600 c: Q)

OJ

400

200

0

416

-200

........ 0\ ........

RC prop

Pile X - Gauges above ~125 & 26 I I I

level of formation/base 23 & 24 slab Temp prop ~ 21 & 22

19 & 20 Base slab 1+1 17 & 18

15&16 0:/: ;\; f~ .. ~ ~ I :If.. * .. .x 15& 16

c:: o

I I 1 Jj co I~ .j "

~ c:: 1 u 01 W'I _

2 ~ C::II ~ ~ c:: c:: 00 §JgI 811

ro 0 o .Q . . .Q t . a.. ~ . '0 . . ~ E:! I .0 I I > ~ o 1:5111:5 E:!1.81 00 1 a311~ ~I ClJII ~ 0:: E 2112 001001 0,)1 liro- I WIIO,) C")

CIJ 001100 T"'IT"'I ';:I~II~~I 0,)110:: I-

17&18

~ I § I I § ~ I ~ I 0,) I 00 I I - _I ~ I I T'"

0110 ClJIClJI 001 ~IIT"'NI OJIII-e> I £ £ CIJ ~ I- I- I 21 &22 £1 NII.q- a.. I a.. I £1a..11 II ~ 0. I a.. I I a.. . I . I a.. I " I I ,-

g- I ~ I I C")- ~ I ~ I d I I I II. . .A 23&24

o : IL:: IL W : W :,,,,,,,,,,~: t'V"~'1: "_..ad'!. ''--25&26

444 472 500 528 556 612

Time in days

Figure 4-35: Pile X bending moments above the base slab plotted against time

5 HORIZONTAL SOIL STRESS MEASUREMENT

5-1 Introduction

In this chapter the instrumentation used to study changes in the total horizontal stress and

pore water pressures around the retaining wall is described (push-in spade-shaped pressure

cells (spade cells)). A study carried out to determine the over-reading given by a spade cell

installed in overconsolidated clay is described in Section 5-4. This chapter also includes a

study of the stress changes measured in the soil around the retaining wall during

installation (Section 5-5) and data is presented which shows the changes in horizontal soil

stress measured due to excavation and construction of the cutting (Section 5-6).

5-2 Spade cell description

The spade cells used in this study were manufactured by Soil Instruments Ltd and

consisted of a pointed rectangular spade-shaped flat jack, approximately 7 mm thick and

100 mm wide, formed from two sheets of steel welded around the edge (Figure 5-1). The

162

narrow gap between the plates is filled with oil, and is connected to a vibrating-wire

pressure transducer by a short length of steel tubing forming a closed hydraulic system.

The spade cell body also houses an integral piezometer, in the form of a porous filter disc

connected to a second vibrating-wire transducer. The use of vibrating-wire transducers

allowed effectively continuous data to be obtained, as discussed in Section 4-6.

Tedd & Charles (1983) and Ryley & Carder (1995) used spade cells containing pneumatic

rather than vibrating-wire transducers. The vibrating-wire spade cell used in this study was

found to be unaffected by temperatures in the range expected to be exhibited in the ground

on site (9-24 degrees Centigrade) when unconfined. Readings from vibrating-wire

embedded strain gauges (incorporating temperature sensors) measuring strain in buried

concrete piles showed that the readings only became sensitive to ambient temperature

changes when the ground level was within 2·5 m of the cell (see Section 4-3-3).

5-3 Calibration

The spade cell used in the study to determine the spade cell over-reading was calibrated by

Soil Instruments. Figure 5-2 shows the equipment used for spade cell calibration. For total

stress measurement the spade cell is lowered into a compression chamber and readings

from the pressure transducer are taken as the pressure in the chamber is increased. During

this process plates are clamped over the ceramic disk (for pore water pressure

measurement) to protect it, as it has a smaller measurement range than the total stress

transducer.) To calibrate the pore water pressure transducer the clamp remains over the

ceramic disk and water pressure is applied to the transducer at its outlet point on the

transducer at the top of the spade cell. Again, transducer readings are taken as the pressure

acting on the transducer is changed.

5-4 Evaluation of spade cell over-reading

5-4-1 Introduction

Due to the considerable uncertainty regarding the magnitude of the over-reading and the

factors that may affect it (described in Section 2-4), a direct determination of the over­

reading seen by spade cells installed in overconsolidated clays should be made wherever

163

possible until more data are available. The spade cell over-reading associated with the

geology at Ashford was evaluated by means of a field study in which a cell was installed

horizontally from the cut-and-cover excavation, aligned to measure the vertical stress, and

its readings were compared with the overburden acting on it. The soil above the spade ce II

was excavated in stages over a period of several weeks, and at each stage the overburden

was calculated and compared with the reading in the spade cell. Additionally,

measurements of in situ horizontal earth pressure taken with the spade cells in the

instrumented section have been compared with pressuremeter readings. In line with other

recent research, the values of over-reading have been expressed in terms of the undrained

shear strength of the soil.

5-4-2 Test procedure

The borehole into which the spade cell was installed was drilled horizontally and

perpendicular to the wall using a continuous flight auger (see Figure 5-3), to a point

approximately half a metre short of the centre of the nadir sump (see Figure 5-4). Owing

to the stiffness and undrained shear strength of the clay it was not necessary to line the

borehole. A jacking platform was then erected, tension bars installed into the piles on

either side of the installation point, and a reaction beam connected to the bars to allow the

spade cell to be pushed into position using a hydraulic jacking system. The installation

barrel was connected to the spade cell which was then moved to the end of the borehole,

so that zero readings could be taken. Considerable care was taken to ensure that the

borehole was horizontal and that the active faces of the spade cell were horizontal aligned.

To complete the installation, the borehole was backfilled and sealed using a cement grout.

The installation barrel was left in place to protect the cables. The exact location and

orientation of the spade cell were confirmed when it was uncovered during excavation

(Figure 5-5).

Prior to installation of the spade cell the instrument was kept submerged in water to

prevent the porous stone from de-saturating and so that it was close to ground temperature

before installation. As an added precaution, a period of fifteen minutes was allowed to

elapse between the transfer of the spade cell to the borehole and installation into the soil to

allow the spade cell temperature to equilibrate fully with the surrounding soil; readings

taken proved this to be the case.

164

The spade cell was pushed approximately 50 mm at each jack stroke during installation

and at each stroke a reading of total vertical stress was obtained using a hand-held

vibrating-wire readout meter. On completion, the transducers were connected to a

datalogger and readings were taken every 15 minutes. Figure 5-6 shows the spade cell

output during the installation phase and during excavation within the sump: the total stress

readings gradually stabilized over a period of approximately 2 weeks prior to excavation.

During excavation careful surveying of the level of the material within the sump at the end

of each day has enabled the overburden to be calculated. However, as excavation

proceeded after the ring beam had been installed access to the base of the excavation was

restricted due to the amount of machinery within the excavation. The excavation phases

are listed in Table 5-1 and the volume of material removed at each excavation phase is

shown schematically in Figure 5-7.

Stage Description Date Day

1 Excavation of3·8 m leaving steps cut in clay for Tues 26/06/01 628 access

2 Steps removed. Mini-digger lowered into Wed 27/06/01 629 excavation

3 Excavate another 2·65 m generally although Thurs 28/06/01 630 only 1 ·4 m directly over spade cell

4 Further excavation less than 1 m Fri 29/06/01 631

5 Excavation and levelling to ring beam formation Mon 02/07/01 634 level

6 Ring beam and wall installation and time for July 2001 633-concrete setting 663

7 Excavation restarted. 1·7 m of material below Wed 01/08/01 644 the ring beam removed.

8 Uncovered spade cell Mon 16:20 06/08/01 669

Table 5-1: Details of nadir sump excavation

5-4-3 Results and Discussion

Figure 5-6 shows that reductions in total stress were measured at every excavation stage

and an increase in total stress resulted from installation of the ring beam. At the end of the

test when there was no load on the spade cell the total stress readings returned to zero. The

pore water pressure was approximately zero throughout the test period.

165

Figure 5-8 shows the measured total vertical stress just prior to excavation and at the end

of each excavation stage plotted against the total vertical stress calculated from the depth

of overburden acting on the spade cell (calculated using a bulk density of 21 kN/m -': see

Chapter 3). Any discrepancy between the initial steady spade cell reading and the in silll

overburden stress must be due to the effect of inserting the cell. Figure 5-8 shows that the

initial overburden is 180 kPa, giving an over-read of 49 kPa or 0·35 x cu. As the

overburden is removed, further discrepancies may arise as a result of the differential

cell/soil stiffness (cell action factor effects), and/or the development of shear stresses

between the soil and the walls of the nadir sump, acting so as to resist heave.

In addition to an over-read of the in situ stress due to the effects of insertion, it is well

known that a pressure cell will generally either over-read or under-read changes in stress

depending on whether the cell is stiffer or less stiff than the soil within which it is

embedded. This was investigated by Peattie and Sparrow (1954) in terms of the relative

cell to soil stiffness. They provide guidance on the design of pressure cells to measure soil

pressures irrespective of the type or condition of the soil. Determination of the relative cell

to soil stiffness is complicated by the fact that, even for a given stress history, stress state

and load path, the stiffness of the soil may vary by several orders of magnitude as strain

occurs.

The data shown in Figure 5-8 suggest an under-read of the reduction in vertical stress,

which is consistent with a cell that is over-compliant (i.e. insufficiently stiff: Clayton &

Bica, 1993) and/or the mobilization of shear stress on the nadir sump walls. Measurement

of the change in thickness of the spade cell blade (by means of high resolution miniature

linear variable displacement transducers (L VDTs)) during calibration in a fluid filled cell

(Figure 5-9) indicated an effective cell stiffness (E) of 4·55 GPa. This is far in excess of

the measured small strain soil stiffness of the lower Atherfield Clay of270 MPa (Xu,

2002); thus over-compliance is not an issue, and the under-read of the reduction in vertical

stress must have been due to the mobilization of shear stresses on the nadir sump walls. To

confirm this, a plane strain undrained finite element analysis was carried out using the

continuum finite element program CRISP (Britto and Gunn, 1987), across the narrow

section of the nadir sump.

The analysis started with the walls already in place and a pre-excavation earth pressure

coefficient of 1. Excavation was modelled by the gradual removal of elements to a depth

of 6·7 m, corresponding to the level at which the intermediate concrete ring beam support

was constructed. The mesh used in the analysis is shown in Figure 5-10 and comprised

166

1904 non-consolidating linear strain triangular elements. The base boundary was fixed in

both the horizontal and vertical directions. The vertical mesh boundaries were fixed in the

horizontal direction only. An elastic-perfectly plastic soil model with a Tresca (undrained

shear strength) failure criterion in terms oftotal stresses, 'tmax = Cu was used for each soil

layer and the wall was modelled as a linear elastic material. An undrained soil stiffness

(Eu) profile was derived from the measured undrained shear strength profile (Figure 3-15

and Equation 3-1) using Eu =1500 cu. This is based on laboratory test data presented by

Jardine et al. (1984) for a variety of clays including London Clay, at a characteristic shear

strain of between 0·01 % and 0·1 % (Jardine et al., 1986; Mair, 1993).

An effective soil stiffness (E') profile was derived from the measured undrained stiffness

(Eu) profile using Equation 5-1, where Poisson's ratio (v') = 0·35.

E'= l+v' E • u·

1.5 Equation 5-1

The effective soil stiffness profile given by Equation 5-2 was used to derive the effective

bulk modulus of the soil (K').

E' K'=----

(3(1- 2v')) Equation 5-2

Undrained conditions were imposed by specifying the bulk modulus of water (Kw) as 100

times the effective bulk modulus ofthe soil (K').

Figure 5-11 shows the calculated total vertical stress as a function of overburden at an

integration point located at the same depth as the spade cell (approximately 8·5 m below

original ground level). The aspect ratio of the sump box (its breath, b, to length, I, ratio =

0·72) is high, suggesting that further arching effects would be expected across the longer

dimension of the box due to the shear stresses on the shorter walls. In very simple terms,

multiplying the calculated over-read by 1·72 gives an upper estimate ofthe possible error

due to arching between the sump walls.

The raw data from the finite element analysis were adjusted for comparison with the spade

cell data by:

• Multiplying the initially calculated over-read by 1·72 to allow for shear stresses on

the short sides of the sump as well, and

167

• Adding the over-read due to spade cell installation of 49 kPa.

Figure 5-11 shows that the adjusted finite element analysis results match closely the

measured data. This tends to confirm that the discrepancy between the calculated and

measured overburden may be accounted for by a combination of (a) spade cell over-read

of 0·35 Cu due to installation effects, and (b) the mobilization of shear stresses at the walls

of the nadir sump.

A simple equilibrium analysis of the mobilized soil/wall adhesion further supports this

conclusion. For an excavation of dimensions 1 x b in a soil of average undrained shear

strength Cu and unit weight y, a mobilized soil/wall adhesion of a clI

and a depth of soil

remaining (above the spade cell location) of h, the downward force acting on the soil in

addition to the weight is:

Equation 5-3

This is equivalent to an average increase in vertical stress of:

Equation 5-4

The undrained shear strength (cu) varies with height above the spade cell location (h) at a

rate of ~ kPa, and as previously stated in Section 3-6-3, the undrained shear strength at the

position of the spade cell Cu = 140 kPa. Thus the average undrained shear strength Gil in

the soil mass remaining above the spade cell is:

- f3h Cu = 140--.

2 Equation 5-5

Substituting equation 5-5 into equation 5-4 provides an indication of the error (over-read)

Err due to the mobilized shear stress on the sump wall perimeter as a function ofthe height

of the remaining soil above the spade cell:

E = (280ah _ aRh2 )(1 + b) . rr f/' lb Equation 5-6

Figure 5-12 shows the measured spade cell over-read and the calculated error (Err) for

adhesion factors (a) between 0 and 0-4. It can be seen that the measured discrepancy could

168

be accounted for by the mobilization of approximately 30% of the undrained shear

strength at the soil/wall interface as the overburden is removed.

Figure 5-13 shows the in situ total horizontal stresses measured at the instrumented

section, both raw and corrected for installation effects by subtracting 0'35 cu, and self ..

boring pressuremeter data obtained 250 m northwest of the nadir sump. The subtraction of

0·35 Cu to account for spade cell installation effects brings the spade cell data largely into

line with the self-boring pressuremeter test results. Consequently all total horizontal stress

measurements have been corrected by 0·35 Cu (unless otherwise stated).

5-5 Installation effects of a bored pile retaining wall in overconsolidated clay

5-5-1 Introduction

This section contains an analysis of the effects of wall installation on the soil surrounding

the contiguous bored pile retaining wall.

5-5-2 Calibration and installation of spade cells

The spade cells in the instrumented section were calibrated and installed by the Transport

Research Laboratory. Figure 5-14 shows the installation of one of these spade cells.

5-5-3 Stabilization of spade cells following insertion

Figure 5-15 shows the readings of total horizontal stress and pore water pressure during

spade cell insertion and for a period of one month afterwards (readings from spade cell or

piezometer number n are referred to as Sn and Pn respectively). The excess pore water

pressures induced during spade cell insertion had dissipated and readings of total

horizontal stress had stabilized prior to the installation of the bored pile wall.

5-5-4 In situ total horizontal stresses and pore water pressures

Figure 5 -16 shows the corrected, stabilized readings of total horizontal stress and pore

water pressure from all spade cells and piezometers, plotted against depth below original

ground level. The initial in situ pore water pressures were slightly less than hydrostatic

169

beneath a water table 1·2 m below the original ground surface in the upper Atherfield

Clay. In the lower part of the stratum the deviation of the pore water pressure from

hydrostatic was more pronounced. This distribution of pore water pressure with depth is

consistent with underdrainage of the Atherfield Clay into the underlying Weald Clay,

within which an ejector well pore pressure control system was in operation about 100m

from the instrumented section (see Section 3-3-5).

The variation of in situ total horizontal stress aho with depth z (below ground level) was

approximated by the linear function given in Equation 5-7.

aho (kPa) = 0·8 + 20·6 x z (z in m). Equation 5-7

5-5-5 Effective stress profile

The profile of effective horizontal stress cr'ho with depth (obtained by subtracting the

measured pore pressure from the total horizontal stress at each location) is shown in

Figure 5-17. The implied in situ earth pressure coefficient Ko (=a'ho/a'yo) is generally

within the range 0·9 to 1· 5 with an average of 1·1. This is rather less than would be

expected on the basis of the one-dimensional stress history of the deposit (as discussed in

Section 3-3-4).

5-5-6 Pore water pressure changes during and after wall installation

Figure 5-18 shows the variations in pore water pressure measured during the period of

wall installation, together with the periods of installation of each pile group. Detailed

interpretation ofthese data is complicated by the sequential nature of pile installation,

together with the fact that the spade cells could not all be installed at one cross section but

had to be spread along the length of the wall. However, the general response of each

transducer comprises a reduction in pore water pressure on excavation followed by an

overcompensating increase in pore water pressure on concreting. The pore water pressure

then gradually falls to the original value over a relatively short period of time (in this case

less than a day) as the concrete begins to set. This response is similar to that measured in

the field by Symons & Carder (1993) and in centrifuge model tests by Powrie & Kantartzi,

(1996). The magnitudes of the measured changes vary, but generally reduce with distance

from the pile being installed. The changes in pore pressure were consistently most

pronounced at piezometer P4. The reason for this is unknown.

170

Pile installation may be idealized as an expanding or contracting circular cylindrical

cavity. Analysis following Gibson & Anderson (1961; reproduced in Powrie, 2004) shows

that if it remains elastic, the soil surrounding the cavity deforms at constant volume and

constant average total stress p (= [O'r + O's + O'v]!3). In an elastic medium, deformation at

constant volume occurs at constant average effective stress pI; hence while p remains

constant there should be no change in pore water pressure. However, a reduction or

increase in cavity pressure (from the in situ condition) equal to the undrained shear

strength of the soil, cu, is sufficient to cause yield and hence potential changes in pore

water pressure. In the present case, the in situ total horizontal stress (given by Equation 5-

7, O'ho = 0·8 + 20·6 z) is greater than the undrained shear strength (given by Equation 3-1,

Cu = 22 + 7 z) for depths z greater than 1·6 m. Unloading the pile bore from an in situ total

horizontal stress O'ho to a cavity support pressure P (where [O'ho - P] ~ cu) will cause yield

of the surrounding soil within a plastic radius rp given by Equation 5-8, where R is the

cavity radius.

[a -c -P] r = R x exp ho u •

p 2c u

Equation 5-8

In the present case, taking the cavity support pressure P = 0, the cavity radius R = 0·525 m

and the profiles of in situ total horizontal stress and undrained shear strength with depth

given by Equations 5-7 and 3-1 respectively, the width of the plastic zone (r p - R) varies

from zero at a depth of 1·6 m to 0·6 m at the bottom of the pile bore.

Even the closest pore water pressure transducers were outside the theoretical plastic zone,

and might therefore have been expected to show no change in pore water pressure.

However, it is well known that soil does not behave as an ideal elastic material. Also, the

neglect in this simple analysis of both the potential for vertical load transfer to the soil

below the pile toe and the effect of any piles already installed introduces further degrees of

approximation. Nonetheless, the measured pore water changes are generally not large, and

decrease rapidly with distance from the wall as the effects of the approximations become

less significant.

For a wall constructed from diaphragm wall panels rather than individual piles, a greater

pore water pressure response would be expected. Excavation in plane strain results in a

reduction in average total stress p of half the reduction in horizontal pressure within the

trench, in the soil close to the trench (Powrie & Kantartzi, 1996). If the soil is again

171

assumed to deform elastically with p' = constant in undrained conditions, the change in

pore water pressure would be expected to be the same as the change in p, i.e. half the

reduction in horizontal stress within the trench. Powrie & Kantartzi (1996) report

centrifuge tests modelling diaphragm wall panel installation in which the measured

changes in pore water pressure clearly reduced with reducing panel length. However. the

difference in pore water pressure response between bored pile and diaphragm walls is not

so clear from the field observations described, but not presented in any great detail, by

Symons & Carder (1993).

Figure 5-19 shows the measured pore water pressures plotted against time for periods

before, during and 10 months after wall installation. When viewed in this way, the

fluctuations in pore water pressure due to pile excavation and concreting are little more

than noise, and the data show clearly that the overall net effect of wall installation on the

in situ pore water pressures was negligible. This is consistent with the observations of

Symons & Carder (1993) and Powrie & Kantartzi (1996), and confirms that, in an

overconsolidated clay, the overall effect of bored pile or diaphragm wall installation on the

in situ pore water pressures may reasonably be neglected. In the 10 months following wall

installation the pore water pressure slowly reduced at the positions of many of the spade

cells. The pore water pressure began to fall in March (approximately Day 150), probably

due to seasonal variations and/or construction dewatering being carried out about 500 m

from the instrumented section.

5-5-7 Total horizontal stress changes due to wall installation

Figure 5-20 shows the total horizontal stresses measured during installation of the wall. As

with the pore water pressures, interpretation is complicated by the sequential installation

ofthe piles and the staggered pattern of the spade cells along the length of the wall.

Nonetheless, the influence of pile construction on the individual traces (generally, a

reduction in total horizontal stress on excavating at least the lower portion of the pile and

an increase in total horizontal stress on concreting) is clear. Overall, pile installation

resulted in a reduction in total horizontal stress to below the in situ value, the magnitude of

the reduction generally decreasing with distance from and along the wall.

The changes in stress measured due to the installation of a single pile can be compared

with values calculated using an elastic analysis, assuming as an upper limit that the change

in stress at the pile bore is from the in situ stress to zero (as the pile was initially bored

172

with no support). The change in radial stress, ~()r, at a distance r from the centre of the

pile, is related to the change in stress at the pile bore, ~()PB, and the radius R of the pile

bore by Equation 5-9 (see for example Fjaer et ai., 1992).

Equation 5-9

The installation of any pile affects the stress state of the ground at the positions where

further piles are to be installed. Assuming that at the depth under consideration pile

installation results in a reduction in stress, then at the centres of piles yet to be installed,

the stress in the direction perpendicular to the wall increases (due to arching) and the

stress in the direction along the wall decreases. Calculation of the changes in stress seen

by a spade cell due to the installation of piles after the first pile is therefore more

complicated. A lower limit to the change in stress may be estimated by assuming that the

in situ stress at the positions ofthe piles not yet installed is reduced (as happens to the

stress in the direction along the wall). An upper limit may be estimated by assuming that

the in situ stress increases (as happens to the stress in the direction perpendicular to the

wall).

Consideration of the Mohr circle of stress shows that the change in horizontal stress acting

on the plane ofthe spade cell, ~crsc, due to the installation of any pile is given by Equation

5-10, where e is the angle between the normal to the wall and the line joining the pile to

the spade cell (Figure 5-21) and ~crr is the change in radial stress at the spade cell (as

given by Equation 5-9).

~crsc = ~crr X cos 28 Equation 5-10

Figure 5-22 shows the measured and calculated changes in stress, normalised with respect

to the in situ stress plotted as a function of the distance along the wall (in pile spacings)

between the spade cell and the nearest pile being installed in a particular pile installation

period. Due to the complexity of the installation sequence, the following rules have been

adopted:

1. Where two piles were installed simultaneously at an equal number of pile spacings

from a spade cell, the change in stress measured at the spade cell attributed to the

installation of a single pile was assumed to be halfthe total. For example two piles,

each four pile spacings from spade cell 3, were installed during Period A, so half

the measured total stress change has been attributed to each.

173

2. Where two piles were installed simultaneously at distances from a spade cell

differing by one or two pile spacings, the change in stress has been halved and

assigned to a pile at the average distance from a spade cell. For example piles were

installed 2 and 3 pile spacings from spade cell 12 in Period E, so half the measured

total stress change has been plotted as a change due to pile installation 2·5 pile

spacings away.

3. Where in any installation period the difference in the distance between the spade

cell and the nearest pile installed and the second nearest pile was more than two

pile spacings, the entire change in stress has been assigned to the nearest pile. For

example, the two piles installed in period A were two and six pile spacings from

spade cell 2, so the change measured at spade cell 2 during period A has been

plotted as a change due to a pile installation two pile spacings away.

The change in stress was calculated over the period from 2 hours before the pile

installation process began to 2 hours after it finished. Figure 5-22 shows that the reduction

in total horizontal stress measured on installation ofthe pile in line with a spade cell was

approximately 0 to 20% of the in situ value 1·275 m from the edge of the wall, 3 to 9%

2·375 m from the edge ofthe wall and 1 to 5% 3·475 m from the edge ofthe wall.

Figure 5-22 shows that the measured reduction in stress due to pile installation is generally

larger than that calculated using the simple elastic analysis - even assuming zero pressure

(rather than the pressure of bentonite or wet concrete) in the pile bore and an in situ stress

corresponding to the upper limit value resulting from the installation of previous piles.

This could be a result of shrinkage ofthe concrete (possibly due to thermal effects) as it

sets, andlor vertical arching and stress transfer below the bottom of the bore, which is not

taken into account in the simple analysis. The underestimation of the measured reductions

in horizontal stress is confirmed by Figure 5-23, which shows the calculated and average

measured reductions in stress (expressed as a percentage of the in situ value), for piles

directly opposite the spade cell (i.e. in Figure 5-21 a distance of zero pile spacings along

the wall).

Figure 5-24 shows the measured in situ and post-installation total horizontal stresses at a

distance of 1·275 m from the wall. The best-fit linear approximation to the in situ total

horizontal stress is given by Equation 5-7. The best-fit linear approximation to the post­

installation total horizontal stress measured by the instruments 1·275 m from the wall is:

crh (kPa) = 7·4 + 16·7 z (m). Equation 5-11

174

On the basis of the best-fit linear approximations, the reduction in total horizontal stress

ranges from 10% at the top ofthe Atherfield Clay to 17% 15·3 m below ground level.

However, this masks the true nature of the variation with depth of the change in effective

horizontal stress, which is much greater (about 30%) at the mid-depth of the wall than at

ground level or at the toe (approximately 4%) as shown in Figure 5-25. This is consistent

with the transfer of lateral stress to the soil below the wall (sometimes referred to as

vertical arching), which results in an increase in horizontal stress below the toe as shown

by Ng et al. (1995). It also suggests that an attempt to quantify the effects of wall

installation as a uniform with depth percentage reduction may be too much of an over

simplification.

5-5-8 Effective stress changes due to wall installation

Figure 5-26 shows all the in situ effective horizontal stresses and the post installation

effective horizontal stresses measured 1·275 m from the wall. The best-fit linear

approximation to the in situ effective horizontal stress below the groundwater level of 1·2

m is given by Equation 5-12, and the best-fit linear approximation to the post-installation

effective horizontal stress below groundwater level is given by Equation 5-13.

cr'ho (kPa) = 15·3 (z - 1'2) + 10.86 (for z ~ 1'2 m). Equation 5-12

cr'h (kPa) = 11·6 (z - 1.2) + 5·52 z (m). (for z ~ 1·2 m). Equation 5-13

The change in effective horizontal stress calculated from the linear best-fit approximations

to the measurements is 30% 3·3 m below ground level, falling to 25% 15·3 m below

ground level (Figure 5-27). However, the linear approximations again mask the true

pattern, and the actual measurements show a similar trend to the total stress changes

(Figure 5-25) with the greatest reduction occurring at 8·3 m below ground level. The

percentage changes in effective stress are greater than the percentage changes in total

stress because there was no overall change in pore water pressures.

Current guidance given in CIRIA Report C580 (Gaba et al., 2003) is that installation of a

bored pile wall might be expected to reduce the in situ earth pressure coefficient by about

10%. This is largely based on work the work by Symons & Carder (1993) (see Section 2-

2-5) who measured a reduction in lateral earth pressure coefficient K from 2·1 to 1·9

(~1 0%). At Ashford the lateral earth pressure coefficient reduced from approximately 1·1

to 0·8 (~25%), on the basis ofthe linear idealizations to the data indicated in Figure 5-26.

175

Although there appears to be a large discrepancy between these changes when expressed

in percentage terms, they are consistent with a similar absolute reduction in total or

effective horizontal stress of about 0'25 yz in each case (where y is the unit weight of the

soil). This suggests that the likely reduction in earth pressure due to in situ wall

installation could be estimated in absolute terms. Given that the process might be expected

to be controlled by the boundary stresses at the pile bore this is perhaps surprising, but

may be more credible than an approach based on a reduction in Ko in percentage terms.

However, it has already been noted that idealisation of the percentage stress change as

uniform with depth may be inappropriate.

5-5-9 Total and effective horizontal stress changes after wall installation

Concern is sometimes expressed that the in situ horizontal stresses become re-established

in the long-term. In this case, measured total horizontal stresses showed no tendency to

change for a period of 10 months after wall installation (Figure 5-28), suggesting that the

reductions in horizontal stress due to wall installation are permanent. This is consistent

with long-term observations of horizontal pressures behind a number of in-service

retaining walls, which show no significant change in total horizontal stress over periods of

between 5 and 24 years after construction (Carder & Darley, 1998).

Although the in situ earth pressure coefficient of the Atherfield and Weald Clay at this site

is low relative to other overconsolidated clay deposits, the reduction in total horizontal

stress was similar to that for a clay with Ko = 2 excavated under bentonite, assuming a

water table at ground level.

5-6 Effect of excavation in front of the wall on horizontal earth stresses

The total horizontal stresses and pore water pressures measured by the spade cells over the

main cutting construction period are shown in Figures 5-29, 5-30 and 5-31 (for spade cells

1·275 m, 2·375 m and 3·475 m behind the wall respectively) and Figure 5-32 (in front of

the wall).

The Figures show that excavation of the cutting caused reductions in total horizontal stress

at all depths and distances behind the wall. Pore water pressures also reduced in all cases.

In front of the wall much larger reductions in total horizontal stress were measured. Base

176

Slab Construction, however, caused an increase in horizontal stress, particularly in spade

cells 11 and 13 which are located only 1·5 m below the base slab.

The compatibility of the total horizontal stress data, reinforced concrete prop loads and

bending moment data was investigated by analysing the data collected on Day 517, just

after Excavation Phase 1. Bending moments were calculated by taking moment up the

wall from the toe by consideration of it as a free body diagram. A spreadsheet was used to

find estimated values of total horizontal stress, within 10% of the spade cell

measurements, which gave a wall bending moment profile which was a best-fit to the

measured bending moment profile, using the least squares method (i.e. the sum of the

difference between the calculated and average measured bending moment squared was

minimised).

The measured and estimated horizontal stress profiles on Day 517 are shown in Table 5-2.

Table 5-3 and Figure 5-34 and the measured and computed bending moments in Piles X

and Y on Day 517 are shown in Table 5-4 and Figure 5-33.

Depth BGL, Spade cell Total horizontal stress, kPa Estimated/ m No. measured, (Yo

Measured Estimated

0 ground level

3.3 1 57 62 110

5.3 2 20 22 110

8.3 3 52 57 110

11.3 4 162 146 90

15.3 5 300 270 90

20.25 Wall toe 431.3

Table 5-2: Measured and estimated total horizontal soil stress behind the wall

177

Depth BGL, Spade cell Total horizontal stress, kPa Estimated/ m No. measured, 0/0

Measured Estimated

6.95 Exc level 0

8.3 82

11.6 11 (13) 101 (107) 111 110

15.3 (12) 17 (162) 331 329 99

20.25 Wall toe 386

Table 5-3: Measured and estimated total horizontal soil stress in front of the wall

Pile Depth, m Measured bending moments, kNm Calculated bending

gauges Pile X Pile Y Average moment, kNm

pile top 0.00 42.0

0.50 58.1

25&26 2.24 102.1 102.1 114.4

3.30 148.6

23&24 3.72 223.5 247.5 235.5 192.2

21&22 5.30 380.0 375.3 377.7 283.7

19&20 6.79 292.8 254.0 273.4 311.4

6.95 298.7

17&18 8.30 222.4 199.8 211.1 254.8

15&16 9.76 306.3 214.6 260.5 225.7

13&14 11.30 149.5 130.0 139.8 181.2

11.60 164.5

11&12 12.70 72.8 35.0 53.9 76.2

15.30 -129.8

9&10 14.20 -49.2 -49.2 -63.0

7&8 15.70 -165.7 -98.2 -131.9 -138.8

5&6 17.20 -94.8 -83.7 -89.3 -111.5

3&4 18.70 -43.7 -43.7 -41.5

1&2 19.70 6.0 -30.8 -12.4 -6.3

pile toe 20.25 0.0

Table 5-4: Measured and calculated bending moments

The RC prop load that would occur under the estimated horizontal soil stress profile

presented in Figure 5-33 can be found by analysis of the horizontal equilibrium, and yields

178

a load of 986 kN. The loads in the RC props indicated by the data presented in Section 4-

4-3 are listed in Table 5-5, along with the loads measured directly before and after the

period of 16 days during Excavation Phase 1 when no excavation was carried out in the

instrumented section. During this time the RC props loads increased by between 168 and

512 kN, and the majority of this apparent increase in load is probably attributable to creep.

Load after Load before Portion of Load excavation excavation load partially indicated after

phase lA, kN phase 1B, kN due to creep, Excavation kN Phase 1, kN

Day 489 503 - 517

PI 857 1250 393 1226

P2 555 1076 512 1406

P4 129 297 168 594

Table 5-5: Increase in prop load over period where no construction activities WCI'C

carried out during Excavation Phase 1

If the measured prop loads were reduced by the amount indicated by the increase in load

over the period of no activity at the instruemented section, the loads in RC props PI and

P2 would compare favorably with those calculated in the simple equilibrium analysis

carried out. The measurement ofRC prop loads is not straightforward and although further

refinements could possibly be made to these data, it is beyond the scope of this work.

5-7 Conclusions

Total stresses measured using push-in spade cells in Atherfield Clay may be corrected for

installation effects by subtracting 0·35 x the undrained shear strength, Cu. This is generally

less than corrections determined for spade cell installation effects in previous studies in

London Clay.

Laboratory calibration ofthe spade cell indicated a very high stiffness in comparison with

the surrounding soil, suggesting that errors due to over-compliance should not be

significant even in a stiff clay. In the experiment reported in this paper, discrepancies

between the measured vertical stress and the overburden could be accounted for by a

combination of cell installation effects and the mobilization of shear stresses on the nadir

sump chamber walls.

179

A comparison of total horizontal stresses measured in the self-boring pressuremeter tests

with spade cell measurements adjusted by 0·35 eu correction showed close agreement.

Measurement of the changes in total horizontal stress and pore water pressure during

installation of a contiguous bored pile retaining wall forming part of the Channel Tunnel

Rail Link at Ashford has shown that:

1. Pore water pressures fell during pile excavation and then increased during

concreting. The magnitude of the changes decreased with increasing distance from

the pile. In theory, the change in pore water pressure should be zero for a circular

cylindrical cavity if the surrounding soil behaves elastically. However, ditTerent

geometries would produce different theoretical results: e.g. in plane strain

l1u = 0·5 I1crh, close to the trench.

2. Overall, wall installation had no effect on pore water pressures, which rapidly

returned to their in situ values on completion of the process. This is in agreement

with previous observations for both diaphragm and bored pile walls. In contrast,

there was a very clear reduction in the effective horizontal stresses and lateral earth

pressure coefficients.

3. The in situ total horizontal stresses did not re-establish during the 10 months that

elapsed between wall installation and excavation in front of the wall.

4. Although the soil almost certainly yielded during installation, the size of the plastic

zone was small. However, the measured changes in horizontal stress due to

individual pile installation were generally greater than those calculated in a simple

elastic analysis, even assuming zero pressure at the pile bore rather than the

pressure of bentonite or wet concrete. This may have been due to shrinkage of the

concrete due to thermal effects, and/or vertical arching not taken into account in

the analysis.

5. The profiles of reduction in total and effective horizontal stress were not linear

with depth, but greatest at the mid depth of the pile. This is consistent with the

transfer of lateral stress to the soil below the wall (sometimes referred to as vertical

arching). In this case study, wall installation caused a reduction in total horizontal

stress varying from about 4% at the top and the bottom of the wall to

approximately 30% at mid-depth. Attempting to quantify the effects of wall

installation as a uniform with depth percentage reduction in total horizontal stress

may be an over-simplification.

180

Figure 5-1: Spade cell

181

Compression chamber

(a) Calibration equipment for total stress measurement

(b) Calibration equipment for pore water pressure measurement

Figure 5-2: Spade cell calibration equipment

182

Figure 5-3: Equipment for drilling horizontal borehole

183

N

+ spade cell

. ,

. ,

00 / SBPM test 00 I

location: 250 mOO /

/°0 cut-and-cover box

00 00

instrumented \0

" . nadir sump

.

section: 600 m \

,---0 , . ,

cut-and-cover box .

contiguous piled walls diameter = 1050

,,41.15 mAOD ----j,f---~

Upper Atherfield Clay

r--------------- ____ sz ~~~1 mAOD

'. ::g-~~_l\~~-.~ .. -,

, spade cell . , "

4 . ,

7500 4

,

T 13575

,

~

.

, .

,

f-------------

. ,

4 •

'---

lower Atherfield Clay

_ ___ __ \.Z 2?:~1 mAOD

Weald Clay

measure ments in mm herwise stated unless ot

Figure 5-4: Nadir sump plan and elevation

184

Figure 5-5: Spade cell: uncovered in excavation

185

......... 00 0'\

450

400

~ 350 .!II::

Q) ... ::s I/) 300 I/) Q) ... a. ... .$ 250 CIS 3: Q) ... &. 200

oI!S I/) I/)

~ 150 -I/) CIS (J

1:: 100 Q)

> n; -0 50 I-

Total stress

L I II II II

IIIIIII I II II II

J IILlII IIIIIII j IIIIII IIIIIII

. IIIIII Excavation 1 a-j I I II

I III III Excavation 2 - I .LII I I

I IIIIII

Excavation 3 I I II II I I I I"H I I

. I IIIIIII Excavation 4 ---LJ II II I I

Pore water I pressure

I TiTl"N I I I IIIII I I I IIIIII I I IIIIII I I IIIII I I IIIIIII I I IIIIII

II I~ II Excavation 5 II II II II

Increase in stress in response to ring beam installation

!~ I

o i j;:=;= ",,"" I" I~ I I Ii: ' 1.1

6 6 620 627 641 613 634 648 655 662 669 I

-50 ...L 11th June 2001

Time in days

Figure 5-6: Total vertical stress measured by spade cell in nadir sump

676

41.145

41.145

Excavation 1

'. //lL3L3§§_J7

! Spade cell 32.590

spade cell

r+~ 00000000~ o o 8

Excavation 4

36.860 3§.109 34.300 v: /7L _____ -I

I Spade cell 32.590

41.145

41.145

o spade cell

OQ/fooo 8 <:) o

~~ 8 ~ 0

00000000 o o o o

Excavation 2 Excavation 3 41.145

[ . .. , _~U42 _____ y-

, I 35.~05 _ 36.8~.: I

7777T/'L_" /:,1 34.698 !. !

! ! ,. ,

I'f Spade cell Spade cell I' : 32.590 32.590 i I

t·]

Q 0 o 0

go)f<000)-OO 8 ?;rooooo~~ 6 spade cell 6 b 0_ 0

8 tYA 8 S oooooo00tS'oB o· 06~s

G 0 o 0

8 8

Excavation 5 Installation of ring beam 41.145

__ ~4.~3.L ___

I Spade cell

I Spade cell

32.590 32.590

8 8 00000000008 00000000008 o 0 0 ........ "0

S 8 spade cellifj. 0 g O -----e-- . 0

- 0 :.'0 8 8 8···· . .'. '" ;',8 000000<00000 88000000000

o 0

8 8 o 0

all measurements in m AOD

.\'

.v

Figure 5-7: (a) Elevations and (b) plans detailing the excavation of material above spade cell in nadir sump

187

CIS D.. ~

Cii C)

CI) "C

250

200

[ 150 II)

c o II) II) e U;

S 100 .s "C e :::l II) CIS CI)

:E 50

o

after excavation 4 .. '

after excavatiofl' 5

LJ1 .' 1

excavatjon under ring beam

50 100

.'

before excavation

.'

150

over-read due to installation

200 250

Total stress acting on spade cell due to overburden, kPa

Figure 5-8: Measured total vertical stress versus overburden

188

0.0009

0.0008 ----------------,----------------- ----------------- ----------------- --------- -------,----.--- ....... .

E E 0.0007 en en CI)

~ 0.0006 CJ .=. -CI) "C ns

0.0005

~ 0.0004 a; CJ CI)

"C ns c. en

0.0003

.= 0.0002 CI) C) c ~ 0.0001 u

-0.0001

Calibration chamber fluid pressure, kPa

Figure 5-9: Nadir sump spade cell stiffness calibration

600

189

E C")

N V

50 m

Figure 5-10: Finite element analysis mesh

Position of spade cell

190

250

225

ca 200 a.. ..liI:

IIJ IIJ 175 Q) ~ -IIJ

ca (.) 150 1:: Q)

> '0

125 Q) -~ :::s

..=! ca 100 (.) ~

0 '0 ~ 75 :::s IIJ ca Q)

:!E 50

25

o

LJ1

25 50 75

-FEA

-A- Nadir sump spade cell

-+- Adjusted FEA

100 125 150 175 200 225 250

Overburden, kPa

Figure 5-11: Measured and calculated vertical stresses against overburden

191

150 Hmax

a=O.4 125

100 a=0.3 -ra a. ~ - 75 ... 0 ... a=O.2 ... w

50

a=0.1 25

a=O o ~~----~------~------~------~~-----.

o 2 4 6 8 10

Height of soil above spade cell (m)

Figure 5-12: Over-read error - measured and calculated from limit equilibrium analysis

192

5

25

30

Total horizontal stress, kPa

o 100 200 300 400 500 600 700 800 900

\ \ \ \ \

\ux. \ \

\ • . \ \ \ \

.~ \ \

\ \ \

,\~x

~ \ \ \ ...

\

.. ~ ... .&<x

• \.

\ \ \ \

\ \ . \

• • \ ~ Overburden

\

• • •

\ \ \ \ \ \ \ \ \ \

• Pressuremeter 3593 - 250 m from nadir sump ... in situ total stress (spade cell): corrected x in situ total stress (spade cell): uncorrected

Figure 5-13: Comparison between spade cell readings corrected by 0-35 Cn and self­boring pressuremeter test results

193

Figure 5-14: Spade cell installation

194

as D. ~

700

600

en 500 III ~ u; 400 S c ~ 300 'i: o = "i 200 '0 I-

100

85

810

810

84

no data for this period

812 817

\ -...

vlv ,,:: ......................................... -----___ 85

812 S13 89 1\ / 811 \ </. .......................................... _84 ____ _

1 .. ·8,-6· .. · .. :::::::::::::::" .. · .. · ...... ·· -~ 816

~~m-:::::::::::::::::::::::::::::::::::::::::::::==f-==~===== 815 ,...-v ~:--3~::!:

58 'S'~2 515 ;'~~"'::' "',::,' ,,:::::::::::: :::::::: ::~'~~89~~~8§1~~ S1

O~~rr~~rr~~"+.~-.r."-.rr,,~rr,,.-r.,,~-.ro~-.,,~

as 300 D. ~

~ 200 c. ... CI) .... ; 100 ~ 0

D. 0

o 5 10

0 5

15 20 25 30 Time in days

(a) Total horizontal stresses

no data for this period I

15 20 25 30 Time in days

(b) Pore water pressures

35 40 45

35 P1 40 45

50

P13 P3 -

50

Figure 5-15: Spade cell measurements taken between their installation and wall installation

195

0

0

1

2

3

4 (2 data points) 5

E 6

-a; 7 > 8 .9! "C 9 c 5 10 ... C» 11 == .E 12 CI)

.c 13 .c -g. 14 c 15

16

17

18

19

20

21

Total horizontal stress, kPa

100 200 300 400 500

¢ Total horizontal stress

t:. Pore water pressure

Hythe Beds

- - - - - -Sesf::fitlfrle -to -po-re water pressure data between 1.2 m and 11.6 m bgl

u = 8.6 z - 10.5

Best-fit line to total stress data aha = 20.6 z + 0.8

upper Atherfield Clay

lower Atherfield Clay

¢ ¢ ¢

Underdrainage Weald Clay

Wall toe

Figure 5-16: Stabilized readings of horizontal total stress and pore water pressure from all spade cells and piezometers, plotted against depth below original ground

level (before wall installation)

196

o Effective horizontal stress, kPa

100 200 300 400 500 O~~~~~~~~~~~~~~~~~

1

2

3

4

5

E 6 a; 7 ~ 8 -g 9

e 10 C) 11 ~ .2 12 Q)

.c 13 .c .. g. 14 C 15

16

17

18

19 20 21

o Effective stress

Hythe Beds

. Best-fit line to in situ effective

/ ·.ODD

o

Wall toe

stress data 0\0 = 15.3 (z - 1.2) + 10.9

upper Atherfield Clay

·· ... Iower Atherfield Clay

o 0 ".

~Weald Clay ..... ~

Figure 5-17: The profile of effective horizontal stress a' ho with depth (obtained by subtracting the measured pore pressure from the total horizontal stress at each

location)

197

CIS C. ~

150

i ;:100

e c. .. ~ 50

~ o c.

spade cell D.1 no. and /\ ,~, ,G,,Q,, , ..... 1;.... , ... f .... ,.Q. ,1-:1.. , ..... 1.. ... depth b.g.l~ : . .

P5-152!P'~: :: :~. : :::: : : :: :~~ 11..:jj a""': f'.+.....: Y·.: : .:.. ::: \ P5 - .j.UIEfE\""\·->~-----"':·- -;--:: --~ ---- :----~:.:--+~~-:---------""P11 P4 11 3 : . : :: :~:---- \ ---: ::: : :

- . Ol :: :: : : : : ::: :: . P4 P3-8.3m : .:: :..:: : : : ::.: :: : P3

"'-'---_." ~--.: : :'1 : : : : '.-. -.. -.. ~:.. ::: ~"""P12 12 153 . . ., . :- . .. . . ......... .. . ................ . ~: : :: : :: .:-< : ::.....: :: ~ ... ~ .. :: : P2

P2-5.3m : : :: ::: l:: .... : : :,:: :: ' P1 ---r:-:; :~', . -:-:-, ~ : P1-3.3m ::: :: ... :... ..

O+-~~-r~~~~~~~~~~~~~~~-r~~~~~~~~

150 CIS c. ~

as' .. ;:100 I/)

~ c. .. CI)

50 ... ; CI) .. 0 c.

0

150 CIS C. ~

i =100 I/) I/)

~ c.

43 50 57 64 71 78 Time in days

(a) 1-275 m from the wall

spade cell A )? ,G p~1 E F G H I no. and ' .... , , ... __ ...... ""--"'-''-'' i'- , ...........

, .. depth b.g.l: : ' .. , .. · .. ~:: ' .. , ..

' .. · .. ' .. ' .. · ..

P17-15.3rr1 ' .. , .. , ..

.. P16-8.3m 17 - .. 16 :

: P15-5.3m .. 15

"

43 50 57 64 71 78 Time in days

(b) 2-375 m from the wall

spade cell no. and ,.A.,.e D91

E ,'0;; i ...•.....• I ............ C

i"'

depth :: I •

~::

43 50 57 64 71 78 Time in days

(c) 3-475 m from the wall

Figure 5-18: Piezometer readings over the period of wall installation

198

period of wall installation

P1-3.3m :

beginning of main construction

~P5

: P4 ~P11

; P12 ~P3 ~P2

~P1 0~~4-~~~~~~~~~~-r~~~~~~~~~~~~~~~

ns 150 a.. ~

a) ... ;100 III ! c. ... CI) ... 50 ; CI) ... 0 a.. 0

... CI)

50 ... ; ! 0 a.. 0

4 32 60 88 116 144 172 200 228 256 284 312 340 368

4

4

period of wall installation

P16-8.3ri.1

P15-5.3m

32 60 88

period of wall installation

Time in days

(a) 1·275 m from the wall beginning of main

construction

~P17

~P16

~~P15

116 144 172 200 228 256 284 312 340 368 Time in days

(b) 2·375 m from the wall beginning of main

construction

~~~~;::;:::;.nh~~ru:::JI!!~~:::=::::::~;;:::::::. = P10

P8-8.3m; P7-5.3m; P6-3.3

32 60

: P13

~P8 ~P7

~P6

88 116 144 172 200 228 256 284 312 340 368 Time in days

(c) 3·475 m from the wall

Figure 5-19: Pore water pressures measured before, during and 10 months wall installation

199

450

400

~ 350 .lII::

~ 300 ~ ... II) 250 S c 2 200 'i: o J: 150 S ~ 100

50

6 7 8 9 10 -------------------------------.. -------.. -------.. -------e-------e-·

1100 15 16 ------------------------------.. ----------------e--------------------

1100 1 2 3 4 5 --------------------------e-------e-------e------... ------... ------

1800

@®®©@®@®®@@©®®0©®®0®® 1800 11 12

------------------------------.. ----------------.. ----------------------1100 __________________________________________________________________ ~---- • Spade cell

11 00 _______________________________________ ~--------------_________________ Dimensions in mm

spade cell A o. and :"':

depth b.g.l~ :

B c

8.1.?:~! .~:~~. ft· .~ ... j .......... : .. :: . " · ... I I " I :' , , I I , ,\

001 E F G H -_.--._-,". -----_ .. _- ... . , I •• , , , , II • , , • II • , . • II • , . • II • , , , .. , , , , ... , . • tl • , , · , .. , , · ,. , , , • t •• , , , ,. , . . I '" I.

85

: : :: : :r' -:' A::: . . . ..:: ., , 84-11.3m, , " , " ''', , , , .. ,' " .' 812 811 11 6 N.'

" , " ,," . ' ................... " .. ,,' '.' .. :' ...... . - . m ~, 'M'::: ,: .. ' : :: :~ ... : .. !' .. : ...•... :

------..-- ., ,. " , .. ~ , "." " ,

: J-~-\i-----~ -~- ; .. ;..... ! -f:-:! ~ : -r-- 84 83-8.3m, (. " " ~I-""----\ '-----'- ...... -.,..~" '--- 811

HJ' : :: : : : 1. : ~:------: :::: ~-:-----: .

81 33 ::: '~' , :: : : : :::: :: : 83 ---' _m : : : : : : " : : : ~ : 81

8=-2--=-5.-=-3m- : G f ~ ~~; . =4+-fF~ : :: 82 I • • • ". I, ,

I , , • ". I, ,

, , . ... '" ... , , . , .. , , .. • , , , .. I '. I • , • , .. I " I , .. , .. , ., ,

O+-.--r~~'-r'~·~'._-.~~~~,--r~~--~.-_r~"T'_,·~'~·_.~~._._~

43 50 57 64 71 78 Time in days

(a) 1,275 m from the wall

spade cell A B 450 o. and :"':

C 001 E F G H ----_ .. _.-_. _." .. __ .". __ .-- -._ .. -- .. _--.

depth b.g.l.

400 817-15. 3m-' -v-"""""'-""-"""""IJ'---':Il':---r---_, - ________ -i--~........;;.-v---:-""':-~ 817

~ 350 .lII::

~ 300 ~ ... II) 250 S c 2200 'i: o J: 150 S ~ 100

50

"

=816 : 815

816-8.3m

=~' [ffi::,' •• :. • •

815-5.3m. ; • ..;...' ___ : : : , . , ,. , I , , " " ,,' ,

"

O+-.--.~~--~~.-~-r~'+' ~--~~._~~~~--~+_.__r~~~

43 50 57 64 71 78 Time in days

(b) 2,375 m from the wall

200

450

400

~ 350 .:.::

~300 e -CI) 250 S c 2 200 'i: o J: 150 S ~ 100

50

6 7 8 9 10 ------------------------------.-------.-------.-------.-------.-.

1100 15 16 ------------------------------.----------------.--------------------

1100 1 2 3 4 5 -------------------------... ------... ------... ------... ------+------

11 12 ------------------------------.----------------.----------------------

11 00 __________________________________________________________________ ~---- • Spade cell

1100 13 Dimensions in mm ---------------------------------------.-------------------------------

spade cell A B C D D1 E F ,;~ H

o. and ;' ; .. .. . .. .. ; ;' .. " ..

depth b.g.l: : :

: : :

: : : :

: : : : :

: : : : :

: : :

: . . .. . '" , , ..., , . S13-11 6rrj : : : : : : :. : : : :::: : : S9~13'~;';' :-·:~j···i··,li .. l ......... d~~ ...... ~ .. ~ : :: .......... S13 . : ': :: : ~ : t : V: .-: '- ,,1:;"'; .. :" .:...... --

8-83m : : :: : : : :: : : : :::: : : S9 • -<----..,;" '\~i:: ' . """

S7-5.3m . . : ~ : r.---:- v.- ,~~ ., ,. S8

;::;;;::::;;:=~ ~, :,' :, :,:, :, :: =S7 S6-3.3m . , . " . ~ S6

, , I ., ,

, I I " I , , , II ,

, .... , , .... , , , I •• ,

I , , II ,

I • , II , ..... , • I , .. , ..

O+-~-.~~~~~_.~~~+_.__.~~,_~.__r~~~~.__r~_.~

43 50 57 64 71 78 Time in days

(c) 3·475 m from the wall

Figure 5-20: Total horizontal stress measurements taken during the period of wall installation

Any SP~]d~e~ r~, _._.-.-. ·_·-e-e·ee·-

Distance alo.ng wall,. I I I I number of pile spacings 0

2 3

--~r-~----~------~~~~cr

~crr ~cre

Figure 5-21: Pile installation sequence for the elastic analysis and Mohr circle showing calculation of correction for stress change measured on spade cell

201

0.5

.5 0.4 -In In

0.3 e In In .... CI) In a.. c .... ._ In 0.2

C =:s o:t::

:;::: II) (.) 0.1 :::l ~ CI) 0 0::

-0.1

0.5

.5 0.4 -In ~ In 0.3 !:; ~ In!:; C In 0.2 c,2 o '-:;::: II) 0.1 :::l

x

•

0

• •

•

o S1

• S2 x S3

• S4 • S5 + S11 x S12

--- average -- elastic solution

- ------;----x 3 4 5 6 t 8 9 10 11 12 Distance along wall (in pile spacings)

• S15 • S16 • S17

--- average --elastic solution

~ ~------~ 0~~~~~~~-4--~---+--~--~_r-_~ __ ~_~~~~~--'---' • o 1 2 3 4 567 8 9 10 11 12

-0.1 Distance along wall (in pile spacings)

0.5

.5 0.4 • S6 -• S7 • S8 x S13

--- average --elastic solution

o 1 234 5 6 7 8 9 10 11 12 -0.1 Distance along wall (in pile spacings)

Figure 5-22: Measured reduction in total horizontal stress normalised with respect to the in situ total horizontal stress (showing prediction from elastic analysis)

202

- 100 c 0

90 :;:; I '" measured I .!! Ci 80 -I/) .5 Ci~

70 ~ 0 ~

60 0 I/) -I/) Calculated maximum Q) ~

50 extent of plastic zone =-'C I/)

I/),a 40 I/) .-

Q) I/)

b c I/) .- 30 .5 c 20 0

:;:; CJ

10 = .. ~ 'C

'" Q) -.- ... -.- .. ~.- ..... r::t: 0 0 1 2 3 4

Distance from edge of pile, m

Figure 5-23: Reduction in total horizontal stress due to installation of pile nearest to spade cell only with distance from the wall compared with the elastic prediction

203

0

0

1

2

3

4

5

E 6

a; 7

~ 8 'C 9 c ::s

10 e tn 11 := .2 12 CI)

.Q 13

.s::. -g. 14 c 15

16

17

18

19

20

21

Total horizontal stress, kPa

100 200 300 400 500

¢ Before pile installation

• After pile installation

Hythe Beds

• ¢ Best-fit line to O"ha measurements ~ O"ha = 20.6 z + 0.8

Wall toe

Best-fit line to O"h measurements 1.275 m from wall after pile

•

installation O"h = 16.7 z + 7.4

upper Atherfield Clay

lower Atherfield Clay

Weald Clay

Figure 5-24: Total horizontal stresses measured by all spade cells before wall installation and by those 1.275 m from the edge of the wall after wall installation

204

0 0 1 2 3 4 5

E 6

Q; 7 > 8 .!!! "C 9 c e 10 CO) 11 ~ .2 12 CI)

.c 13 J:: ... g. 14 c 15

16 17 18 19 20 21

,

1> I

20

, , , ,

I I

I I

, , , I

I I

?

40 60 80 100

- Reduction from linear best­fit approximations

- ~- Reduction in total stress based on measurements

ythe Beds

upper Atherfield Clay ---~--------------------------------

I I

I I ~

I I

I

I

lower Atherfield Clay

Weald Clay

Wall toe

Figure 5-25: Change in total horizontal stress

205

0

1

2

3

4

5

(j) 6 > 7 .Sl! "0 8 c :::s 9 e C) 10 ~ .2 11 CI)

.c 12 E S£ 13 -g. 14 Q 15

16

17

18

19

20

21

0

Effective horizontal stress, kPa

100 200 300 400 500

a Before pile installation I', "

. \.. \. • After pile installation ' .. .... \. .'~ 0 Hythe Beds

" " -..::.- - ------------- - - - -------- - - --Best-fit line to cr'ho measurements

before pile installation cr'ho = 15.3(z -1.2) +10.9

Best-fit line to crh'

measurements 1.275 m from the wall after pile installation

cr'h = 11.6 (z -1.2) + 5.5

upper Atherfield Clay

.... lower Atherfield Clay '.

-----------~----~--------~----------o ll') Weald Clay

Wall toe

Figure 5-26: Effective horizontal stresses measured by all spade cells before wall installation and by those 1.275 m from the edge of the wall after wall installation

206

0 0 1 2 3 4 5

E 6

4i 7 > 8 ~

"C 9 c = 10 e C) 11 ~ ..2 12 Q)

,g 13 .c -g- 14 Q 15

16 17 18 19 20 21

20

, ,

40 60 80 100

- reductionfrom linear best-fit approximations

-i3- Reduction in effective stress based on measurements

Hythe Beds ------,-- --------------------------, ,

I _______ L

I I

I

rf

I I

I I

I

0, " ,

" " "

" " " " " "

Wall toe

"

"

" " " " " "

.P

upper Atherfield Clay

lower Atherfield Clay

Weald Clay

Figure 5-27: Change in effective horizontal stress

207

450 period of pile installation

1--1

Beginning of main construction period

400 85-15.3~ : 1 1 1 1

ns ~350

:2300 ~ .... ~250 S 5200 N 'i:

,g 150 C; 0100 I-

50

812-15.3m

84-11.3m

811-11.6m

~--~~----~-------------: 85 --------....--1-1 -1 1 1 1 1 1

1 812 -~--------~ ~

1

~J~t1-4 -~-------.; 811

1

:= ::::=::::: :~ 2 3

~-S-1 :-s2 O+.~~~r+~~~~~~+O~~~.+~~~~~~40~~~~~~

4

450

400 ns D..350 ..:.:::

:2300 ~ .... ~250 S 5200 N 'i:

,g150 ns 0100 I-

50

0

4

32 60 88 116 144 172 200 228 256 284 312 340 368

period of pile installation

1--1 1

1 1 1 1 1 1 1 1 1 1 1 1

S16-8.3m 1

Time in days

(a) 1·275 m from the wall

~ 15-5~:::'=:======: == I. 1

Beginning of main construction period

1 1

: 817 ----------<""1" ~

32 60 88 116 144 172 200 228 256 284 312 340 368 Time in days

(b) 2·375 m from the wall

208

450

400

CIS D..350 ~

~300 Q) ... +I

~250 JS 5200 N ';: ,g150

CIS cHoo I-

50

0

4

period of pile installation

1----.1 I I 1 1 I I 1 1 1 I 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1

S13-11.6 : 1

--------------~--~----------

S7-5.3ml 1 1

S6-6.~:::· ::::::::::::::::::::::::::::::-:::.:::.:::::=:::::::::;:::;::::::::::~:;::::::::

Beginning of main construction period

1 I I 1 I 1 I 1 1 1 1 1 1 1 1 1 1

~ S13 -:---v-- _~ ~

S9 i S8 ~-

1 S7 ~---

1 -so 1 1

1

32 60 88 116 144 172 200 228 256 284 312 340 368 Time in days

(c) 3·475 m from the wall

Figure 5-28: Total horizontal stress measured before, during and 10 months after wall installation

209

400

350

50

.0 II ro II

- II Cf) ...... 11 Q) (/)11 (/) CII ro 011 coo

o +-~~-+~-.'-~r-r,~-r~~+-r.~-+-.,-.-~~~-r~.-~r.~-.~ 410 438 466 494 522 550 578 606 634 662

Time in days

(a)

(/) C 350 I CII C I I tl I 1091 I II

I .QII.Qtld ~~ I mltl II

I t)11 t)rold 1=1 rol'O II I ::J II ::Jii)Iii1 I rol ...... 1 C II ro

EI L..I Q.\ 1 ...... 1 (/)1 Q) 300 L..II L.. I 5 I ~INlN II > rolC ii) II ii)T"" 1"-; T""I..-I II 0 Q)lo Cll CQ)I~

Q)I ~ :p I -I Q)I Q) 511 ro E CO I.- 011 o (/)1 ~ ro ('1')1 (/)1 (/) .0 ._11 > Q)

It) ..... 1 0 0:: 0 ~:1l = 1-1 ro l ro ~ gl E 0:::

CIS 250 O'll ::J ..r::::1 J9 '01..r::::1..r:::: Il. .!: IlJ NII..;;ta..IO,J

a..:111 ~ c: a..:a.. Cf) lJ I Q) ('I')

~ 0.1 (/) a.. II a.. .1 I Q) ~: 0::: I-o.lC _II -ul (,) <ild rol UIU (/)

N ~200 rolO T""II ('I') XI ><l XI ><lIT"" NI XIX ro 01 T"" 010 a.. II a..WIW WIW l- I-! WIW co 0 I l- I-

~ I II I I I I I I I I/) I II I I I I I I I/) I II I I I I I I CI) I II I I I I I ~ 150 I II I I I I I

I II I I I I I ... I II I I I I I CI) ... I II I I I I I

; 100 P5-f'"V"1~r~ I I I I I I I

... ~ I e P4--I P5 0 I II

Il. 50 P3~ ------------ P4 I II

P2~ P3 0

P1 I II P2 1

4 0 438 466 494 522 550 578 606 634 662 -50

Time in days

(b)

Figure 5-29: Total pressure and pore water pressure measured 1·275 m from the back of the wall during the construction period

210

400 (/) C C C t 0 0 .Q t d CIl ~t

I 1:511 - CIlO iiil -gIl 1= I CIlI"O -'II 1-350 E:

::J II g enen ~I Q)II I~ I iii I ~ (/)11 ICIl l:; II !:; I I I IIC :~:N:N 511 I>

CIlIC (/)11 (/) '<"'j"-I "-1"-110 0 11 - 1 0 Q)IO CII C QlQl Q)I Q)II~ 1- I Q)I Q) II CIl IE ca OJ I+:;

011 0 ~~ ~I~II= I~I(/)I(/) .011 > IQ) c.. 300 ell (.) Oil 0 ..el..ell CIl I I CIlI CIl CIlIIO 10::: ~ CI2 Nil..,. a.;a.; 0....

10....

11 iii

1"0 l..el..e USII E I('t')

1$ '0.. I iii 0...._:: 0...._ ~~

I IIC IClo....lo.... Q)II~ :1-

~250 o.lC r,jI r,jIl- :~: ~:~ (/)11 CIllO ..-11 ('t') Xl Xl 1..- CIlII..- I -

IN ... 0:0 0.... 11 0.... WIW::I- :I-:W:W OJ:: I- II-.... II til II I II I I I II I

S200 II I II I I I II I II I II I I I II I

C II I II I I I II I 0 II I II I I I II I N II I II I I I II I

'§ 150 I II I II I I I II I

S1~ I II I I I II I .c II I I I I

S I II II I I I I I II I I I I I

0100 I II I I I- S1~-

I S16

50 I II I II S15 I II I II I II

0 410 438 466 494 522 550 578 606 634

Time in days

(a)

350 (/)

gO C 0 I :t .• 1 I c' I

0 0

I . :2 ~ I . CIlI II I d t l I I

I 10.-1"011 1._1 I "0 I I I (.)1 (.) 1.8 lGl CII I rol ~I C I I-

E: ::J I ::J rn. I (/) ,-I Q)II C I =1 (/)1 Q) I ICIl

300 .!:J I l:; ..-I I I> 1..- ..-1..-11 0 I CIlI N I N 1 0 CIl I C Ull (/) I I I 11._ I iiil I C 1-Q)I 0 Cil C ~ I Q) Q)I Q)I I ro I ci Q)I Q) 0 ICIl IE OJ I +:;

8:8~ I(/) (/)1(/)11_1_1 (/)1 (/) .0

1:5 I> IQ)

O'J(.) I CIl CIlI CIlII ro I ('t')1 CIlI CIl CIl 10

10::: ca 250 cI2 l..e ..el ..ell _ I 1-1 ..el ..e US ::J IE I('t') c.. ~: ~ 0..; 0.... '-

.~"@ 10.... 0....10....11 (/) I 10....1 - IQ) ~

Icj r,jIr,jIlEI~1 r,jI Q) (/)

10::: :1-I j cj (/) C e200 ~I cvi ~ I X Xl Xii ..- I N I Xl CIl 0 1..- I -CIlI

0 X IN 0 10 ~:o.... :w W:w:: 1-: 1-: w: W OJ 0 :1- II-:::s til I II I I I II I I I I I til I II I I I II I I I I I CI) I II I I I I I I I I I

5.150 I II I I I I I I I I II I I I I I I I ... I II I I I I I I I CI) I II I I I I I I I ....

; 100 I II I I I I I I I I II I I I I I I I I II I I I I I I I

CI) I II I I I I I I I ... P16 I II I I I I I I I 0 50 ~ I I I I I I c.. I II I I I I I

P15 I II I I I I I 1 ________ II P16 I II I I

0 4 0 438 466 494 522 550 578 606 63415

-50

Time in days

(b)

Figure 5-30: Total pressure and pore water pressure measured 2·375 m from the

back of the wall during the construction period

662

662

211

400

350

50

C t "0 ot"O 2 C:;:; 2 C en O).!Q en 0)

EI ...... 11...... c-rol~ enllen...-J.,-l..-I...-JloI2Ir...lN 0) '"' I ~ I 11._ I en I '"1

10) CIIC <U ~III~ICI <U 0) co I _ 0 II 0 ~ v, I I W I I ~ en 0)0 OliO ~I Irol~1 ~ cI2 NII-q- o..;n' 0.. 1 0..;1 t)11-1n:0.. 'gj ~ 0.. : : 0.. '"'1" I I C I I '"'"1

rolo ......... 11(1)~I~ ~:~::::~:~~ 0:0 0..::0.. Lij..,. ill :t:y: 1-:1-: t.Y ill

I ,.11 I I II I I

: S10:: : ::::: I II I I II I I I II I I II I I

S9 I II I I II I I ~~ .1.",,-,11 I III I I

I II I I II I I I II I I II I I I II I I II I I I II I I II I I

S8 ___ ....r,..,..11 1:1:1 : ::::: ~:~ II1I1 S7 --'""'- .~ll- '-/"~- II I I ~- I ------- --rr---- I I I

II

§ "ro .0 ._11 > roullo

- :::JIIE U) 'pliO)

~ ~::o:: ro 0 11 ..-co 0::1-

II II II I

I I I I I I I I I I I I

ro I> 1

0 IE 10) 10:: 1("1)

:1-

IN

:1-I I I I I I I I I I I I I I I I I II S6

~~~~~~~~~:~: --~~~~~:saS7 S6~

I II I II I II II

O+-,,-.~~-.~.-~~,L~-+,L~r+-..-~~~~r.-.,-r.-,~

410 438 466 494 522 550 578 606 634

350

300

~ 250 ~

2f 200 :I III III Q) c.. 150 ... ! ; 100

e £. 50

c.. § . ; I .91 :;:;tl d. : ~: g~:f1

EI Cil.... I VI rol en I t)"-I"<"i 0)1"0 Cli C 0)1 ill .01 0) QlI 0 enl iii ....l U a. 1 0 rol ro gl:::J II -q- .cI..l:J

._1 .... NI 0.. 0..10l 01- D..il .1 l r-o. enc 0 " ..... ~I (1) I....., rol 0 ..-jl 0.. illXI X 01 0 D..il I ill

I II I I I II I I I II I I I II I I I II I I I II I I I II I I I II I I

Time in days

(a)

. c .. I II 1011:1 1"011 1:;:;1 CUI "0 I CII I rol ~ C I 0)11 C I =1 v~ 0)

"-1..-IIOI2INN I II:;:; I enl aj

~I ~II.!Q I EI in ~ rol roll ro I ("1)1 CUI ro .cl.cl l _ II-I .0 .c 0..1 OJI en I 10..10..

.1 .II CI~I I 01 011 - I I d cj XI XII ..- I NI ><l X illilllll-ll-lWill

II I I II I I II I I II I I II I I II I I II I I II I I II I I P10 I II I I

___ ~:~ .r--",--I ~--- II I I I I I

I

II II II II

c"-0" ro .0 --II > ro UII 0E

- :::JII U) .PliO)

~ ~::o:: ro 011"-1-co Oil

II II

I 1-lro I> 10 IE 10)

:0:: 1("1) II­I • IN II­I I I I I I I I I I I I P9U"'1I.Jt r II

, I II ~ ____ --":_~I ---~-P1 0 P~~ :"''W ~ .. - - 159:

662

P I II I :

O~p~~~~~~~--~~~~~~~~~~±;~I ~~~=E~~~~;e~~~ 4 0

-50 438 466 494 522 550 578 606 662

Time in days

(b)

Figure 5-31: Total pressure and pore water pressure measured 3·475 m from the back of the wall during the construction period

212

400

350

50

ro >

S1~ ~ Exc,PI~ .Qll ro> I~ E: ~t:l 0. I:: ~ I ..0 0110 10) ml e .II. ro a I II I I I c75 ::JII E 10:: 1D:2 Cii:: Cii 1111 : 53:: e~ I '!:;II 0) 1«) oj g 5115"""""" IT""II.Q :u;:\ : 3l ~::o:: :1-_ e I '- 011 0 ..J..J I II ..... I I I ro 011 T"" IN ·-1..... II ±-1±-1 l..cll~ lEI I ID 0111- II- S17 g; ~ Nil <;t Q..jQ..j 100001Iro 1«)1 I :: ~ (010 0...110... 99 I .IICiiII-I :~ ~ 0:0 T""1:C'"it.Ljt.Lj :~::E:~: I II I

I 0...110... ~W::T"":N: : ::: S12~ ~ I I 1"1.111-11-1 I II I

I II I I (/)1 II I II I I II I I tl II: :::

S13 ,.! ~NI:~::: II I S12 ~I II I I (/)1 II II >...A ..... """""''W-\O;,\o S11~n ~III I II I I I II

I II I T"" I II I II I 0)1 II I II I (/)1 II I II I rol II I II I oCl II I II I 0...1 II I II I .1 II I II I 01 II I II I XI II I II I WI II

II II

"'rt\\I\~~#WII"'IIS13 .". S11

O+-~ro-+~-r,-~ro~~-r~~+-r.~-r-.,-~~-.~-r,-~~r.-,~

410 438 466 494 522 550

300

250

CIS 200

~ ~ 150 ~

~ Q)

Q. 100 ... Q) -; 50

~ o

Time in days

(a)

(/)

c·· e

~~ 15 II i 5i ti I 0110 I ~II:;:; ID "011 I .-1 J!ll "0 I 011 ° ell I rol Ie

EI 211 2 ~~ 1-1 0)11 I =1 (/)1 0)

ro:e +'II ..... ...-1..-1 ...-I T""II 5 I rol NI N

0) 0 ~:~ U I 0)11 :p I Ciil 0)1 0)

ID L- 0110 !(/)::~:E: (/):(/) I ..... OliO 0i10 ~II ro I «)1 ~I ~ el2 Nil <;t a.; 0..." Cii I 1-10...10... ·c..ICii 0...:: 0... II e I I I o.le

~~ ~ 011 - I ~I cil ci rolO ~I «)- Xii T"" I NI Xl X

0:0 0...:: 0... w:: I- : 1-1 w: W I II I I II I I I II I I II I I I II I I II I I I II I I II I I I II I I II I I

P17 I II I I I II I I ~~~II I I

P13~N I:: I I I I

P12------r ::: I ::: I I I I

I II I I I II I I II I I I II I

578 606 634 662

II i II I II l-II l ro

I> II 10

5 II ro IE ..0

._11 > 10) 011 0 ro ::JII E 10:: (ij .!:; II 0) 1«) 0) (/) II 0:: :1-(/) ell ro 0 1 T"" IN ID 0 I- II-

~ 0 +-~,,-+L,-.,-~ro~L+~~~~~Lo-r~~~~-,~-r,-~~-,-,~

4 0 -50

-100

438 466 494 662

P12 P17

Time in days

(b)

Figure 5-32: Total horizontal stress and pore water pressure measured in front of the wall during the construction period

213

-1000 o

Bending moment, kNm -500 0 500

~ Calculated 2 bending moment

0 PileY

4 " Pile X

IZ1

6

8

10

12

14

16 ~

-t-' c::::

18 Q)

E E

0 E

20 .c 0); - c:::: Co :so:: CD ()

C C\l L..

22 0

1000

Figure 5-33: Measured bending moments and bending moment calculated from horizontal soil stresses within 10% of those measured

214

Horizontal stress, kPa

800 600 400 200 o

2

4

6

8

10

12

14

16

18

20

22

+ Measured horizontal soil stress

-- --- Estimated horizontal soil stress •

+:

~:

E .s:: -c.. Q)

C

o 200 400 600 800

RC prop

Excavation level

.,

+ •

Figure 5-34: Measured and estimated total horizontal stresses

215

6 LONG TERM MONITORING

6-1 Introduction

In this chapter the long-term measurements of total horizontal stress and pore water

pressure in the soil around the wall, and the reinforced concrete prop loads and pile

bending moments in the structure, are presented and discussed.

6-2 Total horizontal stress and pore water pressure measurements

Figures 6-1 to 6-6 show the total horizontal stresses and pore water pressures measured

over the period from before pile installation to late 2005 by the spade cells 1,275,2'375

and 3·475 m behind the wall respectively. Figures 6-7 and 6-8 show the total horizontal

stresses and pore water pressures measured in front of the retaining wall over the same

period. These figures show that installation of the contiguous piled retaining wall and

excavation of the cutting caused the main significant changes in total horizontal stress and

pore water pressure. Since construction of the cutting was completed the total stress and

216

pore water pressure measurements have generally remained constant from year to year,

although a few spade cells have recorded small overall reductions.

Figures 6-1 to 6-8 show that in the years following construction of the cutting the total

stress measurements, and to a smaller extent the pore water pressure measurements,

experience seasonal variation with temperature, particularly those close to ground level

behind the wall and in front of the wall. The recorded loads are smallest in winter and

higher in summer for the spade cells 1·275 m behind the wall, with the highest readings

occurring between approximately July and October. The cells 2·375 m and 3·475 m behind

the wall lag behind this slightly, with the highest readings occurring between

approximately September and November. In front of the wall, the readings from the

shallowest spade cells (SII and S13) are highest between approximately September and

November and from the deeper cells between October and February.

The total stress cells appear to suffer from noise, probably due to temperature effects (the

deepest spade cells exhibit less noise). The pore water pressure transducers appear to be

less affected than the total stress transducers.

There have been a number of other significant changes worth noting. The readings from

total stress cell 4 appear to have shifted down by about 10 kPa over the period between

approximately Days 1630 to 1966 (Figure 6-1). Total stress cell 8 appears to have

malfunctioned (Figure 6-5) and total stress cell 12 performs unusually below about 160

kPa.

6-2-1 Installation of storm drain

Figures 6-1 to 6-6 show that before installation of the storm drain behind the instrumented

wall (see Section 3-5-6 and Figure 3-48) the pore water pressures measured in gauges P 1-

P4, P15-16 and P6-8 increased. This was during late October and early November and was

therefore probably due to seasonal rainfall. Installation of the storm drain generally caused

a fall in pore water pressure. Ever since the storm drain was installed the pore water

pressures have fluctuated very little by comparison. The total stress measurements in the

same gauges also increased over the period before the storm drain was installed, although

generally to a smaller degree, apart from the total horizontal stress measured in spade cell

1 which, if anything, dropped slightly over this period.

217

Over the period of storm drain installation, spade cell (SC) 1, 3·3 m below ground level

and therefore above the bottom of the storm drain trench, recorded a reduction in total

stress of about 15 kPa on excavation of the trench, and on backfilling the stress retmned to

its pre-trench excavation level. SC6, at the same depth as SC 1 but much closer to the

storm drain trench, recorded a reduction in total stress of approximately 50 kPa due to

trench excavation, recovering to 20 kPa. These changes can be seen more closely in

Figures 6-9, 6-10 and 6-11, and are essentially similar to the type of change exhibited as a

result of pile excavation.

6-3 Reinforced concrete prop loads

Figure 6-12 shows the reinforced concrete (RC) prop loads, calculated assuming the

Young's modulus of concrete is constant at 25 MPa, measured from before excavation

within the cutting to late 2005. The average temperature measured in the gauges is also

shown. The figure shows the load in the props increasing by between approximately 200

and 1000 kN each year.

Figure 6-13 shows the RC prop loads adjusted for the effects of creep (after Day 607) as

described in Section 4-4-3. The minimum load recorded each year is constant (in

approximately March to June), and the maximum load (recorded in October each year) is

reducing, showing that the seasonal variation in load is reducing each year.

As more data is collected in the long-term it should be possible to improve the correction

for the effects of creep.

6-4 Pile bending moments

Figures 6-14 to 6-17 show the pile bending moments measured over the period from

before excavation within the cutting to late 2005 by the gauges in Pile Y and Pile X

respectively. The buried gauges exhibit generally constant measurements with time but the

others vary significantly with temperature. Similar bending moments are exhibited in

gauges 15-24 in both piles, except in Pile Y gauges 21&22 and in Pile X 15&16 (and to

some extent 17&18) are higher, because the section is cracked at these locations.

218

6-5 Conclusions

The long-term spade cell measurements are showing generally constant total horizontal

stresses and pore water pressures over the years following construction of the cutting, and

in a few cases overall reductions being recorded. There are some seasonal and daily

variations in the readings, occurring more in the shallower spade cells than the deeper

ones, which are probably due to temperature.

Correction for creep in the reinforced concrete props has produced prop loads which are

generally constant from year to year. These loads vary seasonally like the spade cell data,

and the seasonal variation is reducing from year to year.

The bending moment data are generally constant with time and the effect of cracking on

the pile section is clear in the long-term data.

219

tv tv o

ctS D.. ~

400

350

300

en 250

~ .-I/)

ctS 1: 200 o .~ ... o

.s::. ~ 150 {:.

S5

S4: :

S1

100 t S2

50

2000

Pile installation period

2001 2002 2003

.. Base slab poure

Storm drain installed

~'lJ~ l' ~V~L . . .

o I .' .' \ I I . ' . I I I -50 11 8 286 454 622 790 958 1126 1294

Time in days

2004

1462 1630 1798

2005

S5

1~.J S4

1966 2134

S1 S3 S2

2302

Figure 6-1: Long-term total horizontal stress measured in spade cells 1·275 m behind the wall

tv tv >-'

C1l Il.. ~

eli ... :::l I/) I/)

~ Q. ... Q) ..... C1l 3: Q) ... 0 Il..

300

250

200 , Excavation

period ]--: :1-- ---l: '---150 ] " I , , 1 : . I " Base slab poure

100

~ P ~ P5

50

I ~~~::t:~~~~~~~~~~~ P4 ~ P3

o P1 ~r:::::::::+::::::~::t~:::=::::~±:=::::::::::: P2

P1

-50

2000 2001 2002 2003 2004 2005 -1 00 I I I I

-50 11 8 286 454 622 790 958 11 26 1294 1462 1630 1798 1966 2134 2302

Time in days

Figure 6-2: Long-term pore water pressure measured in spade cells 1·275 m behind the wall

tv tv tv

I'll a. .lII::

350

300

250

Ii 200 1/1 ~ -1/1

I'll 'E 150 o .~ ~

o .s::::: n; 100 -{:.

50

2000 2001

E~cavation period

.-.----

2002 2003

, .. Base slab poure

: .. I Storm drlain installed

2004 2005

~~~~~~~~~~~~~~~~~~MW~~ S16

S15

o ~ I I I I I I

-,·0 18 286 4$4 622 79 958 1126 1294 1462 1630 1798 1966 2134 2302

-50

Time in days

Figure 6-3: Long-term total horizontal stress measured in spade cells 2·375 m behind the wall

N N w

ca c..

350

300

250

~ 200 ~ ::l til til Q) a. 150 ... Q) .... ca ~

~ 100 o c..

50

o

2000

Pile i~stallation perio

-----.

2001

Excavation period

.-.--

2002 2003

.. Base slab poure

: .. I Storm driain installed

-50 I I I I

-50 11 8 286 454 622 790 958 11 26 1294

Time in days

2004 2005

"""t~....-.r-r----------..~_ ............ ~-_ P16

~P1 5

1462 1630 1798 1966 2134 2302

Figure 6-4: Long-term pore water pressures measured in spade cells 2·375 m behind the wall

N N .j:>.

300 ,

1

2000 J 2001 2002 2003 2004

Pile i

250 t peri? , ---+; :~ ,,, .

S10

2001 S9 : : '

m t 1 . . ." : I Base slab poured

--:,,_L'\~~ ' 't o ••

D. ' , : \ . : • . Storm drain inslalled ~ 150 S8::

"'~' I ~ .•. Ifj , : :: ~ .. Ifj , , Q) ,

: ~J~ : ... . 1ii ~7 ' , , .. '

III 100 -I: 0 ' .. 1 ......-~.A. ~ .. , ...... N '':: 0

.c:: III 50 + " I ' , 1 : ~. _ .. --liv-.L. ~._~J. ....... ...IIA..I -0 I- "I

1 :

' ,

1'1.. 0

- 0 18 286 4 4 622 79 95 1798 1966

-50

-100

Time in days

Figure 6-5: Long-term total horizontal stress measured in spade cells 3·475 m behind the wall

2005

,LIL~ 89

. • S7

S6

2134 2302

S8

N N VI

ctI c..

300 2000 :2001 2002

250 + : : I Pile installation period ,

~ '.<--- I " .. Base slab poured 1 : : ~ ..

200

.ll' 150 I , , I ---. , ____

~ ::s

2003

~ : .. I Storm drain installed

2004 2005

~100 ~~':: : CI) , , , ,

.... P9 1 '" , ! P8 : ' ' ~~ : ~'"r'" ' '''''''' ' .. _- ~" 1" -~~" r" P10 '- 50 ' , , " r" , '" -" ... P9 o P7 ' . ~' , ,

~lp6~~l O ~~ " ....... oII!eAt ........ "... P8

" , , " ·"~P7

P6

~ : : I I : -50

-100 I I I ,', I I

-50 118 286 454 622 790 958 11 26 1294 1462 1630 1798 1966 2134 2302

Time in days Figure 6-6: Long-term pore water pressures measured in spade cells 3·475 m behind the wall

tv tv 0'\

'" c.. ~

400

350

300

III~ 250 III ~ -III '" 'E 200 o

.!:! ... o

.£:

~ 150

r=.

100

50

1h :2001

Excavation period

, , , ,

\ , , , ,

'r-' hi • I-~ I~n: : • ~~ 11 ~i

, ,

~~

o I i 'i' l I I I I I

-50 11 8 286 454 622 790

2002 2003 2004

958 11 26 1294 1462 1630 1798 1966 Time in days

Figure 6-7: Long-term total horizontal stress measured in all spade cells in front ofthe wall

2005

S17

~S12

S13

S11

2134 2302

N N -....l

ca C. ~

~ ::s til til ~ Co ... Q) .... ~

300 I

250 +

200

150

100

' ' I 2000

: : I Pile installation period

--;- :-----

Excavati

p~

.1

410,

:2001 2002 2003 2004 2005

,. Base slab poured

,-- :--t---Storm dr+in installed

~ 50 o r-"" P13 P11 c. t

I~

o 1 :': 1 I : l: j . ......,.,,.,..~ . ~~_ ~L _::::;;;;l_ ,- P1 2

P17

-50

-100 I 1'1 I I 1'1 I I

-50 118 286 454 622 790 958 1126 1294 1462 1630 1798 1966 2134 2302

Time in days Figure 6-8: Long-term pore water pressure measured in all spade cells in front of the wall

400

350 datalogger was

refurbished

~300 n: ~ ~ ______________________ /---__ --______ --v ~ ______ ~S=5~

~ 250 ~ Storm drain -tJj

,s200 I: o N

excavation began

S4

'§ 150 . ~~-~--

J:

ca '0100 I-

50~============~==~~~ ~ ~ : S2

O~~~~~~~~~~~~~~~~~~~~~~~~~~~~~

689 696 703 710 717 724 731 738 745 752 759 766 773 780

Time in days

(a)

300

250

-50

:-Storm drain

data logger was ~ refurbished :

excavation began

-100~~~~~~~~~~~~~rrn~ho~~~rlo~oh~~~~~~

689 696 703 710 717 724 731 738 745 752 759 766 773 780

Time in days

(b)

Figure 6-9: Total horizontal stress and pore water pressure measured 1·275 m behind the wall over period of storm drain installation

228

300

250 data logger was :

CIS refurbished ~.' Storm drain ~ 200

j 150 J.-________ -----------rk installed 89 I/)

~ 100 l: o N

~ 50 t:=::::.==:::::=~~=====:::=:::=:::::::::::::::::::==~:::::::~=::':::::;~-:::- :::=-::::: S7

~ S6 o O+-------------------------------------------~--------~ I-

-50

-100~~rtT~~~~~~~~~~hn~~~rrn~n.~~~n+~.rrnT~

689 696 703 710 717 724 731 738 745 752 759 766 773 780

Time in days

(a)

300

250

~ 200 ~- Storm drain ~

cD 5 150 data logger was: excavation

refurbished ~ began I/) I/) Q)

Q. 100 J-___________ ~~--~n~ ~---P-10

P7 -50

-100~~~~~~~~~~~~~~~~~~~~~~~~~~~~~

689 696 703 710 717 724 731 738 745 752 759 766 773 780

Time in days

(b)

Figure 6-10: Total horizontal stress and pore water pressure measured 3·475 m behind the wall over period of storm drain installation

229

300

250

ns c.. 200 .:.I::

If) If) Q) 150 ~ -If) ns 100 -c 0 N .~

50 0 ..c C; - 0 0 I-

-50

-1 00 760

S9 ,e e e e e e e e e e e e e e e e e e. e e e .... 8 e I B 8 Gee e e a e e e e; 1 G I Geeea e a e e, e II a e G e De e s 8 8888 e e ee e e e eo e •• e e a e e 8. a eo. I.ee. eta ea ... I. ell .......... u a U

761

Excavation of trench on one day

762 763

Time in days

Further excavation on next day

764 765

+ S7

766

Figure 6-11: Immediate response of spade cells nearest to excavation of storm drain trench

230

L{) (")

o (")

L{) N

o N o L{)

I

o T"""

I

o N

I

L{) N

I

o (")

I

L{) (")

I

~~~~~~~~~~~~~~~~~~~~~~~~~~G£G

o o L{) (") T"""

~ .. --·1il.~~ -_.

--~~.reo_ - ~ . . .~~;.:,-. ~~~ Il)~-::. __ o · --:-.~

o .---~---.~~,

N .-~§f . -:..;::;ta:-: .. :..-- _ . . ~.~ . . •.

-.:t o o N

o o L{) N T"""

o o L{) ...... T"""

o o L{) o ......

o o L{) (j)

.--;:;~~. ~c.i?~--

o o L{) CO

o 0 0 o 0 0 L{) L{) L{) t- CD L{)

(N)!) peol

o o L{) '<t

o o L{) (")

o o L{) N

o o L{) ......

o o L{)

~LGG

S~GG

6S~G

£O~G

LPOG ~66~

S£6~

6LB~

£GB~

L9L~

~~L~

SS9~

o o L{)

I

231

13500 l T 35 2001 2002 2003 2004 2005

12500 30

11500 25

10500 20

9500 ~IJ 15

8500 ~~ J .J II IIII 10 u

[500 pI ' ' 1~~ " , ~1W'" Inrl'V ~"'V r 0 .

Temperature ~ ::::I

-g6500 o 10 0

-5 ! -' 5500

Q)

-l- -1 0

~

4500 ~ Co~struction Period

3500 -15

2500 -20

1500 -25 P4

500 -30

-500 ~ I.{) ..- "- (\') 0> I.{) ..- "- (\') 0> I.{) ..- "- (\') 0> I.{) ..- "- (\') 0> I.{) ..- "- (\') 0> I.{) ..- "- (\') 0> I.{) ..- ~ -35 (\') 0> ~ 0 I.{) ..- "- N co (\') 0> I.{) 0 CD ..- "- (\') co ~ 0> I.{) ..- CD N "- (\') 0> ~ 0 I.{) ..- "-~ I.{) I.{) CD "- "- co co 0> 0> 0 o ..- N N (\') (\') ~ ~ I.{) I.{) CD "- "- co co 0> 0> o ..- ..- N N (\')

..- ..- ..- ..- ..- ..- ..- ..- ..- ..- ..- ..- ..- ..- ..- ..- N N N N N N

Time in days tv Figure 6-13: Long-term reinforced concrete prop loads corrected for the effects of creep w tv

N W W

1800

1600

1400

1200

E z

Pile Y - Gauges below level of formation/base

slab .

II: J :Construe on Period

1 I ...

~ 1000 .... c: Q)

E 0 800 E en c: "C 600 c: Q)

OJ

400

200

o 1-~ ~ 2001

-200 416 528 640 752 864

2002

976 1088 1200

RC prop

,. Temp prop

Base slab

• 13 & 14 • 11 & 12 • 9 & 10 • 7 & 8 • 5 &6 • 3 &4 • 1 & 2

~ 13&14

:-

r-~~&4 7&8

'- 5&6 2003 2004 2005

i I I ill Iii iii iii I I iii I I I

1312 1424 1536 1648 1760 1872 1984 2096 2208 2320

Time in days

Figure 6-14: Long-term bending moments in Pile Y - gauges 1-14

N w ~

1800

Pile Y - Gauges above

1600 j. level of formation/base slab

1400 I :Construction Period

1200

E z .lII:: 1000 ~ c: Q)

E 0 800 E Ol c:

"C 600 c: Q)

m

400

200

o 1- ~ 2001

-200 416 528 640 752 864

RC prop • 25 & 26 • 23 & 24 • 21 & 22

. Temp prop ____ __ • 19 & 20 • 17 & 18

: Base slab • 15 & 16

. - - -

~ 21&22

15&16 ~--..,...~~~~~~~W' 17&18

19&20

2002 2003 2004 2005

976 1088 1200 1312 1424 1536 1648 1760 1872 1984 2096 2208 2320

Time in days

Figure 6-15: Long-term bending moments in Pile Y - gauges 15-26

tv W Vl

E z

1800

1600

1400

1200 j I bonstruclion Period

Pile X - Gauges below level of formation/base slab

RC prop

Temp prop

Base slab

• 13 & 14 • 11 & 12 • 9 & 10 • 7 & 8

J.I:: 1000 • 5 & 6 ..s c: Q)

E 0 E C) c: '0 c: Q)

m

• 3&4 • 1&2

800

i VP'l ~~ , ,

600

~ ~==~~~:~~ 400

200

:tf o+ptt'~ : _ ~: --== ,V-~p, , ~------------------~----------------

: 2001 : 2002 : 2003 2004 2005 -200 , ',' "

7&8 1&2 5&6

416 528 640 752 864 976 1088 1200 1312 1424 1536 1648 1760 1872 1984 2096 2208 2320

Time in days

Figure 6-16: Long-term bending moments in Pile X - gauges 1-14

tv w 0\

1800

1600

1400

1200

E ~ 1000 ...:-c Q)

E 0 800 E Ol c :c 600 c Q)

III

400

200

0

-200

Construdion Period Pile X - Gauges above level of formation/base

slab

RC prop

• 25 & 26 • 23 & 24 • 21 & 22

:ITemp prop====== ~. 19 & 20

• 17 & 18 :1 Base slab ===:j. 15 & 16

~ 15&16

I

Time in days

Figure 6-17: Long-term bending moments in Pile X - gauges 15-26

7 CONCLUSIONS AND RECOMMENDATIONS

The fieldwork described in this thesis represents one of the most extensively monitored

embedded retaining wall case studies undertaken, particularly outside the vicinity of the

London Clay. Total horizontal soil stress and pore water pressures have been logged

continuously since one month prior to installation of the contiguous bored pile wall,

throughout construction of the propped cutting and into the long-term (6 years to date).

The effects of wall installation, excavation ofthe cutting and the possibility of the in situ

stresses being re-established in the long-term have been assessed. Measurement of the pile

bending moments and the permanent and temporary prop loads has enabled the wall to be

analysed and the overall long-term stress state to be determined.

Detailed conclusions have been drawn at the end of chapters 4, 5, and 6; more general

conclusions arising from this research are stated below and recommendations for further

work arising from this fieldwork are made.

237

7-1 Conclusions

• This study is a complete case record linking the wall behaviour to the change in

horizontal stress in the soil, both behind and in front of the wall.

• There was no reinstatement of the total horizontal stress in the soil after either wall

installation or excavation of the cutting.

• The work undertaken in order to evaluate the spade cell over-read due to

installation in stiff overconsolidated clay adds significantly to the previous data as

spade cells have not previously been used in the Atherfield and Weald clays, and

the performance of a spade cell under a known changing load has not previously

been measured.

• The process of installing an in situ embedded retaining wall in overconsolidated

clay deposits produces a very clear reduction in the effective horizontal stresses

and lateral earth pressure coefficients. The reductions between 15 and 50% were

measured and the reduction was not linear with depth. This is a very significant

result for walls embedded in overconsolidated clays as they are regularly designed

to accommodate unnecessarily high horizontal stresses under working conditions.

• Identification of concreting cracking and analysis ofthe effect this has had on the

bending moment profiles calculated from the inclinometer measurements has

explained the differences in bending moment plots found by two different methods

after cracking had occurred.

• Structural monitoring is regularly used in construction of large projects to

minimise construction costs and maximise safety. Inclinometer measurements are

regularly taken to monitor wall movements. The success of the calculation of

bending moments and prop loads at the CTRL will provide confidence that

structural monitoring can be used more extensively than at present.

• The work undertaken to separate the strain due to shrinkage, creep and applied

load in the reinforced concrete props has allowed prop loads to be estimated which

agree with the measured wall bending moments and total horizontal soil stresses by

means of a simple equilibrium calculation.

238

• There has been no long term increase in the bending moments, reinforced concrek

prop loads or horizontal soil stress.

• Although continuous monitoring is demanding, the benefits of collecting

continuous data using vibrating-wire gauges connected to a datalogging system are

numerous and must form the basis of any future fieldwork studies.

7-2 Recommendations for further work

This research has highlighted several areas of further work which will build on the

important conclusions made above. Suggestions for this work are made in the following

paragraphs.

• Development of a data management system format that assimilates construction

events with the collected instrument responses would be highly desirable.

• The properties of the Atherfield Clay are not well understood. Further studies to

help define its geotechnical parameters would be most welcome. In particular, the

effect of the shear surfaces and fissuring on the permeability and the pore water

pressure changes should be addressed.

• Concrete shrinkage and creep have been studied extensively in the past, however

understanding the effect these factors have on the measurement of prop loads usi ng

embedded strain gauges is obviously crucial. Distinguishing the strain due to load

from the shrinkage strain and creep strain proved difficult in this case, particularly

because, due to the construction events, the props began to take up load very soon

after the concrete was poured. Analysis of the data still being collected will allow a

better understanding of the shrinkage due to creep to be attained.

• Further work should be undertaken to investigate the over-reading exhibited by

spade cells installed in overconsolidated clays. The connection made between the

over-read and the undrained shear strength could be improved upon.

• Measurement of the movement of the south wall would have been very valuable,

helping to answer questions regarding the amount of load taken up by the props,

the movement of the whole wall, issues surrounding the sway exhibited by the

structure and the ground settlement behind the monitored wall.

239

• There is a general requirement for further fieldwork to improve our understanding

of retaining wall behaviour.

• Finally, and possibly most importantly, the data collected in this field study should

be used for input and comparison of results for finite element analyses, not least to

study the overall performance of the wall; the three-dimensional nature of the

bored pile installation process and to investigate the transfer of stress under the

wall in the wall installation process which produced the non-linear stress reduction

observed in this study.

240

Appendix A

Figure A-1: Wireline borehole sample 'catch' mechanism

Figure A-2: Wireline borehole rig

241

Wireline borehole logs

borehole 1 page 1

BH 1

University of Southampton boreholes at the CTRL July 2001

note: stratum boundaries cannot be identified from this 44.99mAOD

80852.76E due to poor recoveries for the top runs. The other borehole is more reliable.

of borehole 200mm dark brown firm sandy CLAY (topsoil/made ground)

400mm brown firm sandy CLAY with rootets (topsail/made ground)

orange firm sandy CLAY with rootlets (made ground)

sandstone with shells and small black speckles

3.7-3.75 white/yellow and black/orange sandstone with shells and small black speckles

firm yellow/grey mottled sandy CLAY with a few pieces of gravel

stiff greylbrown mottled CLAY with some small orange marks

stiff to very stiff yellowish grey sandy CLAY

yellow grey/grey transtion zone. Some peds of very stiff grey CLAY with some yellow colouring. Peds have shiny surface.

stiff to very stiff closely bedded grey sandy CLAY with some small peds of very stiff material, mainly stiff crumbly matrix. Some fissure plains are shiny.

o Z ::J o c::: C> W o « :E

en o W III W J:

!;: J:

o ...J W u:: c::: W)­J:« I-...J «(j c::: W 0. 0. ::J

242

Appendix A - borehole 1 page 2

c: OJ

.><

OJ 19 OJ a: 13 ~~ oil

en - c: " g-o ::;

(t!

en

6.45

82

10.35

11.5

12.5

13.5

14.75

15.75

BH 1

University of Southampton boreholes at the CTRL July 2001

44.99mAOD

80852.76E

22462.84N

very sliff grey extremely closely bedded sandy CLAY. Brown colouring 10 some parts • may be slaining from mud flush

very sliff grey extremely closely bedded sandy CLAY. Brown colouring 10 some parts· may be slaining from mud flush. Some of this run is softened and muddy.

very sliff closely laminaled sandy CLAY. Occasional brown slaining. iron pyrile nodules? Fissures (aboul20mm 10 50mm apart) with shiny faces.

12.3rn - some small black sandy stones which crumble under strong hand pressure. shells.

material here is more sandy than above. Very small pockets of fine white sand.

very sliff laminaled sandy CLAY

lighl brown band of very sliff lighl brown CLAY. chocolate colour.

very sliff grey CLAY. plaslic. not crumbly.

becoming very sliff grey/brown mottled CLAY

very sliff brown CLAY. choclaley appearance. Very closely fissured, moslly horizontally. but some in other directions. Some rare small hard inclusions «5mm). Homogeneous material.

>­<I: ...J U o ...J W u:: 0::: w ::c I­<I: 0::: W Il. Il. :::l

>­<I: ...J U o ...J W u:: 0::: w ::c I­<I: 0::: w ;: o ...J

243

Appendix A - borehole 1 page 3

University of Southampton boreholes at the CTRL July 2001

~~--------~------------------------------------------------------~ o(l

o ::;:

16.85

20.0

21.0

22.0

BH 1

}

A piece offtint was found in top of this run, about 15mm long.

16.7m - shear failure visible in core

16.85m-light grey/brown hd. rock

very stiff brown CLAY. choclatey appearance. Very closely fissured, mostly horizontally, but some in other directions. Some rare small hard inclusions «5mm). Homogeneous material.

bottom 200mm of this run is dark grey very stiff CLAY

top 50mm of this run is light brown CLAY, like the Ath I (it's been softened by drilling). There is a sudden change in colour below it to dark brown/grey very stiff closely fissured CLAY

Honogeneous matenal.

brown band of crumbly material (2mm), pass. organic remnant? Sparkley grains up to 1 mm dia. Under the brown there is a 30mm band of light greylbrown CLAY, which is hard in DISCONTINUITY

layers of shells, about 3 in this section (250mm). Other laminations of siltlfine sand.

very stiff dark grey CLAY

50mm at beginning of run - dk gry soft CLAY (softened by drilling?)

finm to stiff light grey CLAY with laminations of silt

19.4m - bed of mUdstone up to 10mm thick - not complete layer

mainly stiff but soft in places dark grey closely bedded CLAY. Core is stained brown on edges and in open cracks.

grey CLAY laminated with silt grey stiff CLAY laminated with silt.

finm SILT - with some clay. Silt is in bands and pockets up to 10mm wide with small amounts of clay inbetween. Silt is crumbly

stiff dark grey closely bedded CLAY brown band of CLAY about 2mm wide at 20.75 - brown hard layer about 1 Omm wide with small knobbles at 20.8 - band of brown silt about 1 Omm wide silt in pockets/lams w. some clay bet. Some small orange spots dk grey stiff CLAY. 21.03-21.05 - hd brown band -20mm wide made up of thin bands. Below is 40mm of brown hd MUDSTONE. In clay above hd band there v. thin silt lams, below it are silt particles distributed evenly thru the clay. 21.2-21.23 - white silt, 21.23-21.3 - green/grey very stiff CLAY w. silt lams 21.3-21.33 - dk grey very stiff CLAY w. white silt lams white closely bedded SILT w. very closely spaced lams of dk grey clay dk grey v. stiff CLAY w. infrequent silt lams. 21.6-21.65 silt lams close together

green/dk grey CLAY w. very frequent white silt laminations (v. close bedding) 21.9-22.03 - silt lams are less frequent

dk grey CLAY w silt lams and pockets. Much mud staining on the core.

lighter grey _ 22.5 - 22.58 light grey CLAY w silt lams ~ 22.58 - 22.66 very silty band - finm, closely bedded. Some clay. Stained brown.

22.66 - 22.73 dk grey stiff CLAY w les frequent silt lams 22.73 - 22.79 lighter grey with more silt 22.79 - 22.87 light grey CLAY w thin silt lams. Closely bedded. Brown staining

23.12 - 23.2 light grey CLAY with silt lams 23.2 - 23.23 closely bedded SILT light grey stiff closely bedded CLAY w silt lams brown staining

below sample: light grey CLAY laminated with silt

200mm band of white SILT

23.85 - 23.95 light grey CLAY laminated with silt

>­<t: ...J (.)

o ...J W u::: 0:: W J: I­<t: 0:: W :!: o ...J

>-~ (.)

o ...J <t: w :!:

244

Appendix A - borehole 1 page 4

c: Q) j<!

Q) l'l 2 iii en 'iii

" 2 c: en E"0 III en

University of Southampton boreholes at the CTRL July 2001

~~------~--------------------------------------------------------~ o/l U ::;: BH 1

25.0

26.0

27.1

28.0

28.8 sample

29.7

light grey CLAY with silt lams. below 24.17 silt lams are wider. 2/3mm

Igt grey CLAY w. silt in pockets on bedding planes. Frequent but small

from 24.58 - silt in pockets on bedding planes less frequent

at 24.75 -layer of silt 3mm thick at 24.88 orange/brown crumbly layer v thin pass tree derived? very stiff dk grey CLAY with infreq. lams of silt at 24.96 v. thin lam of brown silt (?) with black speckles

dk grey CLAY v infreq white speckles also some infreq bedding planes with sparkles on. Closely bedded

200mm dk grey CLAY with some thin silt lams at 25.6m lump of mudstone 70 x 50mm dk grey circle markings on it. white/beige SILT laminated with clay v. closely bedded

dk grey very closely bedded stiff CLAY

finm - sftnd by drilling? Dk grey v closely bedded w pockets of hard silt

hard layer of brown closely bedded small knobles 15mm

dk greylbrownish (pass stained from mudflush) CLAY. some dk (whitelbrown) silt on bedding planes, but not much.

light grey CLAY closely bedded light grey clay - some v thin & infreq lams of silt

dk grey CLAY w infreq v thin white silt lams. Bedding is angled - 20'

dk grey v closely bedded CLAY with v. little silt

Igt grey CLAY. @28.30rge/brn ROCK - 2/3 dia. Of BH -100mm wide below rock: orange/brown rock fragments in soft clay missing material dk grey stiff closely bedded CLAY at 28.6 - some light grey/dark grey mottling

dk gry stiff v closely bedded CLAY w some silt pockets on bdg planes

at 28.9 - some silt lams one 2mm wide (at end of borehole section)

dk gry CLAY w some hard silt pockets. Some are quite large

at 29.23 - 29.35 - laminated SILT white/brown w. some clay grey CLAY w. lams (@ -20') and pockets of white and stained bm silt

dk grey v closely bedded CLAY at 29.65 - hard silt layer a few mm thick. Stained

>­< -l U o -l < W ~

245

Appendix A - borehole 2 page 1

Q) 0 c ~ Q)

~ ~ en c :::J 0::

0

2

University of Southampton boreholes at the CTRL July 2001

44.83mAOD

soft sandy dark brown CLAY with gravel and cobbles (smelly)

firm to stiff mottled yellow/grey sandy (fine sand) CLAY (some orange staining). Brown speckles.

very stiff fissured mottled yellowish grey sandy CLAY with orange specks

very stiff grey sandy (fine) CLAY. with brown staining (of mud flush?) horizontally thinly laminated

o WZ O:::J <0 ::EO::

(!)

CJl o w m w :I: I­>­:I:

= o ..J w>-U::j ffiu :I: I­<

246

Appendix A - borehole 2 page 2

BH 2

University of Southampton boreholes at the CTRL July 2001

Very stiff grey sandy (fine) CLAY. Less evidence of laminations. Clay is crumbly and has staining. Silty particles sparkle.

Round brown stain on core with a stone in the middle

RUns 1 and 2 very similar. In both runs large parts of the core has been affected by the drilling which has produced stained, softened and crumbly samples. The borehole was left open over the weekend before these runs were taken.

RUn 1 : large peds of very stiff sandy (fine) CLAY in matrix of sandy (fine) clay. There are fissures at intervals of about 50mm in all directions, the surfaces of which are wet and shiny and NOT stained with the mud flush. However much of this material seems to have been softened by the mud flush. Much brown staining. silty particles sparkle.

The bottom 100mm of this run is horizontally thinly laminated.

run 2 : very stiff crumbly silty CLAY with peds of very stiff silty clay Fissures are stained by mud flush.

very stiff grey CLAY. Slightly silty but less so than at lower positions in the stratum.

very stiff grey silty CLAY

Clay crumbles when cut. Lots of brown staining from the flush. The clay feels like it's softened.

First run of day after trouble with casing on day before. Top 25mm is yellow mush. Next 35mm is grey and has usual structure of Ath II but appears to have absorbed a lot of water and expanded.

Material after S 1 0 is very stiff grey silty CLAY.

material is greylbrown mottled above this line, so that the line corresponds to the point at which the clay stops being mottled.

S11 constitutes a light brown band that exists between the Ath I and II, which is very stiff light brown CLAY (choclatey appearance)

>-~ u o ...J W u:: 0:: w J: I­<t: 0:: w a. a. ::::>

247

Appendix A - borehole 2 page 3

University of Southampton boreholes at the CTRL July 2001

BH2

}

s11 is part of a light brown band which exists at the top of the lower Atherfield Clay. Underneath it there is some light brown mottling to the core. Then the material is very stiff brown CLAY. It is chocolatey in appearance and is homogeneous.

darker - the top 350mm of run 3 is darker brown than the rest, but otherwise itis same: very stiff dark brown CLAY

occasional fragments of shell 3-4mm in size

very stiff fissured brown CLAY.

brown/dark grey very stiff closely fissured CLAY.

19.55 - 20.3 very stiff dark grey/brown closely fissured CLAY. Very dark at 20.3m, getting gradually lighter and browner as depth decreases.

60mm band of brown, very hard ROCK

grey very stiff closely fissured CLAY with some silt

light grey stiff CLAY with close silt laminations

300mm very stiff grey closely fissured CLAY

light grey CLAY with close silt laminations

dk grey stiff closely fissured CLAY

light grey stiff closely fissured CLAY with less frequent silt laminations

dk grey very stiff closely fissured CLAY with silt laminations

light grey very stiff closely fissured CLAY with silt laminations

grey very stiff closely fissured CLAY no silt laminations grey very stiff closely fissured CLAY silt laminations grey very stiff closely fissured clay CLAY no silt laminations band of grey ROCK, 100mm thick

dk grey very stiff closely fissured CLAY no silt laminations

200mm band light grey very stiff CLAY

dk grey very stiff closely fissured CLAY no silt laminations

very stiff grey CLAY with silt laminations

>­« ...J u CI ...J W u. 0:: w J: I­« 0:: w ~ o ...J

>­« ...J U CI ...J « W ~

248

Appendix A - borehole 2 page 4

c: ~ University of Southampton boreholes at the CTRL July 2001 ttl (l) 0.:::

i ~ ~~--------~----------------------------------------------------------------~ ~ 5 ~ '" en

25.0

27.5

28.3

BH 2

very stiff closely laminated with silt grey CLAY. Some brown colour in laminations (mud flush?)

very stiff light grey closely laminated with silt CLAY.

dk grey closely fissured CLAY. V close silt laminations - 45° to horiz.

dark grey closely fissured CLAY.

dark grey closely fissured CLAY. With very close silt laminations 15° to the horizontal.

very stiff dark grey CLAY. With very close silt laminations 10-20° to the hOrizontal up to 3mm thick. Silt is greylwhite (or crushed fossils?)

above mudstone - very stiff greenish grey closely fissured CLAY

28.15 - Grey MUDSTONE about 4mm wide.

28.45 - silt lamination

very stiff dark grey horizontally closely fissured CLAY. Some fissures contain white silt but these layers are rare.

very stiff dark grey closely fissured CLAY. With peds about 10-20mm which are stiffer than the crumbly matrix. Some white silt (or crushed fossils?)

249

Appendix B - Piling diary

6 7 8 9 10 .. ~ ~ -- --1100 - - 16 - -1.? .. - -1100 .1 3 1 ~ 5

-1800 Z x y

+- 000000000000000000000 1800 G020 G021 G022 G023 G024G025 G026 G027 G028 G029 G030 G031 G032 G033 G034 G035 G036 G037 G038 G039 G040 .. --

N Dimensions in mm

~ • Spade cell o Pile and pile number

Figure B-1: Plan of piles showing pile reference numbers

Key - start and finish times for the following construction stages

ex. 1 = 1 st stage of excavation, 0 - 8 metres

ex.2 = 2nd stage of excavation, 8 - ~20 metres

Conc. = concrete pour

Date Time interval Pile no. and Procedure Time

Tuesday 23/11/1999 14:00

Wednesday 24/11/1999 10:00 12:00 14:00

Thursday 25/11/1999 14:00 16:00

Friday 26/11/1999 08:00

weekend

Monday 29/11/1999

10:00 12:00

14:00

16:00

G025 - ex.l G033 - ex.l

G025 - ex.2 G033 -ex.2 G025 - Conc. G025 - Conc.

G021 - ex.l G037 - ex.l G028 - ex.l

G021- ex.2 G028-ex.2 G021- Conc. G037 - ex.2 G037 - Conc. G028 - Conc.

G035 - ex.l G031 - ex.l G023 - ex.l

15:10 - 15:40 15:50 - 16:20

11 :55 - 12:35 12:55 - 13:30 14:20 - 14:53 15:50 - 16:38

14:40 - 16:00 16:50-17:10 17:20 - 17:40

08:05 - 08:45 O?:?? - 09:14 09:16 - 10:02 09:30 - 10:01 10:49 - 11 :30 12:25 - 13:11

14:59 - 15:20 15:55 -16:15 16:25 - 16:46

250

Date Time interval Pile no. and Procedure Time

Tuesday 30/1111999 08:00 0035 -ex.2 08:00 - 08:35 0031- ex.2 08:35 - 09:07 0023 - ex.2 09:34 - 10:00 0035 - Conc. 09:55 - 10:31

10:00 0023 - Conc. 11 :08 - 11 :46 12:00 0031- Conc. 12:27 - 13:15

Wednesday 01/12/1999 14:00 0030 - ex.1 15:18 - 15:42

Thursday 02/12/1999 08:00 0030-ex.2 09:10 - 10:00 10:00 0030 - Conc. 11:09 -11:50 12:00 0026 - ex. 1 &2 12:20 - 13:50 14:00 16:00 0026-Conc. 16:26 - 17:24

Friday 03/12/1999 10:00 0039 - ex.1 11:50 -12:18 12:00 0032 - ex.1 12:20 - 12:40

0027 -ex.l 12:50 -13:21 14:00 0022 -ex.l 14:35-15:00

weekend

Monday 06/12/1999 08:00 0022-ex.2 09:06 - 09:41 0027 -ex.2 09:44 - 10:14

10:00 0032-ex.2 11 :04 - 11 :40 0022-Conc. 11:12-11:56

12:00 0027 -Conc. 12:30 -13:15 0039- ex.2 13:10 -13:35

14:00 0032 - Conc. 14:20 - 15:15 0039-Conc. 15:40 - 16:27

Tuesday 07/12/1999

Wednesday 08/12/1999

Thursday 09/12/1999

Friday 10/12/1999 10:00 0040 - ex.1 10:40 - 12:00 0036-ex.1 11 :08 - 11 :30

weekend

Monday 13/12/1999 08:00 0036-ex.2 08:10 - 08:42 0040 - ex.2 08:42 - 09:05 0036-Conc. 09:50 - 10:36

10:00 0040-Conc. 11 :14 - 12:01 12:00 14:00 16:00 0029 - ex.l 17:15 - 17:46

Tuesday 14/12/1999 08:00 0029-ex.2 08:48 - 09:08 10:00 0029 - Conc. 10:58 - 11 :45

251

Date Time interval Pile no. and Procedure Time

Wednesday 1511211999 16:00 G020 - ex.l 16:31 - 16:57 G024 -ex.l 16:59-17:21

Thursday 08:00 G020-ex.2 08:10 - 08:40 G024-ex.2 08:41 - 09:05 G020-Conc. 09:33 - 10:06

10:00 G024 - Conc. 10:40-11:17

Friday 1711211999 10:00 G038 - ex.l 11 :35 - 12:00 12:00 G034 -ex.l 12:03 - 12:25

weekend

Monday 20/1211999 10:00 G038 -ex.2 11 :20 - 11 :46 G034-ex.2 11:48 -12:10

12:00 14:00 G038 - Conc. 14:30 - 15:45 16:00 G034-Conc. 16:05 - 17:05

252

Appendix C - Calculation of flexural rigidity (El) for pile

Upper part of pile (reinforcement bar diameter = 50 mm)

diameter of bar = Area of bar = Area of bar = second moment of area of bar, I = distance to neutral axis of furthest bar, YA = distance to neutral axis of 2nd furthest bar, YB = distance to neutral axis of 2nd nearest bar, Yc = distance to neutral axis of nearest bar, Yo = distance to neutral axis of bar on NA, YE = E (concrete) = E (steel) = I pile = assuming 70 mm cover to rebar

B o A

G

o c

50 mm 1963.5 mm~

0.0019635 m~ 3.0680E-07 m

4

0.43000 m 0.16455 m 0.30406 m

0.39727 m Om

25000000 kN/m~ 205000000 kN/m~

0.059666024 m4

no of bars x (I + Area x Y~ )

IA = 7.2671E-04 m4

18 = 0.000213897 m4

IC = 0.000727328 m4

ID = 0.001240758 m4

IE = 6.1 3592E-07 m4

TOTAL =

IE =

0.002909311 m4

(Ipile x Econc) - (Isteel x Econc) + (Isteel x Esteel) 2015326.594 kNm'

1492834.514 kNm'/m run of wall

Lower part of pile (reinforcement bar diameter = 40 mm)

diameter of bar = Area of bar = Area of bar = second moment of area of bar, I = distance to neutral axis of furthest bar, YA = distance to neutral axis of 2nd furthest bar, YB = distance to neutral axis of 2nd nearest bar, Yc = distance to neutral axis of nearest bar, Yo = distance to neutral axis of bar on NA, YE = E (concrete) = E (steel) = I pile = assuming 70 mm cover to rebar

B o A

o

G C

40 mm 1256.6 mm~

0.0012566 m~ 1.2566E-07 m

4

0.43000 m 0.16455 m

0.30406 m 0.39727 m

Om 25000000 kN/m~

205000000 kN/m~ 0.059666024 m4

no of bars x (I + Area x l) IA = 4.6496E-04 m

4

18 = 0.000136611 m4

IC = 0.000465207 m4

ID = 0.000793803 m4

IE = 2.51327E-07 m4

TOTAL = 0.001860828 m4

IE = (I pile x Econc) • (lsteel x Econc) + (Isteel x Esteel) 1826599.657 kNm'

1353036.783 kNm'/m run of wall

253

References

American Concrete Institute (ACI) (1992). Causes, evaluations and repair of cracks in

concrete structures (ACI 318/318R-89) ACI, Detroit, Mich.

Atkinson, J.H., Richardson, D. and Stallebrass, S.E. (1990). Effect ofrecent on the

stiffness of overconsolidated soils. Geotechnique 40(4): 531-540.

Batten, M. and Powrie, W. (2000). Measurement and analysis of temporary prop loads at

Canary Wharf underground station, east London Proc. Inst. Civ. Engrs. Geotech. Engng

143(July): 151-163.

Batten, M., Powrie, W., Boorman, R., Yu, H.-T. and Leiper, Q. (1999). Use of vibrating

wire strain gauges to measure loads in tubular steel props supporting deep retaining walls

Proc.lnstn. Civ. Engrs, 137(Jan): 3-13.

Bolton. M.D. and Powrie, W. (1987). The collapse of diaphragm walls retaining clay.

Geotechnique 37(3): 335-353.

Bolton. M.D. and Powrie, W. (1988). Behaviour of diaphragm walls in clay prior to

collapse. Geotechnique 38(2): 167-189.

Branson, D.E. (1977). Deformation of concrete structures. McGraw-Hill, New York.

British Standard Institution (1951). Code of Practice No.2: Earth retaining structures.

London: BSI.

British Standard Institution (1986). BS 1881-206: British Testing concrete - Part 206:

Recommendations for determination of strain in concrete, Milton Keynes: BSI.

British Standard Institution (1994). BS 8002: British Standards Institution Code of

practise for earth retaining structures, Milton Keynes: BSI.

British Standards Institution (2004). Eurocode 7: Geotechnical design Part 1, General

rules. EN1997-1 :2004.

British Steel Piling Handbook (1997). British Steel PIc. i h Edition.

Britto, A.M. and Gunn, M.J. (1987). Critical state soil mechanics via finite elements. Ellis

Horwood, Chichester.

254

Burland, J.B., Potts, D.M. and Walsh, N.M. (1981). The overall stability of free and

propped embedded cantilever walls, Ground Engineering 14(5): 28-38.

Burland, J.B., Simpson, B. and St John, H.D. (1979). Movements around excavations in

London Clay Proc. i h Eur. Con! on Soil Mech. and Found. Engng, Brighton. 1: 13-29.

Carder, D.R. and Symons, I.F. (1989). Long-term performance of an embedded cantilever

retaining wall in stiff clay. Geotechnique 39(1):55-75.

Carder, D.R. and Darley, P. (1998). The long term performance of embedded retaining

walls. TRL Report 381. Transport Research Laboratory, Crowthorne.

Carswell, I., Carder, D.R. and Gent, A.J.C (1993). Behaviour during construction of a

propped contiguous bored pile wall in stiff clay at Waltham stow. TRL Project Report 10.

Project Record E468A1BG.

Chandler, R.J. (2000). Clay Sediments in Depositional Basins: the Geotechnical Cycle.

Quarterly Journal of Engineering Geology and Hydrology, 33: 7-39.

Clayton, C.R.I. and Bica, A.V.D. (1993). The design of diaphragm-type boundary total

stress cells Geotechnique 43(4): 523-535.

Clayton, C.R.I. and Milititsky, J. (1983). Installation effects and the performance of bored

piles in stiff clay. Ground Engineering, March.

Coulomb (1776). Essai ... , Paris 7: 343-382.

De Moor, E.K. (1994). An analysis of bored pile/diaphragm wall installation effects.

Geotechinque 44(2): 341-347.

Department of Transport (2000). BD 42: Design manualfor roads and bridges, Vol. 2

Section 1 Part 2: Design of embedded retaining walls and bridge abutments, Department

of Transport, London: HMSO. Crown Copyright.

Dunnicliff, J. (1993). Geotechnical instrumentation for monitoring field performance

(with the assistance ofG.E. Green) Wiley, New York.

Eurocode 2 (1992) Design of concrete structures EN1992.

255

Fjaer, E., Holt, R.M., Horsrud, P., Raaer, RM. and Risnes, R (1992). Petroleum related

Rock mechanics, Elsevier.

Fourie, A.B. and Potts, D.M. (1989). Comparison of finite element and limiting

equilibrium analyses for an embedded cantilever retaining wall. Geotechnique 39(2): 175-

188.

Gaba, A.R. (2002). Presentation given at the ICE, London.

Gaba, A., Simpson, B., Powrie, W. and Beadman, D.R (2003). Embedded retaining

walls: guidance for more economic design Construction Industry Research and

Information Association Report C580. CIRIA, London.

Garrett, C. and Barnes, S.J. (1984). The design and performance of the Dunton Green

retaining wall. Geotechnique 34(4):533-548.

Gibson, RE. and Anderson, W.F. (1961). In situ measurement of soil properties with the

pressuremeter. Civil Engineering and Public Works Review 56(658): 615-618.

Glass and Powderham (1994). Application of the observational method at the limehouse

Link. Geotechnique 44(4):665-679.

Gourvenec, S.M. and Powrie, W. (1999). Three-dimensional finite-element analysis of

diaphragm wall installation Geotechnique 49(6): 801-823.

Gunn, M.J. and Clayton, C.RI. (1992). Installation effects and their importance in the

design of earth-retaining structures. Geotechnique 42(1): 137-141.

Gunn, M.J., Satkunananthan, A. and Clayton, C.RI. (1993). Finite Element modelling of

installation effects. Retaining structures. Thomas Telford, London.

Hayward, T. (2000). Field studies, analysis and numerical modelling of retaining walls

embedded in weak rock. Ph.D. thesis, University of Southampton, UK.

Heyman, G. and Clayton C.RI., (1999) Block sampling of soil: Some practical

considerations Geotechnicsfor developing Africa Wardle, Blight and Fourie, Balkema,

Rotterdam.

256

Hight, D.W. and Leroueil, S. (2003). Characterisation of soils for engineering purposes.

Characterisation and Engineering Properties of Natural Soils. Tan et al. (eds) A A

Balkema ISBN 90 C5809 537 1: 255-360.

Hitchcock, A.R. (2003). Elimination of temporary propping using the observational

method on the Heathrow airside road tunnel project Ground Engineering, May.

HMSO (1978) British Regional Geology - The Wealden District.

Humpage, A.J. and Booth K.A. (2000). The Lower Greensand in the Mid-Kent and

Ashford Contract Sectors of the Channel Tunnel Rail Link Construction Works. British

Geological Survey Commercial Report CRlOO/19.

Jardine, R.J., Symes, M.J. and Burland, J.B. (1984). The measurement of soil stiffness in

the triaxial apparatus. Geotechnique 34(3): 323-340.

Jardine, RJ., Potts, D.M., Fourie, A.B. and Burland, J.B. (1986). Studies of the influence

of non-linear stress-strain characteristics in soil-structure interaction, Geotechnique 36(3):

377-396.

Loveridge, F. (2000). Evaluation of prop loads at CTRL contract 430 Ashford Tunnels.

Ground Engineering

Mair, R.J. (1993). Developments in geotechnical engineering research, application to

tunnels and deep excavations. Proceedings of the Institution of Civil Engineers Civil

Engineering, 97(1): 27-41.

Mair, R.J. and Wood, D.M. (1987). In-situ Pressuremeter Testing: Methods, Testing and

Interpretation. Construction Industry Research and Information Association,

Butterworths, London.

Mantell, G.A. (1833). Geology of south-east England, Longman, London.

Massarsch, K.R. (1975). New method for measurement of lateral earth pressure in

cohesive soils. Can. Geotech. J. 12:142-146.

Massarsch, K.R., Holtz, R.D., Holm, B.G. and Fredriksson, A. (1975). Measurement of

horizontal in situ stresses. Proc. of Speciality Conf on in situ measurement of soil

properties, North Carolina, 1:266-286. American Society of Civil Engineers.

257

Mikkelsen, P.E. (2003). Advances in inclinometer data analysis. Symposium on Field

Measurements in Geomechanics, FMGM, Oslo Norway. September.

Mosley, W.H. and Bungey, J.H. (1976). Reinforced Concrete Design London, Macmillan.

Neville, A.M. (1970). Creep of concrete: plain, reinforced, and prestressed. North

Holland publishing Company, Amsterdam.

Ng, C.W.W. (1992). An Evaluation of Soil-structure Interaction Associated with a Multi­

propped Excavation. PhD thesis, University of Bristol.

Ng, C.W.W., Lings, M.L., Simpson, B. and Nash, D.F.T. (1995). An approximate analysis

of the three-dimensional effects of diaphragm wall installation. Geotechnique 45(3):497-

507.

Ng, C.W.W. and Yan, R.W.M. (1999). Three-dimensional modelling ofa diaphragm wall

construction sequence. Geotechnique, 49(6): 825-834.

Ooi, P.S.K. and Ramsey, T.L. (2003). Curvature and bending moments from inclinometer

data, ASCE International Journal ofGeomechanics, 3(1):64-74.

Padfield, C.J. and Mair, RJ. (1984). Design of retaining walls embedded in stiff clays.

Report 104. London: Construction Industry Research and Information Association.

Peattie, K.R. and Sparrow, R.W. (1954). The fundamental action of earth pressure cells

Journal of the mechanics and physics of solids, 2: 141-155.

Penman, A.D.M. and Charles, J.A. (1981). Assessing the risk of hydraulic fracture in dam

cores. PRc. loth Int. Con! Soil Mech., Stockholm 1: 457-462.

Potts, D.M. and Bond, A.J. (1994). Calculation of Structural forces for propped retaining

walls. XIII ICSMFE, New Delhi, India.

Potts, D.M. and Fourie, A.B. (1984). The behaviour of a propped retaining wall: results of

a numerical experiment. Geotechnique, 34(3): 383-404.

Potts, D .M. and F ourie, A.B. (1985). The effect of wall stiffness on the behaviour of a

propped retaining wall. Geotechnique 35(3): 347-352.

Potts D. M. and Zdravkovic, L. (1999). Finite element analysis in geotechnical

engineering. Thomas Telford, London.

258

Powrie, W. (1985). Discussion on Perfonnance in propped and cantilever rigid walls.

Geotechnique 35(4): 546-548.

Powrie, W. (2004). Soil mechanics: concepts and applications (2nd edition). London: Spon

Press.

Powrie, W. and Kantartzi, C. (1996). Ground response during diaphragm wall installation

in clay: centrifuge model tests. Geotechnique 46(4): 725-739.

Powrie, W. and Li, E.S. (1991). Finite element analyses of an in situ wall propped at

fonnation level. Geotechnique 41(4): 499-514.

Powrie, W, Pantelidou H. and Stallebrass, S.E. (1998). Soil stiffness in stress paths

relevant to diaphragm walls in clay. Geotechnique 48(4): 483-494.

Raphael, J.M. (1984) Tensile strength of concrete ACI Journal, 81(2): 158-165.

Rankine, W. J. M. (1857). On the stability ofloose earth, Phil. Trans. Royal Society

147(2): 9-27.

Richards, D.J. Chandler, R.J. and Lock, A.C. (2003) Electronic data transfer systems for

field monitoring. Proc. Inst. Civ. Engrs 156(1): 47-55.

Richards, D.J., Clark, J., Powrie, W. and Heyman, G. (2005) An evaluation of total

horizontal stress measurements using push-in pressure cells in an overconsolidated clay

deposit. Proc. Inst. Civ. Engrs. Geotech. Engng accepted for publication.

Richards, D.J., Holmes, G. and Beadman, D.R. (1999). Measurement of temporary prop

loads at Mayfair car park. Proc. Inst. Civ. Engrs. Geotech. Engng 137(July): 165-174.

Richards, D.J. and Powrie, W. (1994). Finite element analysis of construction sequences

for propped retaining walls. Proc. Inst. Civ. Engrs, Geotech. Engng. 107(1994): 207-216.

Richards, D.J. and Powrie, W. (1998). Centrifuge tests on doubly propped embedded

retaining walls in overconsolidated kaolin clay, 48(6): 833-846.

Roberts, M. (2003). Private communication.

Roberts, T.O.L., Roscoe, H., Powrie, W. and Butcher, D. (2005). private communication.

Roscoe, H. (2005). Private communication.

259

Roscoe, H. and Twine, D. (2001). Design collaboration speeds Ashford tunnels. World

Tunnelling, 14(5): 237-241.

Ross, A.D. (1937). Concrete Creep Data. The Structural Engineer 15: 314-32.

Rowe, P.W. (1952). Anchored sheet-pile walls. Proc. Inst. Civ. Engrs. 1(1): 27-70.

Rowe, P.W. (1955). A theoretical and experimental analysis of sheet piled walls. Proc.

Inst. Civ. Engrs. 4(1): 32-69.

Rowe, P. W. (1956). Sheet piled walls at failure. Proc. Inst. Civ Engrs. 276-315.

Ryley, M.D. and Carder D.R. (1995). The performance of push-in spade cells installed in

stiff clay. Geotechnique 45(3): 533-539.

Saxena, S.K. (1975). Measured performance of a rigid concrete wall at the World Trade

Center, Proc. Conf. on Diaphragm Walls and Anchorages pp. 107-112.

Skempton, A.W. and Weeks, A.G. (1976) The Quaternary history of the Lower Greensand

escarpment and Weald Clay vale near Sevenoaks, Kent. Phil. Trans. R. Soc. Lond. A. 283:

493-526.

Simpson, B. (1992). Thirty-second Rankine lecture: Retaining structures: displacement

and design. Geotechnique 42(4): 541-576.

Simpson, B. and Powrie, W. (2000). Embedded retaining walls: theory, practise and

understanding. ISSMGE XVth International conference in Turkey.

Smart J.G.O., Bisson G. and Worssam B.C. (1966). Geology of the Country around

Canterbury and Folkestone. Her Majesty's Stationery Office, London.

Smethurst, J.A. 2003. The use of discrete piles for infrastructure slope stabilisation. PhD

thesis, University of Southampton, UK.

Stroud, M.A. (1974). The standard penetration test in insensitive clays and soft rocks

Proc. Eur. Symp. on Penetration Testing (ESOPTI) Volume 2:2, pp. 367-375, Swedish

Geological Society, Stockholm.

Stroud, M.A. and Sweeney, D.J., (1977). Discussion appendix in: A review of diaphragm

walls, pp. 142-148. London: Institution of Civil Engineers.

260

Symons, I.F. (1983). Assessing the stability of a propped, in situ retaining wall in

overconsolidated clay Proc. Inst. Civ. Engrs, 75(2): 617-633.

Symons, I.F. and Carder, D.R. (1990). Long term behaviour of embedded retaining walls

in over-consolidated clay. Geotechnical Instrumentation in Practice, TTL, London.

Symons, I.F. and Carder, D.R. (1992). Field measurements on embedded retaining walls.

Geotechnique 42(1): 117-126.

Symons, I.F. and Carder, D.R. (1993). Stress changes in stiff clay caused by the

installation of embedded retaining walls. Retaining structures Thomas Telford, London.

Symons, I.F. and Tedd, P. (1989). Behaviour ofa propped embedded retaining wall at Bell

Common Tunnel in the longer term. Geotechnique 34(4): 701-710.

Tavenas, F.A., Blanchette, G., Leroueil, S., Roy, M. and La Rochelle, P. (1975).

Difficulties in the in situ determination ofKo in soft sensitive clays Proc. of speciality

conference on in situ measurement of soil properties, North Carolina, 1:450-476. New

York, American Society of Civil Engineers.

Tedd, P., Chard, B.M., Charles, l.A. and Symons, I.F. (1984). Behaviour ofa propped

embedded retaining wall in stiff clay at Bell Common. Geotechnique 34(4): 513-532.

Tedd P. and Charles l.A. (1981). In-situ measurement of horizontal stress in

overconsolidated clay using push-in spade-shaped pressure cells. Geotechnique 31(4): 534

-538.

Tedd P. and Charles l.A. (1983). Evaluation of push-in pressure cell results in stiff clay.

Proceedings of the International Symposium on soil and rock investigation by in-situ

testing. Paris, Vol. 2, 579-584.

Tedd P., Powell lA., Charles l.A. and U glow I.M. (1990). In-situ measurement of earth

pressures using push-in spade-shaped pressure cells - 10 years experience. Geotechnical

Instrumentation in practice, Thomas Telford, London.

Terzaghi, K (1934). Theoretical Soil Mechanics, Wiley and Sons.

Terzaghi, K (1943). Large retaining wall tests, Engineering News Record 112: 36-140.

Terzaghi, K (1954). Anchored Bulk Heads, Trans. Am. Soc. Civ. Engrs 119: 1243-1280.

261

Tschebotarioff, G. (1951). Soil mechanics, foundations and earth structures, McGraw­

Hill.

Union Railways Ltd (1997 A). Central Ashford - Geotechnical Design Basis Report

chainage 87+842 to 92+042. Report number 440-RUP-LCEEH-00043-AA.

Union Railways Ltd (1997 B). Ashford Tunnel Design Criteria. Report number 430-RUG­

RLEEZ-00120-AA.

Weiler, W.A. and Kulhawy, F.H. (1982). Factors affecting stress cell measurements in

soil. J Geotechnical Engineering ASCE, 108, No. GT12 (December): 1529-1548.

Wood, L.A. and Perrin, A.J. (1984). Observations of a strutted diaphragm wall in London

clay: a preliminary assessment. Geotechnique, 34(4): 563-579.

Xu (2005). The behaviour of soil behind full-height integral abutments. PhD thesis,

University of Southampton.

262