Angel ASHIKOV1

Charles CLIFTON2

Borislav BELEV3

BEHAVIOUR AND DESIGN OF EBFs FOR

SEISMIC ACTIONS. THE NEW ZEALAND

EXPERIENCE

Abstract: A steel eccentrically braced frame (EBF) is a relatively new structural system for providing seismic resistance of buildings which employs the structural fuse concept in a ductile building design (i.e. which is designed to undergo controlled damage in a severe earthquake). The conventional EBFs are expected to sustain significant damage during a design level earthquake through repeated inelastic deformations of their active links. In these EBFs, the active link is traditionally made continuous with the collector beam and supports the floor slab, although it is not made composite with the floor slab. This continuity of active link with the adjacent collector beam or beams means that, following yielding of the active links, a repair is expected to be costly and disruptive, even if the structure has met its goal of providing life safety during an earthquake. The first detailed New Zealand design procedures for EBFs were published in 1986 by the Heavy Engineering Research Association (HERA) [6]. Following the 2010/2011 Christchurch earthquake series, in which most EBF structures were pushed into the inelastic range, necessitating replacement of active links in some buildings, the design procedures have been adapted to accommodate the replaceable link concept for EBF. This concept will allow for rapid inspection and replacement of yielded and damaged links following a major earthquake, thereby permitting the structure to be economically brought back to its original safety level. A new HERA design guide [9] has been issued to incorporate these changes in design practice. An overview of the performance of buildings with EBFs in these earthquakes is also included in this paper.

Keywords: EBF, link, seismic resistance, damage, energy dissipation, ductility _____________________________________

1 Angel Ashikov, PhD student at Department of Steel, Timber and Plastic Structures, University of Architecture, Civil Engineering and Geodesy, Bulgaria, e-mail: angel.ashikov@gmail.com

2 Dr. Charles Clifton, Associate Professor at Department of Civil and Environmental Engineering, The University of Auckland, New Zealand, e-mail: c.clifton@auckland.ac.nz

3 Dr. Borislav Belev, Professor at Department of Steel, Timber and Plastic Structures, University of Architecture, Civil Engineering and Geodesy, Sofia, Bulgaria: e-mail: belev_fce@uacg.bg

1. Introduction

A steel eccentrically braced frame (EBF) is a relatively new structural system for providing seismic resistance of buildings designed for controlled damage which employs the structural fuse concept (Fig. 1). Seismic resistant eccentrically braced frames (EBFs) are a lateral load-resisting system for steel building that combine high stiffness in the elastic range, with good ductility and energy dissipation in the inelastic range. EBFs can be considered as a hybrid between moment resisting frames (MRFs) and concentrically braced frames (CBFs). The bracing members and configuration in the EBFs provide the high elastic stiffness characteristic of CBFs, permitting code drift requirements to be easily met economically. The active link is the ductile fuse, which, when well designed and detailed, provides global ductility and energy dissipation capacity comparable to those of MRFs.

Figure 1. Examples of eccentrically braced frames (source: [1]) a - link; b - collector beam; c - diagonal brace; d - column

The distinguishing characteristic of an EBF is that at least one end of every brace is connected so that the brace force is transmitted either into another brace or into a column through shear and bending in a beam segment called an “active link”. Under severe cyclic loading, the inelastic deformation is restricted and localized primarily in the links, which are designed and detailed to sustain large inelastic deformations without loss of strength.

In the design of a seismic-resistant EBF it is necessary to determine the plastic rotation demand on the link. This is most easily accomplished through the use of energy dissipation mechanisms (also commonly called collapse mechanisms), constructed by assuming rigid plastic behaviour of the members. Plastic mechanisms of two types of EBF are illustrated in Fig. 2.

Figure 2. Link rotation angle and frame interstorey drift (source: [1])

The following relationships can be derived from Fig. 2:

(1)

(2)

p

p

p p

h

L

e

θ

γ θ

∆=

=

where: ∆p – plastic interstorey drift; θp – plastic storey drift angle;

γp – link plastic rotation angle; h – storey height; L – bay width

The links act as ductile fuses, dissipating energy through stable hysteretic behaviour, while limiting the forces transmitted to the other components. By applying the capacity design procedures the designer can force the yielding to occur in the ductile link elements, while preventing yielding and failure of other elements. In this way the braces are designed not to buckle, regardless of the severity of the lateral loading on the frame, making the performance insensitive to the level of severe earthquake loading. Capacity design procedures are an integral part of ultimate limit state (ULS) seismic considerations for a structure which is intended to undergo controlled damage at the ULS. This approach is used in most multi-storey normal importance buildings in high seismic regions. Capacity design is a process in which it is pre-selected which members or components of the seismic resisting system are permitted to yield and which are to remain elastic. In capacity design of seismic resisting systems, the principal energy dissipating mechanisms are chosen and are suitably proportioned and detailed. All other elements of the seismic-resisting system are provided with sufficient reserve strength to ensure that the chosen energy dissipating mechanisms are maintained throughout the deformations that may occur.

2. Performance of building with EBFs in recent strong

earthquakes in New Zealand

2.1. Earthquakes series of 2010/2011 in Christchurch, New Zealand.

In the period from September 4th, 2010 to July 13th, 2011, the Christchurch city and surrounding areas have been shaken by six powerful and destructive earthquakes. The series commenced on September 4th, 2010, with an earthquake of Magnitude 7.1 on the Richter scale, centred some 40 km from the Christchurch Central Business District (CBD) and at a focal depth of around 20 km. After that earthquake there have been over 18 aftershocks with magnitude above 5.0 on the Richter scale, including that of February 22nd, 2011 which has been the largest aftershock. On February 22nd, 2011, an M 6.3 earthquake shook Christchurch, New Zealand. The aftershock’s epicentre was 5km from the city and a focal depth of 5km. The very close and shallow proximity to Christchurch city, coupled with the very hard basement rock through which the fault ruptured, resulted in very high Peak Ground Accelerations. This event caused massive damage to the city, collapsing hundreds of buildings in the CBD. Also, while the September 4th earthquake had struck in the night, the February 22nd earthquake struck at 12:51 p.m. local time when people had filled the offices and cafes of the CBD, leading to 184 confirmed deaths. Widespread liquefaction in the CBD and eastern suburbs caused foundation movement in housing and office buildings alike and led to the destruction of thousands of houses and low rise commercial buildings. Two medium rise reinforced concrete office buildings and one parking garage collapsed along with hundreds of unreinforced masonry buildings, including most number of heritage structures (see Fig. 3 and Fig. 4). Many other buildings in the CBD have been severely damaged and some required demolition, which necessitated carefully controlled access to the CBD in the weeks following the earthquake. The total losses are estimated to be over NZ $40 billion.

Figure 3. URM building on the left and The PGC building on the right (source: Elwood and [11])

Figure 4. Examples of damage in reinforced concrete buildings; (a) column shear failure, (b) precast wall damage, (c) beam-column joint, (d) wall damage (source: Sritharan)

Peak ground accelerations over 1,5g in Heathcote Valley Primary School, 1 km from the epicenter and between 0.5 and 0.8g in the CBD were reported by strong ground motion recording stations. Unusually high vertical ground motions, sometimes exceeding the horizontal component were recorded. Ground motions recorded in the CBD generally exceeded the 500-year and even the 2500-year elastic design spectrum of the New Zealand seismic design standard (NZS1170.5 2004). Fig. 5 compares the elastic design spectra with the 5% damped response spectra for the horizontal components of the September 4th, 2010 and February 22nd, 2011 events recorded at the Christchurch Hospital.

Figure 5. Horizontal spectral acceleration for Christchurch Hospital (8km distance of epicenter) form September 4, 2010 and February 22, 2011 events compared with NZS 1170.5 elastic design spectra for Christchurch (source: Elwood, ground motion data from GeoNet).

2.2. Multi-storey steel structures in the Christchurch area

Historically, multi-storey buildings in Christchurch have been constructed from reinforced concrete, however from 2000 onwards structural steel has become more predominant. This means that most of the buildings with steel structure in Christchurch have been designed to the latest seismic provisions in New Zealand. Table 1 provides a listing of the multi-storey steel framed buildings in the CBD and some in the suburbs.

Table 1. Multi-storey steel buildings with EBF system (source: [11])

Seismic-resisting system

Floor system Number

of storeys

Year completed

EBFs and MRFs Composite deck and steel beams 22 2010

EBFs and MRFs Composite deck and steel beams 12 2009

EBFs Composite deck and steel beams 5 2008

EBFs Precast columns and hollow-

core units with topping 4 2003

EBFs Composite deck and steel beams 5 2010

2.3. Seismic performance of multi-storey EBFs in Christchurch

The tallest building in Christchurch, the 22-storey Pacific Residential Tower, consists of perimeter EBFs up to sixth floor, shifting to EBFs around the lift core above that level, with a transfer slab designed to horizontally distribute the seismic loads at that transition point. The design of the EBFs in the building has been governed by the need to limit drift, with a resulting design ductility factor equal to 1.5. This is typical of EBFs in tall buildings in New Zealand’s moderate to low seismic zones. Paint flaking and residual link shear deformations have been observed in the EBF links at that level. The bottom 6 levels of this structure comprise a vertical car stacking system, with very open plan structure and some regions of floor slab omitted. This is followed by floors of hotel rooms, with many internal, non-structural fire and acoustic rated walls, then floors of apartments with a lesser length of such walls. The effects of this variation in non-structural stiffness up the structure modified the earthquake response, concentrating inelastic demand into the active links of the car parking floors. After detailed investigation of these yielded links it was determined that 42 of them would need to be replaced; this was achieved by cutting them out, welding endplates onto the cut ends of the collector beam/brace region, fabricating active links with endplates and bolting these back into the system to complete the replacement. Altogether 42 active links were replaced in this way; details are in [13].

а) Global view b) Flaked paint on EBF link

Figure 6. Pacific Residential Tower in Christchurch (source: [11])

The Club Tower Building (Fig. 7(a)) has eccentrically braced frames located on three sides of a lift core eccentrically located closer to the west side of the building, and a ductile moment resisting frame (DMRF) along the east facade. EBFs designed in compliance with the NZS 3404 provisions are typically sized considering a ductility factor of up to 4, a level of link deformations that would correspond to significant shear distortions of the links. Given the magnitude of the earthquake excitations, with demands above the ULS design level, substantial yielding of the EBF links was expected and observed; however with peak plastic shear strains of 5% as determined from hardness measurements and correlation with plastic shear strain. An example of the visible extent of deformation is shown in (Fig. 7(b)). . The links are free of visible residual distortions or any cracking; subsequent detailed evaluations have shown that they can be left in place with sufficient post-earthquake strength and ductility to resist another design level event. Hair line cracking of non-structural gypsum plaster board finishes has been observed elsewhere throughout the building. The ductile MRF along the east facade did not show any evidence of yielding. Its design had been governed by the interstorey drift limitations, particularly under torsional response due to the eccentricity of the core, and its corresponding effective ductility factor has been as low as 1,25. Following repair of non-structural wall cracking and lift guide rail realignment, the building has been returned to full service in June 2011 and was the first medium-rise building in the city to be returned to service.

а) Global view b) Paint flaking of partially hidden EBF link Figure 7. Club Tower in Christchurch (source: [11])

2.4. Seismic performance of EBFs in parking garages, Christchurch

The EBFs in a hospital parking garage which is very close to the epicenter have performed well, although some link fractures have been observed in two brace bays (Fig. 8 and Fig. 9). This parking structure has been designed to accommodate three additional floors. Yet, some of the links at the first story have shown paint flaking as evidence of inelastic deformations (Fig. 8(b) and Fig. 9(b)). The fractures as shown in close-up in Fig. 9(a) have been of particular concern as these were the first fractures recorded in EBFs worldwide. The most likely explanation lies in the offset of the brace flange from the stiffener. This offset implies that the axial tensile force in the brace fed into the active link/collector beam panel zone through a flexible beam flange rather than directly into the stiffener. No lateral movement or twisting of the ends of the active links (Fig. 9(b)) have been observed, indicating that the lateral resistant provisions are adequate despite that such restraint was only applied to the top flange and the EBFs were not integral with slab above.

а) EBFs b) Evidence of EBF link yielding Figure 8. EBFs in a parking garage of Antigua St., Christchurch (source: [11])

а) Fractured link at lower level EBF b) Evidence of inelastic deformations of link Figure 9. Damaged seismic links in parking garage of Antigua St., Christchurch (source: [11])

3. Historical development of design provisions and guidelines

for EBFs

The first detailed New Zealand design procedures for eccentrically braced frames (EBFs) were published in 1986 by the Heavy Engineering Research Association as a set of notes for a seminar series on the design of seismic-resistant multi-storey steel-framed buildings [6]. A research project commenced at the University of Canterbury to study the inelastic behaviour of steel-framed EBFs and to develop an appropriate design philosophy and procedures for this structural form has been successfully completed in 1990 by MacRae [7]. That has led to the design procedures for EBFs presented in HERA Report R4-76 entitled “Seismic Design Procedures for Steel Structures” [8]. These design procedures covered the full range of expected inelastic demand, from fully ductile response to elastic response for EBFs, moment resisting frames (MRFs) and concentrically braced frames (CBFs).

The last HERA design guide for EBFs was published in 2013 [9]. This guide incorporates changes in design practice that have occurred since the publication of R4-76. The design procedures have been adapted to accommodate the replaceable link concept. This concept was developed following the performance of EBFs in the Christchurch earthquakes series 2010 and 2011. It allows for rapid inspection replacement of yielded and damaged links following a major earthquake, thereby permitting the structure to be economically brought back to its original safety level.

In AISC Seismic provisions 2010 [1], Eurocode 8 [2] and NZS3404 [3], the active links are classified into 3 categories according to the type of plastic mechanism developed:

-short links, which dissipate energy by yielding essentially in shear;

, ,1,6 / (3)s p link p linke e M V< =

-long links, which dissipate energy by yielding essentially in bending;

, ,3,0 / (4)L p link p linke e M V> =

-intermediate links, where plastic mechanism involves bending and shear.

(5)s Le e e< <

A classification in three groups is made in terms of link rotation angle θp between the link and the element outside of the link. The classification’s purpose is to limit local rotation consistent with global deformation.

-short links

0,08 (6)p pR radθ θ≤ =

-long links

0,02 (7)p pR radθ θ≤ =

-intermediate links

(8)p pR linear interpolationbetweentheabovevaluesθ θ≤ =

In the majority of the buildings in New Zealand with EBFs the active links are chosen as short links as they do not generate a large bending moment in the collector beam, permitting the frame to be economically designed. Structural displacement ductility demands on the 4 categories of seismic-resistant systems for the ultimate limit state are specified in Table 2.

Table 2. Relationship between category of structure and structural displacement ductility demand for the ultimate limit state (source: [3])

Category Description Displacement ductility demand

1 Fully ductile m > 3.0

2 Limited ductile 3.0 > m > 1.25

3 Nominally ductile m = 1.25

4 Elastic m = 1.0

Based on observations of the performance of steel buildings in the 22 February 2011 Christchurch earthquake [8], it is suggested to limit the ductility

demand to m=3 when minimization of post-earthquake repair is an important

design criterion. An alternative approach to reduce repair costs and disruption is to use removable links for EBFs and this has now become standard practice in New Zealand.

4. New concepts and research findings

4.1. Bolted replaceable active links

Conventional steel eccentrically braced frames (EBFs) are expected to sustain significant damage during a design level earthquake through repeated inelastic deformation of the active link. In these EBFs, the yielding link is traditionally continuous with the collector beam and supports the floor slab, although it is not made composite with the floor slab. Repair is otherwise expected to be costly and disruptive, even if the structure has met its goal of providing life safety during an earthquake. These drawbacks can be mitigated by designing EBFs with replaceable active links. The replaceable active link concept allows for quick inspection and replacement of damaged links following a major earthquake, significantly minimising time to reoccupy the building. A bolted replaceable active link allows the link element to be fabricated from a lower steel grade, or modified section dimensions, thereby assuring an elastic response of the collector beam outside the removable link element. With the replaceable link the designer has greater flexibility to choose a section that best meets the required strength, without automatically changing the floor collector beam section. This concept was studied by Mansour [15]. HERA have also undertaken finite element analysis to verify the design procedures for EBFs with replaceable links. The results showed that the proposed design procedure achieves the objectives of suppressing inelastic demand away from the active link. Furthermore, built-up sections with relatively thin webs and thick flanges can be used to optimize the design EBFs and, if deemed beneficial, links made of different grades of steel can also be used. This new type of a replaceable link with built-up cross-section, connection configurations, welding details and intermediate stiffener spacing were studied by the first author under the guidance of Professor Charles Clifton. The active links exhibited a very good ductile behaviour, developing repeatable and stable yielding.

4.2. Research activities on EBFs performed at University of Auckland

The second author is currently involved in PhD supervision either at first or second supervisor level with students who are looking at the following aspects of EBF system performance in severe earthquakes:

a. Slab participation to the strength and stiffness including self-centring capability. This work has been completed [14]. When the out-of-plane stiffness of the composite floor slabs is taken into account, the building exhibits self-centring by showing a tendency to go back to its original position. b. Enhanced performance of EBF active links - improving the performance of bolted replaceable active links through changes to stiffener and

end-plate details. It is a joint research effort between University of Auckland and Portland State University, USA. c. Studies on soil-structure-foundation interaction and its influence on EBFs on pad foundations. d. Determining a field-based method for establishing the post-earthquake residual capacity of yielded EBF active links. This PhD project is nearing completion.

5. Conclusions

Steel structures with EBFs generally performed very well during the Christchurch earthquake series of 2010 and 2011. However, poor design and/or detailing resulted in a few fractures at link zones. This shows the importance of good detailing, load path development and robust connections [9].

The replaceable active link concept is very promising because it allows the structural fuses to be designed with desired cross-section dimensions and different steel grade, if needed. The damaged active links can be easily replaced following a major earthquake without costly and time consuming repair works.

Acknowledgement

The authors gratefully acknowledge the funding provided by AUSMIP+ program of European Union for the PhD mobility of the first author to the University of Auckland.

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