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DEVELOPMENT OF AN EARTH PRESSURE MODEL FOR DESIGN OF RETAINING STRUCTURES IN PIEDMONT SOILS
FINAL REPORT
University of North Carolina at Charlotte Department of Civil and Environmental Engineering
October, 2008
J. Brian Anderson, Ph.D. P.E., Principal Investigator Vincent Ogunro, Ph.D., Co-Principal Investigator
Jonathan M. Detwiler, Graduate Student James R. Starnes Jr., Graduate Student
Richard E. Burrage Jr., Graduate Student
TECHNICAL REPORT DOCUMENTATION PAGE
1. Report No. FHWA/NC/2006-51
2. Government Accession No.
3. Recipient’s Catalog No.
4. Title and Subtitle Development of an Earth Pressure Model for Design of Earth Retaining Structures in Piedmont Soil
5. Report Date October 2008
6. Performing Organization Code
7. Author(s) J. B. Anderson and V. O. Ogunro
8. Performing Organization Report No.
9. Performing Organization Name and Address University of North Carolina at Charlotte College of Engineering
10. Work Unit No. (TRAIS)
9201 University City Boulevard Charlotte, NC 28223-0001
11. Contract or Grant No.
12. Sponsoring Agency Name and Address North Carolina Department of Transportation Research and Development Unit
13. Type of Report and Period Covered Final Report
1549 Mail Service Center Raleigh, 27699-1549
July 2004 – June 2007
14. Sponsoring Agency Code 2005-16
Supplementary Notes:
16. Abstract Anecdotal evidence suggests that earth pressure in Piedmont residual soils is typically over estimated. Such estimates of earth pressure impact the design of earth retaining structures used on highway projects. Thus, the development of an appropriate model for estimating earth pressure would result in more rational design of retaining structures in Piedmont residual soils. Accordingly, the objective of this research was to develop an earth pressure model for Piedmont residual soil.
An experimental program to estimate, model, and measure earth pressure in Piedmont residual soils was carried out by the University of North Carolina at Charlotte. This study centered around the instrumentation, construction, and load testing of four sheet pile retaining walls at two sites in the Charlotte Belt and Carolina Slate belt regions of the Piedmont. The scope of work included extensive insitu and laboratory soil testing to estimate soil strength parameters for the residual soils; and numerical models to plan the load testing program and evaluate the results. Results of the load tests showed little or no earth pressure due to Piedmont residual soil. Interpretation of data from the sites using theoretical and numerical methods supports this findings.
Conclusions from this study include: 1) The earth pressure currently used in design of retaining structures in Piedmont soils is greater than earth pressure measured during load tests. Field measurements from the instrumented wall load tests demonstrated that the retained soils exerted little or no pressure on the structure. 2) The Piedmont soils that were tested in this research had significant strength. The average drained friction angle was 28o and the average drained cohesion intercept was above 300 psf. These values were consistent with those found in the literature for similar soils. 3) Based on the soil test results as well as the minimal earth pressure detected during the load tests, the soil strength parameters, φ’and c’ should be used together in Rankine’s earth pressure equation to predict the earth pressure in Piedmont soils. And 4) Triaxial tests provided the most consistent measurement of φ’ and c’. The borehole shear tests also measure φ and c’ but should only be used when triaxial testing is unavailable.
17. Key Words Retaining structure, residual soil, sheet pile, insitu test, load test, instrumentation, dilatometer, borehole shear, triaxial
18. Distribution Statement
19. Security Classif. (of this report) Unclassified
20. Security Classif. (of this page) Unclassified
21. No. of Pages 147
22. Price
Form DOT F 1700.7 (8-72) Reproduction of completed page authorized
ii
DISCLAIMER
The contents of this report reflect the views of the Authors and not necessarily the views of the University. The Authors are responsible for the facts and the accuracy of the data presented herein. The contents do not necessarily reflect the official views or policies of the North Carolina Department of Transportation or the Federal Highway Administration. This report does not constitute a standard, specification, or regulation.
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ACKNOWLEDGMENTS
This project would not have been possible without major contributions from
several individuals and groups. The researchers would like to express their sincerest
gratitude to following individuals and organizations:
The North Carolina Department of Transportation Geotechnical Engineering Unit
Njoroge Wainaina, Mohammed Mulla, Chris Kreider, Scott Hidden, Chris Chen, Scott Webb, Clint Little, Jay Stickney, Chad Smith, and Mike Smith.
Materials and Tests Unit C.K. Su and Randy Privette Division 12 Construction and Operations
Max Buchanan, Dan Grissom, and Marty Plyler. Division 10 Construction and Operations
Barry Moose, Garland Haywood, Gary Huneycutt, and Mark Smith.
S&ME, Inc.
Billy Camp and Tim Cleary Insitu Soil LLC.
Roger Failmezger
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SUMMARY
Anecdotal evidence suggests that earth pressure in piedmont residual soils is
typically over estimated. Increased or overly conservative estimates of earth pressure
impact the design of earth retaining structures used on highway projects. Thus, the
development of an appropriate model for estimating earth pressure would result in more
rational design of retaining structures in Piedmont residual soils. Accordingly, the
objective of this research was to develop an earth pressure model for Piedmont residual
soil.
The University of North Carolina at Charlotte carried out an experimental
program to estimate, model, and measure earth pressure in Piedmont residual soils. This
study centered around the instrumentation, construction, and load testing of four sheet
pile retaining walls at two sites in the Charlotte Belt and Carolina Slate belt regions of the
Piedmont geologic province. The scope of work included extensive insitu and laboratory
soil testing to determine soil strength parameters for the residual soils. Numerical models
were used to both plan the load testing program and evaluate the results. Results of the
load tests showed little or no earth pressure due to Piedmont residual soil. Data
interpretation through theoretical and numerical methods supports this findings.
Conclusions from this study include:
1) The earth pressure currently used in design of retaining structures in Piedmont
soils is significantly greater than earth pressure measured during load tests. Field
measurements from the instrumented wall load tests demonstrated that the
retained soils exerted little or no pressure on the structure.
2) The Piedmont soils that were tested in this research had significant strength.
The average effective friction angle was 28o and the average drained cohesion
v
vi
intercept was above 300 psf. These values were consistent with those found in
the literature for similar soils.
3) Based on the soil test results as well as the minimal earth pressure detected
during the load tests, the soil strength parameters φ’ and c’ should be used together
in Rankine’s earth pressure equation to predict the earth pressure in Piedmont
soils.
4) Triaxial tests provided the most consistent measurement of φ’ and c’. The
borehole shear tests also measure φ and c’ but should only be used when triaxial
testing is unavailable.
TABLE OF CONTENTS page TECHNICAL REPORT DOCUMENTATION PAGE...................................................... ii DISCLAIMER ................................................................................................................... iii ACKNOWLEDGMENTS ................................................................................................. iv SUMMARY.........................................................................................................................v 1 INTRODUCTION ............................................................................................................1
1.1 Introduction........................................................................................................1 1.2 Background........................................................................................................4
1.2.1 Retaining Structures............................................................................4 1.2.2 Methods for Estimating Lateral Earth Pressure ..................................4
1.2.2.1 Theoretical Approaches .......................................................4 1.2.2.2 Soil Structure Interaction Approach ..................................10 1.2.2.3 Finite Element Method ......................................................10
1.3 Retaining Structure Design ..............................................................................12 1.4 NCDOT Retaining Structures ..........................................................................12
2 PROJECT OVERVIEW .................................................................................................14
2.1 Objectives ........................................................................................................14 2.2 Research Methodology and Tasks ...................................................................14 2.3 Significance of Work .......................................................................................18 2.4 Limitations of Research ...................................................................................19
3 REVIEW OF RELEVANT LITERATURE ...................................................................20
3.1 Residual Soil ....................................................................................................20 3.1.1 Comprehensive test site studies ........................................................20 3.1.2 Other works involving insitu and laboratory testing.........................22 3.1.3 Specific Literature on Triaxial Testing in Residual Soils ................24
3.2 Load Testing of Retaining Structures ..............................................................24 4 SITE LOCATION AND CHARACTERIZATION........................................................26
4.1 Site Selection and Investigation Plan...............................................................26 4.2 Statesville.........................................................................................................27
4.2.1 Insitu Tests ........................................................................................27 4.2.2 Laboratory Soil Tests........................................................................32
4.3 Monroe.............................................................................................................36 4.3.1 Insitu Tests ........................................................................................36 4.3.2 Laboratory Soil Tests........................................................................40
4.4 Summary of Soil Testing .................................................................................40
vii
5 INSTRUMENTED WALL LOAD TESTS....................................................................43 5.1 Preliminary Modeling ......................................................................................43
5.1.1 FEM Initial Conditions .....................................................................43 5.1.2 FEM Modeling..................................................................................44
5.2 Selected Wall Section ......................................................................................46 5.3 Instrumentation Plan ........................................................................................47
5.3.1 Overview...........................................................................................47 5.3.2 Strain Gages and Data Acquisition System ......................................47 5.3.3 Inclinometer ......................................................................................49
5.4 Statesville Load Test........................................................................................51 5.4.1 Instrumentation System ....................................................................51 5.4.2 Pile Installation .................................................................................53 5.4.3 Excavation.........................................................................................57 5.4.4 Challenges and Lessons Learned During the Statesville Test ..........58
5.5 Monroe Load Test............................................................................................60 5.5.1 Instrumentation System ....................................................................60 5.5.2 Pile Installation .................................................................................61 5.5.3 Construction Stop and Re-evaluation ...............................................63 5.5.4 Excavation.........................................................................................64
6 RESULTS OF FIELD LOAD TESTING PROGRAM ..................................................70
6.1 Statesville Test Site..........................................................................................70 6.1.1 Quantitative Results ..........................................................................70 6.1.2 Field Observations ............................................................................70
6.2 Monroe Test Site..............................................................................................74 6.2.1 Quantitative Results ..........................................................................74 6.2.2 Field Observations ............................................................................74
7 MODEL DEVELOPMENT............................................................................................79
7.1 Soil Strength Parameters..................................................................................79 7.2 Numerical Earth Pressure Models ...................................................................82 7.3 Earth Pressure Model Development ................................................................82 7.4 Impact on Numerical Models...........................................................................88 7.5 Impact on Retaining Structure Design.............................................................88 7.6 Parameter Selection .........................................................................................91
8 CONCLUSION...............................................................................................................93 8.1 Research Summary ..........................................................................................93 8.2 Conclusions......................................................................................................93 8.3 Recommendations............................................................................................94
9 LIST OF REFERENCES................................................................................................95
viii
ix
APPENDICES A STATESVILLE SOIL TEST DATA..........................................................................100 B MONROE SOIL TEST DATA...................................................................................101 C STRAIN GAGE DATA ..............................................................................................102 D INCLINOMETER DATA...........................................................................................118 E RECOMMENDATIONS FOR TRIAXIAL TESTS IN PIEDMONT RESIDUAL SOILS ..........................................................................................................................135
CHAPTER 1 INTRODUCTION
1.1 Introduction
Earth retention is an integral part of transportation infrastructure. Mechanically
stabilized earth, sheet pile, soldier pile, pile panel, soil nail, gravity and cantilevered are
different types of retaining structures utilized by the North Carolina Department of
Transportation. Regardless of the earth retaining method, the design of a retaining
structure is driven by the earth pressure exerted by the soil that must be retained.
Normal and shearing stresses exist within soil masses as well as at the interface
between soil and structure. The magnitude and orientation of the resultant forces depend
upon the strength and compressibility properties of the soil mass. In the case of a
retaining wall, the amount of soil stress that is transmitted to the structure is the net
amount that is not carried by the shearing strength within the soil mass. In the worst
case, the entire stress due to the soil overburden is imposed on the wall, this would be the
hydrostatic stress. However, most soils have internal shearing strength that will carry the
stress and reduce the amount of lateral earth pressure. Therefore, knowing the strength
and compressibility properties of the soil, the earth pressure on a retaining structure can
be determined theoretically or analytically.
Of course, the magnitude of earth pressure is affected by the type and nature of soil.
North Carolina has a diverse geologic footprint. Subsurface materials in the state range
from rock found in the mountains to sands of the coastal plain. However, several zones
of residual soils, known collectively as the Piedmont Physiographic Province, underlay a
large portion of the state. Figure 1.1 shows the location of the Piedmont and that a
1
number of major U.S. cities, from Trenton, New Jersey to Atlanta, Georgia, are all
located in the zone.
Figure 1.1 - Map of the Piedmont physiographic province (USGS 2001)
Within North Carolina, the Piedmont soils can be divided into three major zones:
the Inner Piedmont (Alexander, Buncombe, Burke, Caldwell, Catawba, Cleveland,
Henderson, Iredell, Lincoln, McDowell, Polk, Rockingham, Rutherford, Stokes, Surry,
Transylvania, Yadkin, and Wilkes counties), the Charlotte Belt (Cabarrus, Davidson,
Davie, Gaston, Guilford, Forsyth, Iredell, Lincoln, Mecklenburg, and Rowan counties)
and the Carolina Slate Belt (Alamance, Anson, Cabarrus, Caswell, Chatham, Davidson,
Durham, Granville, Guilford, Montgomery, Moore, Orange, Person, Randolph, Rowan,
2
Stanley, Union, and Vance counties). Figure 1.2 is a rough geologic map of North
Carolina showing the Piedmont zones with county and NCDOT division boundaries.
11
13
14
12
10
9
7
8
54
6
1
2
3
Figure 1.2 - County and NCDOT Division maps over North Carolina geology.
As a consequence of the formation process, residual soil often exhibits an increased
strength over comparable transported materials. However, this strength often goes
undetected by traditional sampling and insitu testing methods. Thus, the strength of
residual soil is often poorly characterized and underestimated. This is particularly
apparent when performing the standard penetration test, where many strength correlations
were developed for transported soils that were assumed to either be cohesive or
cohesionless, but not both. This assumption could mean that though a residual soil may
be classified as a cohesionless material, the additional strength available due to the fabric
structure and particle to particle bonding is ignored.
In terms of an engineering equation, the shearing strength of soil is represented by
the Mohr-Coulomb criteria τ = c’ + σ' tanφ’, where the shear strength τ is the sum of a
3
cohesive component c’ and geostatic stress σ’ multiplied by the tangent of the drained
friction angle φ’. Within this context, most soils are typically assumed to have c’ = 0.
However, residual soils tend to possess an appreciable amount of drained cohesion that
can have a substantial impact on the design of retaining structures.
1.2 Background
1.2.1 Retaining Structures
Retaining structures are used when a difference in elevation is required and a
constraint such as right of way, limited horizontal distance, or excessive elevation
change, prevents utilizing a slope. In terms of typical highway applications, retaining
structures are often used in bridge abutments, culverts and pipe wing walls, and roadway
cuts and fills.
Retaining structures can be categorized by construction method as either bottom up
or top down type. Bottom up would commonly be used for retaining walls required for
fills, but can be used in cut applications with temporary shoring. Such walls include
gravity, cantilevered concrete, mechanically stabilized or reinforced earth. Top down
walls are used for cut applications where the wall is installed during the cut process
without the need for shoring. Common examples include soldier pile/pile panel
(cantilevered or tied back), soil nails, and sheet piling.
1.2.2 Methods for Estimating Lateral Earth Pressure
Common to methods for retaining structure design is the determination of lateral
earth pressure. In the majority of cases, this is provided through the use of limiting
equilibrium earth pressure coefficients. When the displacement of the structure is
critical, the use of numerical techniques may be required.
4
1.2.2.1 Theoretical Approaches
There are three categories of lateral earth pressure - earth pressure at-rest, active
earth pressure, and passive earth pressure. Earth pressure at rest, σ0, refers to lateral earth
pressure caused by an unyielding wall preventing earth from any lateral movement. If a
wall is permitted to move away from the retained soil mass a slight distance, the soil will
expand laterally following the wall. Shearing resistance developed within the soil mass
acts opposite to the direction of the expansion, resulting in a decrease in lateral earth
pressure. The minimum lateral earth pressure is the active pressure, σ'a. Conversely, if a
wall moves into the retained soil mass, the soil will be compressed laterally with the soil
shearing resistance acting to oppose the lateral compression. The maximum lateral earth
pressure condition is the passive pressure, σ'p. The relationship between vertical stress
and the active and passive lateral stress for a soil with cohesion and internal friction is
shown on the Mohr’s circle presented in figure 1.3.
Normal Stress
φ’
c’
Shea
r Str
ess
σ 0 σ' pσ' a σv’
Figure 1.3 - Mohr’s circle relationship for soil with cohesion and internal friction
5
The ratio of horizontal effective stress to vertical effective stress in a soil mass is
referred to as the coefficient of lateral earth pressure, K. For the at-rest condition the
earth pressure coefficient is designated K0. For the active and passive pressure conditions
the earth pressure coefficient is designated Ka and Kp, respectively. Therefore,
KcK vh '2 '' ±= σσ (1.1)
where K is the appropriate earth pressure coefficient, depending on stress
conditions. The at-rest stresses for cohesionless soils are often determined using the
modified Jaky (1948) equation:
( )φσσ ′−= sin1'' v0 (1.2)
Traditionally, two general methods are used to estimate lateral earth pressure – the
Rankine (stress) and Coulomb (force). Use of these theories has resulted in successful
retaining wall design, but measured lateral earth pressure often differs largely from the
expected values calculated from these models. These methods and many others are
presented in a comprehensive text by Hansen (1961).
The Rankine theory for determining lateral earth pressure is based on several
assumptions, but the most important is that there is no adhesion or friction between wall
and soil. Pressures computed from Rankine theory are limited to vertical walls (with
vertical blackface), and failure is said to occur in the form of a sliding wedge along an
assumed failure plane that is a function of the friction angle of the soil. The equations
derived from these assumptions are widely used and, propitiously, the results may not
differ appreciably from those from other more fundamentally accurate analyses. Results
obtained from the Rankine method generally are slightly more conservative, resulting in a
small additional safety factor (Liu and Evett 2001).
6
The Rankine earth pressure model is illustrated in figure 1.4. The general equation
for the magnitude of the resultant lateral force, Pa, in the active pressure case is given as:
)Kc2KH()zH(21P aaca ′−−= γ (1.3)
where:
′⎛= −⎜⎝ ⎠
2tan 452aK ⎞
⎟φ
(1.4)
a
c Kc2z
γ′
= (1.5)
The resultant force acts at (H-zc)/3 from the base of the wall.
Pa
HKa - 2c Ka
-2c Ka
zc
H
H-zc
(H-zc)/3
Figure 1.4 – Rankine earth pressure theory
7
Coulomb’s theory for lateral earth pressure resulting from a retained mass of
cohesionless soil considers that a failure wedge forms behind the wall by sliding along a
plane. As the retaining structure moves away from the soil mass, lateral expansion is
permitted and results in a relative movement between the wall and soil causing friction to
develop on the back face of the wall (McCarthy, 2002). And so Coulomb, like Rankine,
assumed that the failure surface due to lateral earth pressure would be planar. The key
difference in these two methods, however, is that Coulomb took into account friction
between the back face of the wall and retained soil. For Coulomb’s method, the resultant
of this friction and lateral pressure, PA, acts at an angle, δ, measured normal to the back
face of the wall. When the failure wedge is satisfactorily retained by the wall, the forces
acting on the wedge are in equilibrium. Therefore, when the unit weight, γt, and friction
angle, φ, for a retained soil are known, the force imposed on the wall as a result of the
active pressure wedge, PA, can be determined by vector addition, shown in figure 1.5.
This force is calculated using the following equation:
( )
( ) ( ) ( )( ) ( ) ⎥
⎥⎥⎥⎥⎥
⎦
⎤
⎢⎢⎢⎢⎢⎢
⎣
⎡
⎥⎥⎦
⎤
⎢⎢⎣
⎡
−+−′′+
++
−′== 2
2
22
tA2
tA
coscossinsin1coscos
cosH21KH
21P
αθθδαφφδθδθ
θφγγ (1.6)
Similarly, the force imposed by the passive earth pressure wedge, Pp, can be
calculated for the Coulomb case using the following equation:
( )
( ) ( ) ( )( ) ( )
22 2
2
2
cos1 12 2 sin ' sin '
cos cos 1cos cos
p t p tP H K Hφ θ
γ γδ φ φ α
θ θ δθ δ θ α
⎡ ⎤⎢ ⎥⎢ ⎥′ +⎢ ⎥= =⎢ ⎥⎡ ⎤+ +⎢ ⎥− −⎢ ⎥
− −⎢ ⎥⎢ ⎥⎣ ⎦⎣ ⎦
(1.7)
8
``
`
Pa
H
A
B
C
D
F
θ
θ
θ
φ
W
W
F
Pa
θ φ
θ
φ
Figure 1.5 – Coulomb’s active pressure trial failure wedge and force polygon
The actual active pressure condition for a given wall results from a unique failure
wedge that provides the largest numerical value for the force PA. But the above equation
is only true for a uniform backfill slope and where the retaining wall is a plane surface. If
this is not the case, PA can be determined by trial-and-error analyzing a series of different
sized failure wedges (McCarthy 2002).
When the retained soil mass has both cohesion and friction, the vector addition
representing equilibrium forces acting on the failure wedge must include the additional
vector of cohesive resistance, cL, acting on the failure plane.
Because the Coulomb theory assumes a planar failure, whereas the actual failure
surface is typically curved, the lateral force calculated is slightly low. This discrepancy is
typically minor, and so the Coulomb procedure for active pressure determination
9
provides a practical accuracy. The value of PA computed from the Coulomb method will
be slightly larger than that of the Rankine method, but due to the difference in direction
that these forces act, the Rankine method typically creates the more severe condition and
results in a slightly more conservative value (McCarthy 2002).
1.2.2.2 Soil Structure Interaction Approach
Given that Rankine and others have provided a convenient method for
determining the lateral earth pressure, in any case, the soil mass must translate to achieve
the minimum active pressure. If the soil mass does not translate one of two cases
develop: either the at-rest pressure dominates or there is no lateral earth pressure on the
structure. The soil structure interaction approach (SSI) can take this effect into account.
Furthermore, this method allows for the beam behavior of a flexible wall to be included,
which is essential to this study.
The most common application of the SSI method is for laterally loaded piles.
Wang and Reese (1993) popularized the solution of the differential equation governing
the response of a pile under lateral load in COM624. The soil in that case was reduced
into a set of nonlinear slip springs connected to nodes on the pile, like those shown in
figure 1.7. The pile was treated as a bending element. The response curves (p-y curves)
represent a combination of active, passive, and shearing stresses on a single pile.
1.2.2.3 Finite Element Method
The finite element method (FEM) is often used in the analysis of complicated
geotechnical problems based on the stress-deformation concept. In addition, it provides a
rational method of attacking problems that are not readily solved using classic limiting
equilibrium analysis under elastic conditions in infinite half spaces.
10
P
y
Pu
k
Figure 1.7 - Simple p-y relationship
Simply stated, the FEM is a way to solve partial differential equations (PDEs)
numerically. Extended to civil-structural-geotechnical problems, the PDEs arise from
complex elemental shape functions, the determination of strain from stress, and
constitutive models that incorporate viscosity, plasticity, and pore fluid dynamics.
Taking a step back, a simpler perspective is that the finite element method is nothing
more than a complex way of expressing Hooke’s Law E σ=
ε. A model is developed
where either forces or displacements are imposed on boundaries. The resulting
displacements and stresses are determined based on the shape functions and constitutive
models selected.
In the context of the determination of earth pressure on a retaining structure, the
stresses are then determined based upon the propagation of stresses within continuum,
rotation of planes of primary stress, and deformation/translation of the soil mass. This
method allows for the behavior of discrete soil layers to be included while at the same
time considering the stress and deformation and their impact on one another.
11
1.3 Retaining Structure Design
rmination of lateral earth pressure, the design of a
tainin or
s.
g Structures
allows the
use of g
35
Given the methods for dete
re g structure is typically completed using limiting equilibrium concepts. Take f
example a gravity retaining wall. The wall is sized to resist overturning and sliding,
while the base is safe for bearing and settlement. MSE structures use similar concept
Sheet pile and solider pile walls are designed using a mixture of limiting equilibrium and
soil-structure-interaction. In any case, the driving horizontal earth pressure is typically
determined using Rankine’s active earth pressure coefficient.
1.4 North Carolina Department of Transportation Retainin
According to the 2002 NCDOT Roadway Design Manual, the NCDOT
ravity/semi-gravity, cantilever concrete, mechanically stabilized earth (MSE),
prefabricated modular, and pile/panel (can be cantilevered or tieback) wall systems. A
sampling of recent wall design projects within the Piedmont are shown in table 1.1. The
majority of these are soldier pile or gravity structures. The standard parameters used in
the design are also reported. It is worth noting that for the projects, the walls were
designed to retain cohesionless soils with the drained friction angles between 30 and
degrees.
12
Table 1.1 Recent retaining wall projects in the Piedmont
Project # County Type γ (pcf) φ' c’ (psf) B‐31 0 19 Buncombe Soldier 110 34 B‐1019 Anson Soldier 120 30 0 B‐3157 Davidson Soldier 120 30 0 B‐3406
Durham Fabric
B‐3446 Davidson Gravity 120 35 0 B‐3601
A 120 3 lamance Soldier 0 0
B‐3667 Davie Gravity 120 35 0 B‐3872 McDowell MSE 120 30 0 B‐4011 Ashe Soldier 120 30 0
B‐4043 Burke
Gravity 120 35 0
B‐4095
Davidson Gravity 120 35 0 B‐4258
Ru d therfor Soil Nail
B‐4317 Watauga Gravity 120 35 0 B‐3340 Haywood Soldier
120 3 0 0
I4025A Yad/Sur Soldier 120 30 0 NC112 Buncombe Soil Nail 120 30
R‐2107B
M y ontgomer Gravity 120 35 0 R‐
2233AA Rutherford MSE 120 30 0
R‐2248D Mecklenburg Soldier 120 30 0 R‐2518A Yancey Soldier 120 30 0 R‐2710 Watauga Fabric 120 30 0 R‐3415 Yadkin Soldier 120 30 0 R‐4758 Jackson Soldier 120 30 0 R‐3101
Alleghany Soil Nail
U M g‐2510A ecklenbur Gravity 120 35 0
U‐3401 Randolph
M y 120 3 SE/Gravit 5 0 U‐3456 Richmond Soldier 120 30 0
U‐3601 Buncombe Gravity 120 35 0
13
CHAPTER 2 PROJECT OVERVIEW
2.1 Objectives
objective of this research program was to develop a model for lateral
earth
ral earth pressures by earth pressure cells was
ruled o
s
the
The main
pressure in Piedmont residual soil. In order to develop the model, this program
examined both the properties of Piedmont residual soils and the field performance of
retaining structures at two research sites.
2.2 Research Methodology & Tasks
Since direct measurement of late
ut due to construction complications and survivability concerns, instrumented
sheet piles were used as “moment-cells” adapted from the analysis of deep foundation
under lateral loading. Measured earth pressure was correlated back to insitu and
laboratory determined soil parameters. Figure 2.1 shows a schematic diagram of
walls and the excavation.
Figure 2.1 - Conceptualization of the field test walls
Excavation
Variable Soil profile
14
The following outlines the different tasks of the project, and explains the variations
between the proposed and actual scope of work where applicable
1. Overview Retaining Structures: A detailed and extensive review of widely
accepted works and literature associated with retaining structures will be
completed.
Commentary: This task was completed unchanged.
2. Collection of Existing Data: The historic records of earth retaining structures
implemented in Piedmont residual soils and their performance will be collected
and analyzed. The performance of these structures will be examined and
correlated with the corresponding soil parameters
Commentary: This task was completed unchanged.
3. Site Selection: Working in close partnership with engineers and personnel of the
NCDOT Geotechnical Engineering Unit, three suitable test sites located in the
Carolina Slate Belt, the Charlotte Belt and the Inner Piedmont will be selected.
The selection of these sites will be based on the thickness of residual soils,
bedding and joint structure.
Commentary: Two sites, one in Statesville in the Carolina Slate Belt and the other
in Monroe in the Charlotte Belt were chosen by the NCDOT Geotechnical
Engineering Unit in cooperation with UNCC based on availability of funds.
4. Insitu and Laboratory Soil Testing: After the wall sites are selected, a rigorous site
investigation program will be completed. NCDOT and UNCC personnel will
work to perform standard penetration tests, Iowa Borehole Shear Tests, additional
specialized insitu soil tests to determine the engineering properties and geologic
15
fabric of the residuum. In addition, the probable penetration depth of the sheet
piles will be evaluated. Where possible, undisturbed samples of residual materials
will be collected for determination of engineering properties.
Commentary: Additional specialized insitu soil tests successfully carried out
included Dilatometer (Statesville and Monroe), Cone Penetration Test
(Statesville). Specialized tests to determine soil fabric were not available.
5. Preliminary Numerical Simulations: Finite element method models will be
developed to predict the earth pressure behind and deflection of the sheet pile
walls based on insitu test parameters and geologic information at each site. These
models will help fine tune the instrumentation and measurements for each test
wall to best match the field conditions.
Commentary: Finite element analyses were used to simulate the behavior and
response of walls for different excavation depths. The simulations were used to
select the sheet pile wall sizes, and assess wall deformation prior to field
instrumentation.
6. Prototype Wall Testing: In order to verify the function of field instrumentation,
two prototype sheet piles, one strain instrumented and one with the slope
inclinometer casing, will be tested in the laboratory. The sheets will be installed
with a fixed base and a point load will be applied at the opposite end to induce a
moment. The strain gages will be checked and monitored during the loading
process. The function of the slope inclinometer will also be verified. A third
uninstrumented sheet will be tested in series with the prototype sheets to compare
16
the bending stiffness of the instrumented and non instrumented sheet pile sections
and the effects of other environmental factors on the measurement system.
Commentary: This task was not performed because of delays in project
implementation.
7. Full-Scale Instrumented Field Test Walls: The primary task of the research
project will be the construction and excavation of the test retaining walls. At each
of the selected sites from task 3, a pair of parallel instrumented retaining walls
will be constructed, approximately 20 feet apart. Each wall will be 20 to 30 feet
in length and will be constructed of sheets driven 20 feet into the ground or
refusal. The soil between the two walls will be excavated 10 feet from the
original ground surface. Each wall will contain three strain instrumented sheets
and two slope instrumented sheets. Prior to and after excavation, slope
inclinometer readings will be taken. The strain gages continually will be
monitored over the duration of the excavation. After completing the excavation,
the sheets will be removed and transported to the next site.
Commentary: Apart from installation of the walls and constraints due to
subsurface soil conditions, this task was carried out as planned with some changes
to ensure for practical constructability of walls and safety.
8. Analysis of Results: Based on the results from task 6 and the research strategy
discussed above, lateral earth pressure distribution for each field sheet pile test
will be determined. The results obtained for each test site will be compared and
correlated to characteristics of the residual soil (thickness, bedding and joint
17
structure). A simple displacement and earth pressure relationship based on the
joint structure orientation will be proposed.
Commentary: This task was completed unchanged. It should be noted that
characteristics of the residual soil such as joint structure orientation could not be
determined.
9. Model Validation: The models developed in task 7 above will be correlated to the
engineering properties of the residuum obtained from both laboratory and insitu
tests. A relationship among earth pressure and measurable insitu soil properties
will be developed and recommended. Results of the models will also be compared
with classical earth pressure theory.
Commentary: This task was completed in accordance with the plan.
10. Reporting and Publication: The data from each phase of the study, the results of
analysis, and the project findings will be documented in a comprehensive report
to the NCDOT. In addition, findings will be presented as publications and
presentation to the TRB, STGEC, and Geo3 T2.
11. Commentary: This task was completed in accordance with the plan.
2.3 Significance of Work
Anecdotally speaking, this project was suggested based on the observation that
Piedmont residual soils produce little or no lateral earth pressure. This can be seen in
steep natural slopes or temporary cuts in these soils that appear to be stable over long
periods of time. Most engineers recognize this and account for some reduction in earth
pressure in their designs, however, this effect is almost purely by judgment.
18
The objective of this project was to develop a model that predicts earth pressure in
Piedmont residual soils for use in the design of typical retaining structures. Classical
earth pressure theory as well as numerical modeling using parameters derived from soil
tests at both research sites were used to predict earth pressure at these sites. The results
of the full scale load tests were used to validated the findings.
Based on these findings, the use of a reduced amount of earth pressure for
retaining structures in Piedmont soils appears justified. This will impact the type and size
of earth retaining structures that are available on a given project.
2.4 Limitations of Research
The limitations of this program bear mention. The original two-year duration of
the overall project, as well as the finite schedule of the sheet pile wall construction
projects, did not allow time to assess the long term changes in lateral earth pressure.
Unaccounted aspects of unsaturated soil behavior may have contributed to the behavior
observed. The second significant limitation was the number of research sites tested. The
proposed research program involved installation of sheet pile walls at three or more sites.
However, due to multiple factors, only two research sites were tested. Findings and
conclusions herein are limited by these factors.
19
CHAPTER 3 REVIEW OF RELEVANT LITERATURE
3.1 Residual Soil
Previous studies of residual soils have been performed by researchers in the mid-
Atlantic and southeastern United States, as well as the tropics, Europe, and China. The
bulk of the published research, in the United States and Europe, falls into one of three
groups: investigation at a research farm location by North Carolina State University
(1980s), field and laboratory testing at a National Geotechnical Research Site in Alabama
by Auburn University and Georgia Institute of Technology (1990s), and recent activity
on Portuguese soils attributed to Viana de Fonseca and Cruz (2000s).
3.1.1 Comprehensive test site studies
An extensive investigation of a single Piedmont residual soil test site was
documented in a doctoral dissertation by Heartz (1986) and a technical publication by
Lambe and Heartz (1988). The work was performed at a research farm near the North
Carolina State University campus, in Raleigh, NC. The goal was to characterize the site
by measuring compressibility, shear strength, and permeability, with special emphasis on
the effects of anisotropy, mica content, stress history, suction, and sample disturbance.
The soil at the site was found to be composed primarily of flat kaolinized biotite
particles oriented parallel to each other in fairly distinct tilted bands. One-dimensional
consolidation tests, drained triaxial tests, and unconfined compression tests along with
classification and index testing were performed on Shelby tube and hand trimmed block
specimens. The drained triaxial tests were isotropically consolidated, and falling head
permeability tests were performed before and after consolidation. Suction was measured
20
on unconfined compression test specimens before loading using a fine porous stone
mounted in a triaxial cell. Results revealed suction was well below one atmosphere.
Exploration efforts by Wilson (1988) and Harshman (1989) expanded on work by
Heartz at the test site. Wang (1995) and Wang and Borden (1996) describe the
weathering profile and deformation characteristics of the soils and attempted to evaluate
pressuremeter, dilatometer, standard penetration tests, and oedometer tests to predict
foundation settlements. The laboratory testing program included oedometer, triaxial
stress path and multistage compression, resonant column, and torsional shear tests. From
the same site, Gupta (1995) and Borden et al. (1996) developed relationships for residual
soil shear modulus, damping ratio, and volume change/densification characteristics based
on the influencing parameters of confining pressure, shear strain amplitude, and number
of loading cycles.
Similar effort was conducted by Vinson and Brown (1997) at a level III National
Geotechnical Experimentation Site near Opelika, Alabama at the southern extent of the
Piedmont province (Mayne et al. 2000). The original work was intended to serve as a
reference for soil characteristics for other research at the site and to compare different
types of insitu and laboratory measurements of physical properties of the soils at the site.
The insitu testing program included standard penetration, piezocone penetration,
pressuremeter, cone pressuremeter, dilatometer, borehole shear, cross- hole seismic,
seismic cone penetration, and seismic dilatometer tests. The laboratory testing program
included undrained triaxial tests, unconsolidated undrained triaxial tests, routine water
content, unit weight, grain size analysis, Atterberg limits, and other classification tests.
The results provide a basis for comparison of strength and stiffness measurements.
21
Analyses of the results of insitu and dynamic laboratory tests on soils from the site are
included in Martin and Mayne (1998), Hoyos and Macari (1999), Finke et al. (1999),
Finke et al. (2001), and Schneider et al. (1999).
3.1.2 Other works involving insitu and laboratory testing of residual soils
Some of the earliest published work on engineering aspects of residual soils was by
Sowers (1954 and 1963). Masters theses at the Georgia Institute of Technology during
the same period were published by Miller (1957) who investigated the behavior of a shear
vane in Piedmont residual soil and Crowther (1963) who correlated data from cone and
standard penetration testing in Piedmont residual soil to the bearing capacity of shallow
foundations.
In 1982, a conference was held on “Engineering and Construction in Tropical and
Residual Soils” that generated 37 papers (ASCE, 1982). The goal of the conference was
to discuss the unique properties of tropical residual soils by emphasizing aspects such as
composition and structure. The conference focused on the geotechnical properties, soil
investigation and field testing, slope stability, excavation and dewatering, along with
construction case histories.
In 1985, Brand and Phillipson (1985) published Sampling and Testing of Residual
Soils which included current practices from 19 countries around the world. The goal of
the book was to detail localized residual soil classifications, insitu and laboratory testing,
geotechnical issues, and specialized residual soil problems. One of the main conclusions
was that several authors place a larger reliance on insitu tests for residual soils than other
soil types, because of sampling and testing problems. The difficulty of sampling reduces
the reliability of lab testing and creates a discontinuity in results.
22
During the 1990s there was an increase in the use of insitu tests to characterize
residual soils. Frank (1990) at the Georgia Institute of Technology revisited the
relationship between SPT N and CPT qc values. A conference track at the ASCE annual
meeting in Charlotte, North Carolina in October of 1999 resulted in published GSP with
14 papers on residual soils. Of those, Failmezger et al. (1999) took a critical look at the
misuse of the standard penetration test for residual soil characterization and the use of
alternative insitu tests to characterize residuum. Kelley and Lutenegger (1999) used the
standard penetration test with torque and dynamic cone penetrometer to develop
relationships for engineering properties.
In the current decade, a great deal of literature has been generated concerning insitu
testing of Portuguese residual soils. These soils are derived from some of the same
ancient rocks as the Piedmont. Viana da Fonseca (2001) details a load testing of a
shallow foundation in residual soil. Cruz et al. (2004a) used dilatometer and cone
penetration tests to estimate the effective cohesive intercept, c’, in Portuguese residual
soils. Cruz et al. (2004b) took the same approach and developed it a step further to
conclude that residual soils are characterized by a true drained cohesion as well as a
multi-level yielding behavior that reflects the breaking down of residual structure under
stress. Gomes Correia et al. (2004) used insitu tests to obtain geotechnical parameters.
They reported that the bonded structure and fabric of residual saprolitic soils from granite
have a significant influence on geomechanical behavior. Rodrigues and Lemos (2004)
performed standard penetration, cone penetration, and cross-hole seismic tests and
discovered significant differences between those correlations and the ones established for
transported soils with identical grading curves. Viana da Fonseca et al. (2004) details the
23
investigation of a research site by extensive insitu and laboratory tests and presents
correlative results of qc/N and shear modulus (G0). Cruz and Viana da Fonseca (2006)
used the dilatometer to estimate the at rest earth pressure coefficient (K0), shear strength
parameters (c’ and φ'), and stiffness parameters (G0, E, and M).
3.1.3 Specific Literature on Triaxial Testing in Residual Soils
Between 1988 and 2006, 13 triaxial shear test studies have been published
specifically in residual soils. A summary of these authors, date, location, soil type, and
triaxial shear type are presented in Table 3.1.
3.2 Load Testing of Retaining Structures
Researchers such as Peck (1969) and Clough and O’Rourke (1990) have
investigated retaining wall pressure and earth movements due to excavations. Other case
studies of worldwide experience are also presented by Long (2001). In addition, studies
seeking to determine lateral earth pressures have been presented by researchers like Fang
et al. (1994) who utilized a movable model retaining wall to study passive pressures. The
same moveable retaining wall was presented by Fang et al. (1997) to study the effect of
sloping backfill on earth pressures. Similarly, Georgiadis and Anagnostopoulos (1998)
used a model cantilever sheet pile wall in a tank to study surcharge effects. Both studies
dealt with carefully placed sand in a controlled environment.
Full scale tests have been rare. A case was discovered where a full scale sheet
pile supported excavation was instrumented. Kort (2002) presents a study carried out on
an anchored sheet pile wall in Rotterdam, Holland. The soils were predominantly weak
clay and peat, and the study was concerned with oblique bending and plastic hinging of
sheet piles. However, there is extensive documentation of the field instrumentation plan.
24
Table 3.1 - Comprehensive summary of published residual soils triaxial testing Author Date Location Soil Type Test Type Numerical Results
Gan 1996 Hong Kong Undisturbed completely decomposed granite and fine ash tuff (saprolite)
CD Saturated and Unsaturated
Remolded: c’ = 1.45 psi and φ’ = 35.5o CDG: c’=3.05 psi and φ’ = 31o
Garga 1988 Brazil “Dense” Basaltic soil and “vesicular” basalt
CU Peak CU Dense: c’= 8.53 psi and φ’ = 32o Vesicular: c’= 5.26 psi and φ’ = 23o
Heartz 1986 North Carolina
Piedmont (Gneiss and Schist bedrock)
CD
Lambe and Heartz
1988 North Carolina
Piedmont CD
Lee and Coop 1995 Korea Decomposed granite (Compacted samples, highly organized)
CD
Mayne et al. 2000 Alabama Piedmont CD; CU Disturbed and Undisturbed
c’ = 2.47 psi and φ’ = 31o if c’=0 psi then φ = 35.3 degrees
Mohamedzein and Mohammed
2006 Sudan sandstone and mudstone
UU, CIU (Depth < 4m) c’ = 0 psi and φ = 29 -35o (Depth > 4m) c’ = 0 psi and average φ’ = 42o
Rahardjo et al.
2004a Singapore Reconstituted CD φ’ = 31.5o
Rahardjo et al.
2004b Singapore Residual CD Saturated and Unsaturated
φ’ = 41.3o or φ’ = 36o
Viana da Fonseca et al.
2006 Portugal Granite (saprolite with weak relict structure)
CD c’ = 0.65 psi and φ’ = 45.8o
Vinson and Brown
1997 Alabama Piedmont (micaceous sandy silt)
CD, CU, UU
Wang 1996 North Carolina
Igneous and Metamorphic rocks
CD Unsaturated
Wang and Yan
2006 Hong Kong 2 Saprolites Weathered volcanic tuff an Weathered Granite
CD CU
WT: φ’ = 36.6o WG: φ’ = 34.1o
25
CHAPTER 4 SITE LOCATION AND CHARACTERIZATION
4.1 Site Selection and Investigation Plan
Selection of the research sites was based on multiple criteria that were not
necessarily evident at the outset of the project. Foremost, the sites were to be located in
the Piedmont physiographic province. Based on the geology and funding opportunities,
research sites were suggested by the NCDOT Geotechnical Unit in the Charlotte
Belt/Inner Piedmont near Statesville and the Carolina Slate Belt, outside of Monroe.
Figure 4.1 shows the same map from figure 1.2 with the locations of the selected sites.
The Statesville site was located in a borrow pit being excavated for the adjacent (project
R-2911) US 70 bypass project. The Monroe site was part of the right of way being
purchased for the US 601 widening, (project R-2616). Other sites that had been initially
considered but were ruled out for various reasons included the Shelby Bypass, Monroe
Bypass, and another near Winston-Salem.
1
Figure 4.1 - North Carolina Geologic Map with counties, DOT divisions, and test site locations superimposed
1
13
14
12
10
9
7
8
54
6
1
2
3
26
To verify the existence of residual soils with sufficient thickness for wall testing,
several Standard Penetration Test (SPT) borings were conducted at targeted locations
within the test sites. Using the information from these borings, the sites were chosen and
the location for each wall construction was established.
In order to pinpoint the stratigraphy, a second round of SPT was completed at the
approximate location of measuring elements of the test walls. These borings were
extended forty feet below the ground surface. Further insitu tests included Marchetti
Dilatometer (DMT) to determine soil properties including lateral stiffness and earth
pressure coefficients as well as Iowa Borehole Shear Tests (BST) to determine insitu
strength. Cone Penetration Tests (CPT) and K0 Stepped Blade (K0SBT) were conducted
only at the Statesville Site with the latter being ineffective.
Shelby tube specimens that were collected in the preparation of boreholes for the
BST were sent to the Materials and Tests Unit for consolidated undrained triaxial and
consolidation testing. The same soils were classified and their specific gravities were
determined. The full complement of data is included in Appendices A and B for
Statesville and Monroe, respectively.
4.2 Statesville
4.2.1 Insitu Tests
The first test site was near Statesville inside a contractor controlled borrow pit
being excavated for use as fill for the US 70 Bypass. Three initial boring locations
indicated thick horizons of Piedmont Residual soil ideal for installation of sheet piles for
the project. Any of the tested locations would have been suitable for wall installation.
Targeted site exploration began just after the contractor began site clearing
operations. Due to the anticipated use of the site, the contractor dictated that only a small
27
portion of the borrow pit was available for the research project, and thus the location was
constrained near the northwest corner of the property. The approximate location for the
test walls was laid out and the second phase SPT was conducted. The boring locations
overlain on the walls that were eventually constructed are shown in Figure 4.2.
CPT-1
CPT-2
BST-1
CPT-4
CPT-3
BST-3
BST-2
KoSBT-1
SPT-1
SPT-3
SPT-2
SPT-4
DMT-2
DMT-1
NORTH
DMT-4
DMT-5
DMT-3
NW-I
NW-S
SW-I
SW-S
SE-S
SE-I
NE-S
NE-I
Figure 4.2 - Location of in-situ soil tests with respect to wall placement
Four SPT showed a profile consisting of tan and red clayey silt. Uncorrected
blowcounts ranged from 5 to 15 with an average around 8. The raw and corrected
numerical results of the SPT are show in Table 4.1. Two Marchetti Dilatometer (DMT)
soundings were completed to determine strength and compressibility parameters. Three
additional DMT soundings, including thrust measurements that were not available during
28
29
the initial tests, were conducted after the wall load tests. All of the soundings were
consistent as seen in Figure 4.3. Borehole shear tests (BST) were performed at several
depths in three boreholes. Key to the performance of the BST was the creation of a test
pocket using a Shelby tube. Of the twelve potential BST locations, eight were successful.
Results of these tests show generally friction angle of 30-40o and cohesion intercept
averaging 250 psf. The results are shown graphically and numerically in figure 4.4 and
table 4.2, respectively.
Additional tests, only conducted at the Statesville site were the Cone Penetration
Test (CPT) and the K0 Stepped Blade Test (K0SBT). S&ME Inc. performed four CPT
soundings with their truck mounted 20 ton rig, the results of which are shown in figure
4.5. The CPT showed a consistent tip resistance value of 20 tsf and sleeve friction
between 2 and 5 tsf. Only one K0SBT test was attempted. Multiple pressure cells were
damaged during penetration, therefore test was abandoned and not considered further.
Table 4.1 - SPT results (Statesville)
SPT-1 SPT-2 SPT-3 SPT-4 Depth
(ft) N N’60 Depth
(ft) N N’60 Depth
(ft) N N’60 Depth
(ft) N N’60
4.2 7 18.2 4.5 7 18.2 4.3 6 15.6 3.4 11 28.6 9.2 5 9.1 9.5 8 14.4 9.3 6 10.9 8.4 14 26.8
14.2 6 8.8 14.5 10 14.6 14.3 6 8.8 13.4 13 19.7 19.2 5 6.4 19.5 8 10.0 19.3 8 10.1 18.4 9 11.6 24.2 5 6.1 24.5 8 9.0 24.3 6 6.9 23.4 16 18.3 29.2 5 5.8 29.5 8 8.2 29.3 5 5.5 28.4 11 11.5 34.2 7 7.8 34.5 8 7.6 34.3 8 8.5 33.4 10 10.1 39.2 8 8.5 39.5 7 6.4 39.3 10 10.2 38.4 11 10.7
0 100 200(psi)P1
0 50 100(psi)
P0
0 25 50(psi)KD
0 2 4(psi)ID
0 2500 5000(psi)ED
0
5
10
15
20
25
30
35
40
0 4000 8000
Dep
th (f
t)
qD (lb)Thrust
DMT-1
DMT-2
DMT-3
DMT-4
DMT-5
30
Figure 4.3 - Summary of DMT results (Statesville)
0.00
5.00
10.00
15.00
20.00
0.0 5.0 10.0 15.0 20.0 25.0 30.0
Shea
r Str
ess
at F
ailu
re (p
si)
Normal Stress (psi)
Shear Strength Envelope
5ft Depth
10ft Depth
15ft Depth
0.00
5.00
10.00
15.00
20.00
0.0 5.0 10.0 15.0 20.0 25.0
Shea
r Str
ess
at F
ailu
re (p
si)
Normal Stress (psi)
Shear Strength Envelope
5ft Depth
9.5ft Depth
14.5ft Depth
0.00
5.00
10.00
15.00
20.00
0.0 5.0 10.0 15.0 20.0 25.0
Shea
r Str
ess
at F
ailu
re (p
si)
Normal Stress (psi)
Shear Strength Envelope
14.5ft Depth
20ft Depth
Figure 4.4 – Summary of Statesville BST results
31
32
Table 4.2 - BST Measured φ and c (Statesville) Depth φ c Test (ft) (deg) (psf)
BST #1 5 41.7 32 BST #1 10 41.2 130 BST #1 15 32.8 227 BST #2 5 31.2 767 BST #2 9.5 23.4 166 BST #2 14.5 4.8 522 BST #3 14.5 36 259 BST #3 20 32.8 144
4.2.2 Laboratory Tests
The specimens tested were from depths between 5 and 21 feet below ground level,
and all were collected in preparation of pockets for the BST. Classification tests were
performed on nineteen samples using Atterberg limits, sieve, and hydrometer analysis.
The specific gravity was determined as well.
Six one dimensional consolidation tests were conducted to determine
compressibility parameters. Nineteen triaxial tests, representing seven test locations,
were completed for the Statesville Site. All but three of these specimens were saturated
and tested in the consolidated-undrained (CU) condition. The others, from a shallow
depth, above the water table, were not saturated and tested consolidated drained (CD)
condition to ascertain the impact of unsaturated behavior.
The index tests, summarized in table 4.3, showed that the vast majority of the
soils were A-7. Consolidation tests suggest that the soils were all highly over-
consolidated, especially at shallow depths. Triaxial tests reported effective friction
angles between 25o and 35o with cohesion ranging between 200 and 400 psf. Modulus,
E50, values ranged from 200 to 3000 psi, for the confining pressures between 5 and 35 psi.
Tables 4.4 and 4.5 show the consolidation and triaxial test results, respectively.
Friction Ratio
0 6 12(%)
Tip Stress
0 50 100qc (tsf)
u
-2 2 6(tsf)
Sleeve Stress
0
5
10
15
20
25
30
35
40
0 2 4fs (tsf)
Dep
th (f
t)
CPT-1
CPT-2
CPT-3
CPT-4
33
34
Figure 4.5 – Summary of CPT results (Statesville)
Table 4.3 - Summary of index properties and classification testing (Statesville) Grain-Size Analysis Atterberg Limits
Minus #10 Fraction Borehole No.
Sample No.
Sample Depth
Specimen No.
#4 #10 #40 #200 #60 #270 Silt Clay
Liquid Limit,
LL
Plastic Limit,
PL
Plasticity Index,
PI
AASHTO Specific Gravity,
Gs
[ - ] [ - ] [ft] [ - ] [% passing] [%] [%] [%] [ - ] [ - ]
1 100 99 78 28 39.5 37.5 21 2 54 NP NP A-2-5(0) 2.643
2 ST-1 5.0 - 7.0
3 - 100 89 52 22.6 30.8 36.6 10.1 53 42 11 A-7-5(5) 2.717
1
2 ST-2 9.3 - 11.3
3
- 100 94 65 13.9 31.8 38.3 16.1 52 36 16 A-7-5(11) 2.74
1
BST-1
ST-3 19.3 - 21.3 2
- 100 89 58 18.2 34.2 35.5 12.1 56 40 16 A-7-5(9) 2.715
1 ST-2 9.0 - 11.0
2 - 100 97 82 6.3 16.2 41.2 36.4 73 43 30 A-7-5(31) 2.723
1 - 100 96 81 8.5 13.3 35.7 42.5 72 42 30 A-7-5(30) -
2 - 100 94 65 14.8 25.5 29.6 30.1 51 32 19 A-7-5(12) 2.671
BST-2
ST-3 14.0 - 16.0
3 100 99 93 59 15.6 32.4 35.9 16.2 51 42 9 A-5(6) 2.678
1 100 90 78 49 26.6 23 18.1 32.3 52 33 19 A-7-5(7) 2.684
2 100 96 80 44 31.1 29.1 21.7 18.2 48 38 10 A-5(2) 2.668 ST-3 14.0 - 16.0
3 - 100 94 54 15.7 39 33.2 12.1 54 47 7 A-5(4) 2.643
1 - 100 93 54 18.2 34.9 26.7 20.2 53 40 13 A-7-5(6)
2
BST-3
ST-4 19.0 - 21.0
3 - 100 91 50 19.6 39 29.4 12.1 53 39 14 A-7-5(5)
2.713
Table 4.4 - Summary of consolidation testing (Statesville)
Borehole No.
Sample No. Depth AASHTO
Compression Index,
cc
Recompression Index,
cr
Pre- consolidation
Pressure, Pc
Over- consolidation
Ratio, OCR
[ - ] [ - ] [ft.] [ - ] [ - ] [ - ] [psf] [ - ]
ST-1 5.0 – 7.0 A-2-5(0) 0.528 0.011 7,000 12.1
BST-1 ST-3
19.3 – 21.3
A-7-5(9) 0.515 0.035 9,000 4.0
ST-2 9.0 - 11.0
A-7-5(31) 0.225 0.003 6,400 5.8
BST-2 ST-3 14.0 -
16.0 A-7-5(30) 0.266 0.005 6,000 3.5
ST-3 14.0 - 16.0 A-7-5(7) 0.264 0.005 4,000 2.3
BST-3 ST-4 19.0 -
21.0 A-7-5(6) 0.421 0.006 7,000 3.3
Table 4.5 - Summary of triaxial shear testing (Statesville) Depth c’ φ' E50
Location (ft) Type Saturated (psf) (deg) (psi)
BST-1 6 CD No 318 24.3 933, 1791, 926
BST-1 10.3 CU Yes 255 29.5 1219, 2256, 2596
BST-1 20.3 CU Yes 410 28.7 2126, 3346
BST-2 10 CU Yes 411 27.1 588, 673
BST-2 15 CU Yes 0 34.4 192, 1477, 2459
BST-3 15 CU Yes 300 27.8 266, 1765, 2361
BST-3 20 CU Yes 404 26.1 607, 1861, 2372
35
4.3 Monroe
4.3.1 Insitu Tests
The second research site was chosen using a different set of criteria from
Statesville. The selection of the Monroe site was done primarily to be in line with the
future US601 widening project. Within that project, two locations of highest elevation
along the alignment were identified as potential test sites. One was a small field and the
other was a residence, both of which would be part of the right-of-way to be acquired in
the widening project. Based upon the initial SPT borings, both sites were potential test
locations. However, due to delays in the US601 project, right-of-way would not be
purchased in time, therefore, the southerly of the two potential sites was chosen and the
small amount of right-of-way was pre-purchased for use on the project.
Beginning in June of 2006, in depth insitu testing and soil sampling program was
carried out at the Monroe site. As before, the location of the test walls was established
before testing. Figure 4.6 shows the locations of the insitu tests with respect to the final
walls installed. Four SPT, depicted in table 4.6, were conducted down to 40 feet with
SPT1 refusing at around 28.5 feet. The soil type at the Monroe site appeared to be a
mixture of red clayey silt with uncorrected blowcounts that ranged between 9 and 19
blows per foot. The soils were markedly stiffer than those found in Statesville and more
pockets of quartz were found than previously. Four DMT were performed in proximity
to the test walls. In several cases, the DMT pushing was refused, however, the
obstruction was easily augered through and the test resumed. The results of the DMT are
shown in figure 4.7. Finally, BST at four depths were conducted in two locations. The
data in table 4.7 and figure 4.8 includes friction angles near 40o and cohesion as high 200
psf for the Slate Belt soil.
36
37
Figure 4.6 – Monroe soil test locations
Table 4.6 - SPT results (Monroe) SPT-1 SPT-2 SPT-3 SPT-4
Depth (ft) N N’60
Depth (ft) N N’60
Depth (ft) N N’60
Depth (ft) N N’60
3.5 30 78.0 3.9 17 44.2 4 17 44.2 4.1 10 26 8.5 15 28.5 8.9 13 24.2 9 15 27.7 9.1 9 16.5
13.5 11 16.6 13.9 11 16.4 14 13 19.3 14.1 12 17.7 18.5 10 12.9 18.9 9 11.5 19 14 17.8 19.1 12 15.2 23.5 10 11.4 23.9 9 10.2 24 10 11.3 24.1 15 16.9 28.5 *100 28.9 13 13.4 29 17 17.5 29.1 10 10.3
*Refusal 33.9 18 17.1 34 16 15.2 34.1 11 10.4 38.9 19 16.9 19 15 13.3 39.1 19 16.8
0 100 200(psi)
P0
0 300 600(psi)
P1
0 60 120(psi)KD
0 5 10(psi)
ID
0 10000 20000(psi)ED
0
5
10
15
20
25
30
35
40
0 5000 10000
Dep
th (f
t)
qD (lb)Thrust
DMT-1
DMT-2
DMT-3
DMT-4
38
Figure 4.7 - Summary of DMT results (Monroe)
0.00
5.00
10.00
15.00
20.00
0.0 5.0 10.0 15.0 20.0 25.0
Shea
r Str
ess
at F
ailu
re (p
si)
Normal Stress (psi)
Shear Strength Envelope
10ft Depth
15ft Depth
20ft Depth
0.00
5.00
10.00
15.00
20.00
25.00
30.00
0.0 5.0 10.0 15.0 20.0 25.0 30.0 35.0
Shea
r Str
ess
at F
ailu
re (p
si)
Normal Stress (psi)
Shear Strength Envelope
5ft Depth
10ft Depth
15ft Depth
20ft Depth
Figure 4.8 - Summary of Monroe BST results
Table 4.7 - BST calculated φ and c (Monroe)
Depth φ c Test (ft) (deg) (psf)
BST #1 10 43.0 0 BST #1 15 43.2 0 BST #1 20 40.1 201.6 BST #2 5 40.6 205.9 BST #2 10 23.4 165.6 BST #2 15 40.4 24.5 BST #2 20 42.0 144.0
39
40
4.3.2 Laboratory Tests
The specimens tested were from depths between 5 and 21 feet below ground level,
and were all collected in preparation of pockets for the BST. Classification tests were
performed on nine samples using Atterberg limits, sieve, and hydrometer analysis. The
specific gravities were determined as well.
Three one dimensional consolidation tests were conducted to determine
compressibility parameters. Nineteen triaxial tests, representing seven test locations,
were completed for the Monroe Site. All specimens were saturated and tested in the
consolidated-undrained (CU) condition.
The index tests, summarized in table 4.8, showed that the soils were
predominantly A-7. Consolidation tests suggest that the soils were all apparently highly
overconsolidated. Triaxial tests reported drained friction angle between 30 and 40o with
cohesion intercepts ranging between 100 and 500 psf. Modulus, E50, values ranged from
700 to 4000 psi, for confining pressures between 5 and 15 psi. Tables 4.9 and 4.10 show
the consolidation and triaxial test results, respectively.
4.4 Summary of Soil Testing
It is worth noting that even though both sites were in different sections of the
Piedmont, soil parameters determined from both were remarkably similar. The insitu and
laboratory tests both showed friction angles and a cohesive component to the strength.
Soils at both sites were AASHTO A-7 and appeared highly over consolidated.
Table 4.8 - Summary of index properties and classification testing (Monroe)
Grain-Size Analysis Atterberg Limits
Minus #10 Fraction Borehole No.
Sample No.
Sample Depth
Specimen No.
#4 #10 #40 #200 #60 #270 Silt Clay
Liquid Limit,
LL
Plastic Limit,
PL
Plasticity Index,
PI
AASHTO Specific Gravity,
Gs
[ - ] [ - ] [ft] [ - ] [% passing] [%] [%] [%] [ - ] [ - ]
ST-1 9.2-11.2 1 1 97 94 85 4.4 11.7 37.4 46.5 56 43 13 A-7-5(16) 2.8
1 - 98 89 73 13 20.1 38.6 28.4 43 29 14 A-7-6(10) 2.78 ST-2 14.2-16.2
2 2 96 93 80 6.1 14.2 33.1 46.6 63 43 20 A-7-5(21) 2.77 BST-1
ST-3 19.2-21.2 1 - 100 98 84 4.2 15.3 46.1 34.3 46 38 8 A-5-(10) 2.76
ST-1 4-6 1 - 97 92 81 8.1 10.9 26.2 54.7 58 35 23 A-7-5(22) 2.83
ST-2 9-11 1 - 99 97 88 4.0 9.1 24.4 62.5 65 46 19 A-7-5(24) 2.84
1 8 84 75 66 13.8 9.6 27.8 48.8 67 41 26 A-7-5(19) 2.77 ST-3 14-16
2 - 100 100 94 1.4 6.9 36.8 54.9 75 50 25 A-7-5(34) 2.82
BST-2
ST-4 19-21 1 - 99 98 90 1.6 9.9 43.8 44.7 64 50 14 A-7-5(21) 2.82
41
Table 4.9 - Summary of consolidation testing (Monroe)
Borehole No.
Sample No. Depth AASHTO
Compression Index,
cc
Recompression Index,
cr
Pre- consolidation
Pressure, Pc
Over- consolidation
Ratio, OCR
[ - ] [ - ] [ft.] [ - ] [ - ] [ - ] [psf] [ - ]
ST-1 4.0 - 6.0
A-7-5(16) 0.171 0.029 9000 16.3
ST-2 9.0-11.0
A-7-5(24) 0.350 0.042 3800 3.45
BST-2
ST-3 14.0-16.0
A-7-5(19) 0.353 0.032 3600 2.18
Table 4.10 - Triaxial results (Monroe) Depth c’ φ' E50
Location (ft) Type Saturated (psf) (deg) (psi)
BST-1 10.2 CU Yes 122.4 28.6 4029, 1518, 1945
BST-1 15.2 CU Yes 230.4 36.9 3404, 1683, 1301
BST-1 20.2 CU Yes 0 41.4 1078, 1559
BST-2 5 CU Yes 504 30.5 1765, 1071, 2604
BST-2 10 CU Yes 382 29.9 1760, 1051, 1505
BST-2 15 CU Yes 295.2 36.0 2229, 984, 1351
BST-2 20 CU Yes 417.6 28.6 722, 1283
42
CHAPTER 5 INSTRUMENTED WALL LOAD TESTS
5.1 Preliminary Modeling
The selection of the wall section and geometry was based upon a preliminary
numerical modeling program. Considered were length of wall, spacing between walls,
depth of walls, depth of excavation, type of pile, and instrumentation. In order to
simulate the plane strain condition, the walls ideally would have been several hundred
feet in length, driven to a depth of one hundred feet and excavated at least sixty feet deep.
This scenario, of course, was impractical from a financial standpoint, therefore the
configuration was optimized by modeling the potential configurations using the finite
element program PLAXIS (1998).
5.1.1 FEM Initial Conditions
Since the optimization was done before the test sites were chosen, average values
for soil properties such as cohesion, friction angle, modulus, and unit weight were based
on previous work in residual soils in North Carolina by Heartz (1986). Table 5.1
provides a summary of these values. Since there is some question concerning the use of
cohesion in the analysis, alternative values of 270 and 0 psf were also considered in these
preliminary models.
Table 5.1 - Summary of soil properties for modeling
Unit Weight 100 pcf φ’ 24.9 deg c’ 540 psf Permeability 0.85 ft/day E 1.440E+05 psf ν 0.15
Three sheet piles, PZ27, PZ22, and a lightweight 5-5 arch pile were considered.
The PZ27 was the sheet pile most commonly used by NCDOT. The PZ22 sheet pile had
43
44
the same cross sectional area as the PZ27, but wider section resulting in a lower moment
of inertia. The lightweight 5-5 arch pile, a flat sheet, was considered as a low cost and
high deflection option. Table 5.2 summarizes the structural properties of the sheet piles.
Table 5.2 – Summary of key pile properties used for modeling
PZ27 Moment of Inertia (I) 184.2 in4 Cross Sectional Area (A) 7.94 in²/ft Young’s Modulus (E) 29000000 psi Weight per foot of wall (w) 27 lb/ft
PZ22 I 84.7 in4 A 6.46 in²/ft E 29000000 psi W 22 lb/ft
5-5
I 5.7 in4 A 3.35 in²/ft E 29000000 psi W 11.3 lb/ft
5.1.2 FEM Modeling
FEM models were created for pile lengths of 20 ft., 25 ft., 30 ft. 35 ft. and 40 ft.
Excavation of the soil behind the sheet piles was simulated by removing 2.5 ft.
increments of soil to within 5 ft. of the tip of the pile. Each model was analyzed using
the full c’ = 540 psf, a half c’ = 270 psf, and a worst case scenario of c’ = 0. The
maximum deflection was compiled for each scenario, The results are shown in Table 5.3.
The models showed the PZ22 and PZ27 were similar in behavior, though as
expected the PZ22 exhibited greater deflection at the same length and excavation depth.
The 5-5 arch sheet pile, a much more flexible pile than the PZ22 and PZ27, was modeled
at 20 ft only. The 5-5 pile deflects much more than the PZ22 and PZ27.
Table 5.3- Summary of deflection versus depth
45
5.2 Selected Wall Section
The PZ22 appeared to be the best pile to use because it was similar to the PZ27
typically used by DOT and still provided ample measurable deflection. Though an
extreme amount of deflection was not desirable, for safety reasons, the larger the safe
deflection that could be obtained the better because the measurements are more distinct
and easier to interpret. The Arch 5-5 was ruled out because of the large deflections at
short pile lengths. Through additional models, the minimum separation between the
parallel walls was found to be 30 ft. for PZ22 sheet piles to prevent the overlap of the
passive soil wedges inside the excavation.
The appropriate length for the walls was determined by engineering judgment and
economic considerations. The wall length needed to be sufficient to ensure that the
instrumented piles in the center would exhibit plane strain behavior. If the walls were not
long enough, the piles at the end of the excavation, that were embedded in the access
slopes, would influence the wall sections. Thus, through discussion it was decided that
the test sections would be 40 feet.
Based on the results of the finite element simulations, 35ft. PZ 22 sheet piles were
selected. The minimum separation between the parallel walls was 30 feet. The test walls
would be 40 feet long and the soil between the walls would be excavated to 20 ft. When
working with the Geotechnical Engineering Unit to develop a plan for the excavation, the
issues of safety and access were raised. Therefore, in order to provide a 40 foot test
section, the walls would need to be extended further to make provision for safe access for
equipment and personnel to perform the excavation. The final plan included the addition
of 20 foot wing walls on each end of the test walls to protect slopes angled at 2:1 into the
excavation. There would be a total of 132 PZ 22 sheet piles (66 each side). Of the total,
46
80 were the full 35 feet long to be used for the test sections and partially for the access
ramps. The remaining 52 piles were 20 feet long reflecting the decreasing depth of the
access ramps.
5.3 Instrumentation Plan
5.3.1 Overview
The data acquisition program was comprised of three components, each designed
to provide redundant measurements that would allow for earth pressure calculations or
estimation. The most involved component was a network of strain gages on four piles
intended to give a representation of the strain along the length of the piles. The second
component was an inclinometer to determine the deflection and slope of four additional
piles. For redundancy, and a general check, points on the wall and within the excavation
were surveyed using conventional equipment. The strain monitoring system and
inclinometer are described in the following sections.
5.3.2 Strain Gages and Data Acquisition System
Measurement of bending moment was done indirectly through the measurement of
strain. Since the piles were steel and likely to remain in the elastic range throughout the
test, the bending moment was determined through conversion of strain to curvature
(Equation 5.1) then multiplying the curvature by the EI (Equation 5.2).
C T
yε − ε
κ = (5.1)
where κ is the curvature of the beam/pile, εC and εT are compression and tension strains, y is the distance between the gages ( 8.825 in)
M E= κ I where M is the bending moment (5.2)
E is Young’s modulus I is the moment of inertia
47
Strain was measured using arc weldable vibrating wire (VW) strain gages
manufactured by Geokon Inc. like the one shown in figure 5.1. The gages were installed
in tension-compression pairs on the flanges of the sheet piles, as shown in figure 5.2, at
regular intervals of length. The mounts were welded using a gauging block and blank.
After the mounts cooled, the gage was inserted into the mounts and set at the appropriate
amount of pre-strain by stretching or compression before locking with set screws. The
“plucker” which causes the wire to vibrate, and measures the frequency, was slipped over
the gage and secured with a hose clamp. When all of the gages and pluckers were
attached to the pile, the gages were covered with steel angle to protect them from damage
during pile driving. The strain was recorded using a pair of Campbell Scientific CR1000
dataloggers.
Plucker
Mount
Gauge Rod
Figure 5.1 - Geokon model 4000 strain gage
48
C
T direction of bending
49
Retained soil
Excavation
Figure 5.2 – Diagram of strain gage attachment 5.3.3 Inclinometer
In addition to the strain gages, the pile displacements were measured with an
inclinometer. Typically, an inclinometer operates by lowering the instrument into a
grooved PVC pipe set into soil. In the case of a sheet pile, a square steel tube section was
welded directly to the piles to substitute for the inclinometer casing, like the one shown in
figure 5.3. The measurement of deflection is made along the diagonal of the tube instead
of the oriented casing grooves. This skew was accounted for in the inclinometer
reduction software by mathematically rotating the readings such that the output was the
deflection along the bending axis of the wall. The primary measurement is along the A
axis and the rotated measurement is the X axis.
In the field, the A direction was always toward the northwest corner of the steel
tube. Therefore to rotate west side measurements to the wall bending axis, the skew
angle is 135o. Similarly east side measurements are rotated -45o. Figures 5.4 and 5.5
demonstrate this.
Figure 5.3 - Steel tube welded to a sheet pile as an inclinometer casing
50
135o Skew
X
A
Retained soil
Excavation
Figure 5.4 - Rotation of west side measurements to the axis of wall bending
51
Figure 5.5 - Rotation of east side measurements to the axis of wall bending
5.4 Statesville Load Test
The test site at Statesville was the first opportunity to put the concept into action.
Installation of the sheet piles was done through a purchase order contract between
Division 12 Construction/Bridge Maintenance and Burns and Sons Construction.
5.4.1 Instrumentation System
The initial instrumentation plan at the Statesville site included strain gages on four
piles and inclinometer tubes on four additional piles. Each side of the excavation had
four instrumented piles, 2 for strain and 2 for displacement, near the center of the wall
(within the 40 foot test section). Eight strain gages were attached to the non-excavation
side of both flanges of the strain instrumented piles at four foot spacing. A total of 64
gages was used – 16 per pile. Inclinometer tubes were attached to non excavation side of
the outermost flange of the displacement-measurement piles. The wall layout is shown in
figure 5.6.
-45o Skew
Retained soil A
X
Excavation
52
4’
train Gage S
Inclinometer Tube
Figure 5.6 – Instrumentation system for wall
5.4.2 Pile Installation
Work at the Statesville site started on Monday September 12, 2005. While
UNCC personnel were installing strain gages, the contractor began work on driving the
sheet piles. Initially, the contractor leveled the site to provide a good working platform.
Next, string lines were pulled to locate the walls and a jig was set up to get the first panel
of piles in position and plumb prior to driving. Initial driving was done using a relatively
small vibratory hammer.
The panel method of driving three piles as a set was used to help with pile
alignment, decrease the chance of pile wander, and insure proper set. However, many
piles refused under vibratory driving. An attempt was made to drive many of these piles
further with a diesel hammer. The diesel hammer was able to drive some of the piles
deeper, however due to the makeshift nature of the pile helmet, and lack of any driving
criteria, several pile tops were overstressed and yielded. Near the end of driving, a proper
sheet-pile driving helmet was implemented. The new helmet and maintaining a driving
criteria of 10 blows per inch, made a significant difference in not over stressing the piles
though several of the piles were still not able to be driven completely without damage.
Thus, several of the piles were partially driven. However, all of the sheet piles on the
original plan were eventually installed. Burns and Sons wrapped pile driving on
September 30, 2005. Several photos from the construction phase in Statesville are shown
in figure 5.7. The heights of the piles above the ground surface is shown in figure 5.8.
This is also reflected in the position of the strain gages. The relative positions of the
strain gages with respect to the ground surface are shown in figure 5.9.
53
54
F ure .7 – hoto raph from wall nstal tion n St esviig 5 P g s i la i at lle
55
end x x x x x x x x x x x 7' 7' 6' 14' 14' 4' 4' 3' 4' 3' 3' 3' 2' 2' 7' 7' 3' x x x x x 3' 3' x x 3' 3' x x x 2' x x x x x wall continues but mostly driven
PIT Entrance --->
end x x x x x x x x x x x 9' 8' 7' 7' 6.5' 6.5' 11' 11' 11' 11' x 6.5' 7' 7' 7' 7' 9' 9' x x 6' 6' 6' x 4' 4' 2' 9' 9' 3' 3' 3' 3' x x x x x wall continues but mostly driven
= Inclinometer= Vibrating Wire
Figure 5.8 - Height of piles above ground (x = fully driven)
N
Figure 5.9 - Gage diagram of pile elevations and excavation depths
Pile 1 (NW)
Pile 2 (SW)
Pile 3 (SE) Pile 4
(NE)
1
2
3
4
8
7
6
5
9
10
11
12
16
15
14
13
17
18
19
20
24
23
22
21
25
26
27
28
32
31
30
29
33
34
41
42
43
35
36
4
4
40
39
4
4
38
37
4
4
8
7
6
5
49
50
57
58
59
51
52
60
64
56
55
63
62
54
53
61
0 ft
4 ft
8 ft
12 ft
16 ft
20 ft
56
5.4.3 Excavation
After the completion of pile driving, the strain gages were connected to the
dataloggers. The dataloggers and excess cabling were secured in weather resistant (traffic
signal) boxes. The dataloggers were set to take readings continuously every one minute
until the load test was complete. Two sets of baseline inclinometer r gs were taken.
Notably, heavy rains due to the remnants of Tropical Storm Ophelia delayed the start of
excavation by nearly two weeks. The excavation process proceeded as follows:
- Monday October 17, 2005 ~ Excavation began early morning but stopped at
11:00am after excavating one quarter of the first 4 ft. lift, due to a hydraulic
failure. Excavation resumed in the afternoon to bring the total progress on the lift
to about 1/3.
- Tuesday October 18, 2005 ~ First lift (4 ft.) was completed by 11:00am. The
second lift started around 2:00pm.
- Wednesday October 19, 2005 ~ Excavation for the second lift was resumed in
the morning and was complete by noon. The third excavation lift was started
around 3:00pm.
- Thursday October 20, 2005 ~ Third layer (12 ft.) was completed by 3:00pm..
- Friday October 21, 2005 ~ Fourth excavation step (16 ft.) began at 7:30am and
was completed at 3:00pm.
- Wednesday October 26, 2005 ~ Excavation of final 4 ft. lift (20ft) was started
and finished at 2:00pm.
- Thu ing.
eadin
rsday October 27, 2005 ~ 20 ft. lift readings were taken in the morn
57
58
nal reading has been taken, an opportunity arose to explore the
reed
ade to ensure the sheet piles are driven to proper depth, especially
o
driven nine feet short of the required depth.
of
that even though it was confirmed with the
e
throughout the test. It is difficult to say if the haul
e
piles. Near the depth where soil makes contact with the wall, there should be sufficient
After the fi
unsaturated effect in qualitative terms. The contractor working on the site ag
to flood the soil behind the east wall with several thousand gallons of water. .
- November 1, 2005 ~ All instrumentation removed and walls decommissioned
- November 5, 2005 ~ Burns and Sons remove sheet piles and transports
them to Statesville Maintenance Facility.
Figure 5.10 shows highlights of the excavation.
5.4.4 Challenges and Lessons Learned During the Statesville Test
Being this was the first wall installation of a proposed three to four walls, there were
several observations made that would be carried forward to future load tests. First, a
better effort could be m
the instrumented sheets. Only one strain gage and one inclinometer sheet were driven t
full depth. One inclinometer sheet pile was
Second, the test wall sheets should be disconnected from those protecting the access
ramp. Next, the research team and NCDOT should have better control of the location
the test walls on the site. It was unfortunate
grading contractor that the area between the west wall and the property line would not b
used for hauling, and any existing haul road would be abandoned when testing started,
the haul road continued to be used
road had any impact on the tests. However, data collection and site access were far mor
challenging.
In terms of instrumentation, there should be more strain gages on the measuring
59
Figure 5.10 – Photographs of c io S v
ex avat n in tates ille
gages to capture the function that defines bending moment with depth. In Statesville,
there were strain gages at 8 depths – spaced 4 feet apart. This was done, primarily, to
provide full depth gages as well as work within the project budget. The gages should
have been clustered around the potential depth of soil contact on the wall or used in a
tighter configuration. In general, the inclinometer was under-used on the project. There
could have been many more inclinometer instrumented piles with little additional cost to
the project.
Overall, there should have been sufficient provision to try to get “the most bang for
the buck” on the test walls. Since deflections, stresses, and moments were less than
expected, addition of a surcharge load behind one or both walls would have given an
additional learning opportunity for the test. In addition, truly saturating the backfill to
relieve any negative pore pressure would help rule out soil suction as a source of the
increased soil strength/decreased earth pressure.
5.5 Monroe Load Test
Building on the experience in Statesville, the instrumentation plan was modified
for the Monroe test site. In addition, it was proposed to provide a surcharge on one side
of the excavation to force an active earth pressure wedge to develop and drill wells
through which the retained soil on the other side could be saturated. Unfortunately,
insufficient right of way, unavoidable overhead utilities, the location of a drinking water
well, and proximity to residences prevented these additional measures. Sheet pile
installation was conducted by NCDOT Division 10 Bridge Maintenance personnel.
5.5.1 Instrumentation System
Based upon lessons learned e Statesville test, the instrumentation p w
amended to maximize the ability to measure earth pressure at the second site. From the
from th lan as
ve
rest microstrain was set for every gage (new and existing). In addition to the
ame
lp
Per recommendation based on
a larger vibratory hammer, ICE model 22 hydraulic, was
icult
he
of the sheet piles were lubricated to minimize friction in
the
previous test four of the sheet piles were already instrumented with strain gages every
four feet. To increase the resolution of the data for subsequent tests, additional strain
gages were installed between the original gages, such that there would be a strain gage
every 2 feet along the pile (Figure 5.11). In order to install the new gages, the protecti
covers were removed. The new gages were welded on the sheet pile in line with the
previous gages at a spacing of two feet on center. Before reinstalling the protective
cover, the at-
nded strain gage plan, inclinometer tubes were attached to four additional piles to
bring the total inclinometer piles to eight.
5.5.2 Pile Installation
After the piles were instrumented, NCDOT Division 10 bridge maintenance
began driving the sheet piles on Monday October 2, 2006. A string line was set to he
keep the wall aligned during the driving process.
experience in Statesville,
utilized to drive the sheet piles. Unlike the previous work in Statesville, pile driving in
Monroe proved to be rather difficult due to many factors. The short piles used to protect
the access ramps were installed quickly. Starting with the first long piles, driving
difficulties began to surface.
The most obvious factor is the piles were not “new” this time. Thus any damage
from the work in Statesville, due to driving and extraction, would contribute to diff
driving. Thus, some piles were slightly bent, and had rust and other imperfections in t
sockets. The balls and sockets
joints, and help the piles drive easier. Geologically, thin layers of quartz that were
found during the borings may have hindered driving in some locations.
62
Figure 5 – nded strain g pl for Monroe t it
est s e
S e train Gag
2’
Inclino eter Tube
m
.11 Ame age an
Thus, the ends of many piles were cut on an angle to resemble a pile point. Likely the
biggest factors proved to be the lack of the panel method of driving and inexperience of the
crew with sheet piling. The Division 10 t preferred using the set and drive method
whereby each sheet is continuously driven to its full length without stopping. This does not
provide the confinement necessary to keep the piles in line during driving.
Although the remedial measures improved the pile driving, the process was still very
slow, and it became very difficult to keep the piles plumb. In several instances, the top of the
pile would break off in the jaws of the hammer while driving, and the pile would have to be
cut off in order to continue driving.
5.5.3 Construction Stop and Re-evaluation
After several working days attempting to drive sheet piles, the effort was stopped on
October 18, 2006 in order to regroup and determine if continuation work in Monroe wa
feasible. During a meeting with the Division Engineer on October 19, 2006, the consen
was to make every attempt to drive the piles at the research site. To that end, additional
auger borings were completed on October 26 and 27, 2006 within the wall alignment to
determine the location of rock that might hinder driving. In addition, a diesel hammer w
borrowed from Division 9 bridge maintenance.
Rock was found in the south western portion of the wall footprint, however the
alignment of test sheets could be amended such that its impact could be minimized.
Knowing that rock was not the limiting factor, pile installation was restarted on Monday
October 30, 2006. Still, the driving difficulties persisted. Driving operations were
suspended on Wednesday ber 29, 2006 for budgetary reasons and complaints from
eam
s
sus
as
Novem
63
64
ws
and twi
ng out of plumb and alignment. The stickup heights
.15, and the relative locations of the strain
gages a
er,
f
, the excavation process was started and
precede
nd
r a
m
g
and was completed by 11:00am. The third lift (12ft) was started after lunch, and was
completed by 3:30pm.
residents living near the site. Figure 5.13 shows photographs of the wall construction in
Monroe.
The resulting test walls were far shorter that those in Statesville. Figure 5.14 sho
the 26 short piles and 36 long piles that were eventually driven. Most of the long piles were
damaged during installation. The NW inclinometer casing broke loose from the sheet pile
sted rendering the data unusable. During installation of the NW strain gage pile, the
protective cover vibrated loose damaging many wires in the process. The final piles were
driven without supervision by UNCC and without any attempt to plumb them with a template
resulting in several measuring piles bei
of the sheet piles installed is shown in figure 5
re shown in 5.16.
5.5.4 Excavation
Once the piles were driven, the strain gages were connected to the datalogger, and
baseline readings were taken. Also, the baseline readings were taken with the inclinomet
and the location of each instrumented pile was surveyed using a total station. All three o
these methods were used throughout the excavation sequence to determine the deflection of
the wall. With all of the baseline readings taken
d as follows:
- Monday December 11, 2006 ~ The excavation began in the early morning, a
continued until about noon. There was a slight delay while the operator waited fo
truck to haul the material on site. The first lift (4ft) was completed by 4:00p
- Tuesday December 12, 2006 ~ The second lift (8ft) was started in the early mornin
65
llati n in onr
wall insta o M oeFigure 5.13 – Photographs from
NO
RTH
NW-I
SW-I
SE-SSE-I
NE-S
SW-S
NW-S
NE-I
13 Short Piles13 Short Piles
US601
unlock
Figure 5.14 – Sheet piles as installed in Monroe
Figure 5.15 - Height of sheet piles above ground level
N
66
67
Figure 5.16 - Gage diagram of pile elevations and excavation depths
- Wednesday December 13, 2006 ~ Excavation for the fourth lift (16ft) was
started in the early morning, and was completed by 1:30pm. The fifth and final
lift (20ft) was started, but the excavation was halted because one of the dump
trucks was damaged in the stockpile area and required towing.
- Thursday December 14, 2006 ~ The final lift (20ft) was continued in the early
morning and completed by 12:00pm. Readings were taken after the final lift was
completed. After the final readings were taken, an 8ft triangular surcharge was
built behind the west wall. Readings were also taken after the surcharge was
completed at 3:30pm.
- Sunday December 17, 2006 ~ The final readings were taken in the afternoon, and
the dataloggers were disconnected and walls decommissioned.
The excavation process is shown in figure 5.17.
68
69
Figure 5.17 – Photographs of excavation in Monroe
CHAPTER 6 RESULTS OF FIELD LOAD TEST PROGRAM
6.1 Statesville Test Site
As detailed in Chapter 5, the sheet piles were instrumented with strain gages and
inclinometer tubes. In addition to the numerical data, there were qualitative observations that
provided important information about the behavior of the soil.
6.1.1 Quantitative Results
The quantitative results are presented as plots of bending moment and lateral
deflection with depth. Figure 6.1 shows the bending moments induced as the excavation was
completed. Based on the sign convention shown in figure 5.2, bending into the excavation
should be a positive moment. Negative moments would then mean that the wall bent
backward into the wall of the excavation. In several cases, the bending moment appeared to
be negative. This is likely due to the fact that the residual stresses locked in during driving
were released when the soil was excavated, and there was not sufficient earth pressure on the
wall to reverse this effect.
Deflections shown in figure 6.2 are affected by the pile stick-up height. The SE
inclinometer pile was grossly underdriven and the deflections were magnified. The fully
driven piles on the west side of the excavation showed displacements that were less than one
inch.
6.1.2 Field Observations
Two important observations were made during the course of the load test. First, the
sheet piles, were disconnecting from the soil mass. This was visible and could also be
deduced in the strain gage data. Figure 6.3 is a magnified view of the disconnection.
70
-20000 0 20000 40000 60000Moment (lb*ft)
-60000 -40000 -20000 0Moment (lb*ft)
-10
-5
0
5
10
15
20
25
30
35
20000 40000 600
Dep
th (f
t)
-10
0
5
-60000 -40000
(ft)
NW - S NE -S
-5
10
15Dep
th
20
25
30
35
-60000 -40000 -20000 0 20000 40000 60000Moment (lb*ft) Moment (lb*ft)
-60000 -40000 -20000 0 20000 40000 6000-10
-5
0
5
10
15
20
25
30
SW - S SE - S
35
Dep
th (f
t)
Post Driving4ft Excavation 18-Oct8ft Excavation 19-Oct12ft Excavation 20-Oct16ft Excavation 21-Oct20ft Excavation 27-Oct
-10
-5
0
5
10
15
20
25
Figure 6.1 - Measured bending moments per pile and excavation step without residual
stresses
Dep
th (f
t)
30
35
71
-10
-5
0
5
10
15
20
25
30
35
-0.5 0 0.5 1 1.5 2
Dep
th (f
t)
Displacement (in)
4 ft Excavation 18-Oct 8 ft Excavation 19-Oct
12 ft Excavation 20-Oct 16 ft Excavation 21-Oct
-10
-5
0
5
10
15
20
25
30
35
-0.50.51.52.53.5
Dep
th (f
t)
Displacement (in)
20 ft Excavation 27-Oct
-10
-5
0
5
10
15
20
25
30
35
-0.5 0 0.5 1 1.5 2
Dep
th (f
t)
Displacement (in)
-10
-5
0
5
10
15
20
25
30
35
-0.50.51.52.53.54.5
Dep
th (f
t)
Displacement (in)
NW - I
SW - I SE - I
NE - I
Figure 6.2 - Measured lateral dis
placements per pile and excavation step.
72
As previously mentioned, the Statesville site was inside an active borrow pit for the US
70 bypass project. The contractor planned to excavate 40 feet of material from the site
through the course of the construction project. The contractor’s excavator can be seen
loading a truck in figure 6.4. Note the scale of the truck versus the excavated depth of the
haul road.
Figure 6.3 - Disconnection of sheet piles from soil mass
Figure 6.4 - Vertical excavation of soil onsite
73
6.2 M
oe
in Monroe impacted the results.
6.2.1 Quantitative Results
The quantitative results are presented as plots of bending moment and lateral
deflection with depth. The first two plots show the measured bending moment versus depth
for two cases. Figure 6.5 shows the bending moments induced as the excavation was
completed. The detected bending moments were nearly zero. The negative bending
moments again were likely due to the fact that the residual stresses locked in during driving
were released as happened in Statesville.
Lateral deflections from the slope inclinometer are shown in figure 6.6. Construction
difficulties resulted in damage to the sheet piles that more than likely forced them out of
horizontal and vertical alignment. As mentioned previously, one of the inclinometer casings
was twisted loose from the pile when installed. The most properly driven inclinometer pile
was in the north east location. Measured deflections are still likely due only to release from
the wall and tilting rather than earth pressure.
6.2.2 Field Observations
During the excavation, the piles were separated from the soil mass, leaving the
retained soil standing without support. In order to estimate the depth of separation, the end
was removed from a tape measure, and the tape was inserted behind the sheet pile wall in
several locations. In every case, the tape measure extended the full 20 excavation depth
onroe Test Site
As before, bending moment and lateral displacement were measured at the Monr
test site. The number of strain gages was roughly doubled in an attempt to increase the
resolution of the bending moment distributions. However, the construction challenges
experienced
74
-10.00
-5.
15.00
20.00
25.00
30.00
35.00
-60000 -40000 -20000 0 20000 40000 60000
Dep
t
Moment (lb*ft) Moment (lb*ft)
-10.00
.00
15.00
20.00
25.00
30.00
35.00
-60000 -40000 -20000 0 20000 40000 60000
Dep
t
NE -SNW - S
00
0.00
5.00
10.00
h (ft
)
-5.00
0.00
5.00
10
h (ft
)
Post Driving4ft Excavation 18-Oct8ft Excavation 19-Oct12ft Excavation 20-Oct16ft Excavation 21-Oct20ft Excavation 27-Oct8ft Surcharge West
-10.00
-5.00
Moment (lb*ft)
0.
10.00
15.00
20.00
25.00
30.00
35.00
-60000 -40000 -20000 0 20000 40000 60000
De
SW - S SE - S-10.00
Moment (lb*ft)
-5.00
10.00
15.00
20.00
25.00
30.00
35.00
-60000 -40000 -20000 0 20000 40000 60000
De
stresses
00
5.00
pth
(ft)
0.00
5.00
pth
(ft)
Figure 6.5 - Measured bending moments per pile and excavation step without residual
75
-10
-5
0
5
10
15
20
25
30
35
-0.5 0 0.5 1 1.5 2D
epth
(ft)
Displacement (in)
4 ft Excavation 11-Dec 8 ft Excavation 12-Dec
12 ft Excavation 12-Dec 16 ft Excavation 12-Dec
20 ft Excavation 14-Dec 20 ft Excavation + Surcharge 14-Dec
-10
-5
0
5
10
15
20
25
30
35
-0.500.511.52
Dep
th (f
t)
Displacement (in)
-10
-5
0
5
10
15
20
25
30
35
-0.5 0 0.5 1 1.5 2
Dep
th (f
t)
Displacement (in)
-10
-5
0
5
10
15
20
25
30
35
-0.500.511.52
Dep
th (f
t)
Displacement (in)
NW - I
SW - I SE - I
NE - I
placements per pile and excavation step. Figure 6.6 - Measured lateral dis
76
before the soil came in contact with the sheet pile wall. Figure 6.7 shows the separation of
the sheet pile wall from the retained soil.
The reduced wall configuration resulted in the elimination of one of the access ramps,
thus there was a section of the excavation with a near vertical slope. Figure 6.8 shows this
steepened slope that illustrates the strength of the soil mass in the field.
From the field observations, it was determined that there was no lateral earth pressure
acting on the retaining walls. Therefore, the soil mass would have stood by itself
unrestrained.
Figure 6.7 - Wall separation
77
Figure 6.8 - Unsupported soil mass (20ft excavation)
78
79
CHAPTER 7 MODEL DEVELOPMENT
The strength and compressibility parameters obtained from the in-situ soil tests and
lab tests were used to estimate the amount of lateral earth pressure that would occur in the
field. Several methods were used to quantify the lateral earth pressure and estimate the
deflection of the test wall, including Rankine’s earth pressure theory, soil structure
interaction, and the finite element method.
7.1 Soil Strength Parameters
Since the model would be based upon soil parameters measured in the lab or insitu, a
first step was to develop composite profiles for both sites. The borehole shear or triaxial tests
produced direct measurements of soil parameters c and φ or c’ and φ’. However, the other
insitu tests report intermediate or index parameters that must be correlated to engineering
properties. There are a multitude of correlations to the SPT test. For the sake of simplicity,
the equation published by Peck et al. (1974) was used to determine the friction angle from
the N’60 value. A co ulate su from N’60
as an indicator of c’. Young’s modulus was determined from an equation in the Soil
Properties Manual by Kulhawy and Mayne (1990). Similarly, constrained modulus, friction
angle, undrained shear strength were correlated from the dilatometer soundings based upon
work by Marchetti (1980, 1997). And finally, for the CPT tests in Statesville, Robertson and
Campanella (1983) was employed to determine friction angle from qc, and Baladi et al.
(1981) was used to find Young’s Modulus. In order to develop composite useful profiles, the
soil parameters, measured and correlated, were averaged at 5 feet depth intervals. Tables 7.1
and 7.2 summarize the soil properties for Statesville and Monroe, respectively.
rrelation published by Sowers (1979) was used to calc
80
r a g l e f e es ar it D X
Table 7.1 - Summa y of vera e soi prop rties or th Stat villeSPT CPT MT
rese ch s e BST T L
Dept E q c φ' c’ E Depth h N N’60 φ su E qc φ su D M φ su E φ (ft) ( ) f) ) ) f) si ts (p ( f o (p (ft) bl./ft) (bl./ft (o) (ps (psi (tsf (o) (ps (p ) ( f) si) o) (psf) (psi) (o) (ps ) ( ) (psf) si)
GS=0 GS=0 5.0 8 20.7 5 .6 3 8 4 2. 52 4 7 5 0 4.3 1 12 5.0 7. 33.5 1653 301 20 39. 270 63 9 19 1 17 3877 36. 40 2 3 8 17
10.0 3 15.5 6 .2 3 1 0 3. 39 3 8 32.3 8 8.3 3 148. 31.9 1237 225 19 35. 248 58 2 24 8 03 2915 14 2 3 3 66 10.0
15.0 8 13.2 6 .4 8 0 4 3. 33 3 9 24.5 6 1.1 5 148. 31.1 1056 192 17 32. 221 54 7 09 7 13 2458 33 3 1 0 20 15.0
20.0 5 9.9 6 .5 5 9 6 3. 34 32 4 7.4 0 207. 30.0 793 144 19 31. 247 61 9 26 36 1079 2545 .8 14 2 4 7 62 20.0
25.0 8 10.9 .4 5 6 .6 3 8 16 4. 27 - - - - 8. 30 87 159 19 30. 247 6 7 80 35 1235 2065 - 25.0
30.0 3 8.6 0 8 .0 3 2 4 4. 30 - - - - 7. 29.6 69 125 20 30. 252 63 3 27 36 1185 2249 - 30.0
35.0 3 9.4 0 8 .9 5 3 89 6. 32 - - - - 8. 29.8 75 136 22 29. 289 6 2 30 35 1343 2400 - 35.0
40.0 0 9.8 5 2 .4 0 3 53 7. 23 - - - - 40.0 9. 30.0 78 143 21 29. 268 6 3 02 34 1471 1710 -
G = .) below the S round Surface
WT 18 (ft G (G )
81
T DM BST T
Table 7.2 - Summary of average soil propertie he Monroe research site s for t SP T XL
D h ept N N’60 s E q h φ u D M φ su E φ c φ' c’ E Dept(ft) ( l./fb t ft) (o (psi ( ) (ps f) ( f) ) (ft) ) bl./( ) (psf) ) (tsf) ps ) i (o f) (psi ) (o ) ( sp (o) ps (psi
GS=0 GS=0
5.0 .5 70 1 130 4 6 6 .5 4 0 18 48.1 40.3 3848 18 5.4 75 05 43 2511 37 0. 20 30 50 18 31 5.
10.0 13.0 35 1 79 6 .3 2 24.0 34.5 1922 05 5.8 06 45 39 1469 1 4 34 1 . 16 29 25 19 86 10.0
15.0 11.8 25 6 10 469 .5 3 17.4 32.5 1393 40 4.5 324 36 06 8 41 9 . 24 36 26 18 52 15.0
20.0 11.3 11 20 5 43 3 9 14.3 31.5 45 87 4.7 830 35 1028 3 1 41 1 . 17 35 20 11 16 20.0
25.0 11.0 99 18 3 285 12.4 30.9 6 16 3.8 849 35 975 9 - - - - - 25.0
30.0 13.3 10 20 3 2613.7 31.3 98 02 3.7 593 34 1177 6 9 - - - - - 30.0
35.0 15.0 11 20 4 313 14.3 31.5 41 81 5.6 213 34 1353 0 - - - - - 35.0
40.0 17.7 12 22 2 1815.7 32.0 55 88 4.2 553 33 1201 9 7 - - - - - 40.0
GWT not present
7.2 Numerical Earth Pressure Models
The starting point for the development of an earth pressure model was numerical
simulation. Two cases were considered: 1) a typical assumption of φ' = 30° and c’ = 0 psf
that is often used in geotechnical design in North Carolina, and 2) φ' = 28° and c’ = 250 psf
representing lower bo from ria l te at e r ites. Numerical
predictions were mad nt and deflection
of a 34 feet em ation stages (the
20 feet excavation m u b he ' = ° and c’ = 0, and therefore was not used
subsequently). Figures 7.1 and 7.2 present the ulations. From figure 7.1,
the maximum bending moment on the sheet piles when φ' = 30° and c’ = 0 psf ranges
between 40,000 and 50,000 lb*ft with a maximum deflection on the order of 10 inches. In
contrast, the m um bending mo t w fi e fo ' = ° and c’ = 250 psf is
2500 – 7500 lb it to fl on g b ee 5 1 nches.
The me th of
16 feet for both te sults from the two
scenarios are s n e C pl i e li ve e the load
tests presented in Chapter 6, the me re sp ilar to that of
soil with φ' = 2 nd c’ 50 f, e T fo ws logically from the soil test results
summarized in Tables 7.1 and 7.2. The residual soils at the research si
that, on average, is greater than 30o and cohesion intercept of at least 250 psf.
7.3 Earth Pressure Model Develop ent
Rather than s
from the load tests, a ive pro h w to nsi r cl sica s using
und values t xia sts th esearch s
e using both SSI and FEM to show the bending mome
bed
axim
*ft w
asu
Sta
how
8° a
ded PZ22 sheet pile wall with 4, 8, 12, and 16 feet excav
odel as nsta le w n φ 30
results of the sim
men sho n in gur 7.2 r φ 28
h pile p de ecti ran ing etw n 0. and .5 i
red bending moments and lateral displacements at an excavation dep
sville and Monroe superimposed over the expected re
in figur 7.3. ou ed w th th qua tati obs rvations from
asu d re onse supports the behavior sim
= 2 ps or b tter. his llo
tes have fiction angle
m
et out to develop a new earth pressure model using the results generated
more effect ap ac as co de as l model
82
-10
-5
0
5
10
15
35
-60000 -40000 -20000 0 20000 40000 60000
h (ft
)
20
25
30
Moment (lb*ft)
Dep
t
-10
-5
0
5
10
35
Dep
th (f
t)
Moment (lb*ft)
15
20
25
30
-60000 -40000 -20000 0 20000 40000 60000
4ft
8f
12ft
16ft
-10
-5
Lateral Deflection (in)
SSI
SSI FEM
FEM
0
5
35
th (f
t)10
15
20
25
30
-15 -10 -5 0 5 10 15
Dep
-10
-5
Lateral Deflection (in)
0
5
10
15
(ft)
20
25
30
35
-15 -10 -5 0 5 10 15
Dep
th
Figure 7.1 - Numerical prediction of bending moment and deflection for φ′ = 30° and c’ = 0
83
-10
-5
0
5
10
15
20
25
30
35
-60000 -40000 -20000 0 20000 40000 60000
Dep
th (f
t)
Moment (lb*ft)
4ft
8f
12ft
16ft
-10
-5
0
5
10
15
20
25
30
35
-60000 -40000 -20000 0 20000 40000 60000
Dep
th (f
t)
Moment (lb*ft)
-10
-5
0
5
10
15
20
25
30
35
-15 -10 -5 0 5 10 15
Dep
th (f
t)
Lateral Deflection (in)
-10
-5
0
5
10
15
20
25
30
35
-15 -10 -5 0 5 10 15
Dep
th (f
t)
Lateral Deflection (in)
SSI
SSI FEM
FEM
igure 7.2 - Numerical prediction of bending moment and deflection for φ′ = 28 and c’ = 250 Fpsf
84
85
-10
-5
0
5
10
15
20
25
30
35
-60000 -40000 -20000 0 20000 40000 60000D
epth
(ft)
Moment (lb*ft)
Statesville Field Monroe Field FEM f = 30 c' = 0 FEM f = 28 c' = 250
-10
-5
0
5
10
15
20
25
30
35
-15 -10 -5 0 5 10 15
Dep
th (f
t)
Lateral Deflection (in)
φφ
Figure 7.3 - Numerical prediction of bending moment and deflection at 16 foot excavation with field data superimposed.
parameters appropriate for the soils encountered at the sites. Rankine coefficients are often
used in the determination of active earth pressure for many types of retaining structures.
Furthermore, in the simplest cases, Coulomb’s theory reduces to the same equations as
Rankine. Therefore, Rankine’s equation became the basis for the model.
First, using the φ and c profiles from the triaxial and BST, the Rankine earth pressure
distribution is determined for a cut 20 feet deep at both Statesville and Monroe and is plotted
in figure 7.4. These earth pressure distributions clearly show the impact of including c’ in
the analysis. The Rankine earth pressure distributions where φ′ = 30 and c’ = 0; and φ′
7.3 shows the same results in terms of the ne The resulting earth
= 28
and c’ = 250 were then superimposed on top of those in figure 7.4 to obtain figure 7.5. Table
t earth pressure force.
86
0
2
4
6
8
1
1
1
1
1
2
Dept
h (ft
)
0
2
4
6
8
0
2 40 600 800
h P re
0 00 0
Eart ressu (psf)
0
2
4
6
8
10
12
14
16
18
20
2 40 60
Earth P essure sf)
0 00 0 0 800
r (p
Dept
h (ft
)
0
2
4
6
8
10
12
14
16
18
20
0 200 400 600 800
Earth Pressure (psf)
Dept
h (ft
)
0
2
4
6
8
10
12
14
16
18
20
0 200 400 600 800
Earth Pressure (psf)
Dept
h (ft
)
)
e 7.4 – Ra kine arth ressu
(a b) (c) (d)
r n e p re distr o a n triax nd BST parameters for Statesville (a and b) and Monroe (c and d)
(
ibuti ns b sed o
ial aFigu
Figure 7.5 – Rankine earth pressure distributions based on triaxial and BST parameters for Statesville (a and b) and Monroe (c and d) o o
87
0
2
4
6
8
10
12
14
16
18
20
-500 0 500 1000
Dep
th (f
t)
Earth Pressure (psf)
Triaxial Based
phi = 28 c' = 250
phi = 30 c' = 0
0
2
4
6
8
10
12
14
16
18
20
-500 0 500 1000
Dept
h (ft
)
Earth Pressure (psf)
BST Based
phi = 28 c' = 250
phi = 30 c' = 0
0
2
4
6
8
10
12
14
16
18
20
-500 0 500 1000
Dep
th (f
t)
Earth Pressure (psf)
Triaxial Based
phi = 28 c' = 250
phi = 30 c' = 0
0
2
4
6
8
10
12
14
16
18
20
-500 0 500 1000
Dep
th (f
t)
Earth Pressure (psf)
BST Based
phi = 28 c' = 250
phi = 30 c' = 0
with superimposed predictions of distributions for cases of φ′ = 30 , c’ = 0 and φ′ = 28 , c’ = 250 psf distributions overlain
(a) (b) (c) (d)
Table 7.3 – Net earth pressure forces for cases shown in figures 7.4 and 7.5. Description Force (lbs/ft) Statesville Triaxial 2459 BST 3380 Monroe Triaxial 2093 BST 2904 Typical φ′ = 30°, c’ = 0 psf 7333 φ′ = 28°, c’ = 250 psf 3075
pressure forces for the 2-parameter Triaxial and BST based distributions fall within the
range of 2100 to 3400 lbs/ft with an average of 2800 lbs/ft. In contrast, th = ’ =
0 psf resultant is nearly twice that value.
Based on this comparison, it is proposed that the parameters from ri
and/or BST be used in the Rankine earth pressure equations for computing earth pressure
in residual soils. In the context of the load test results where little or no tr a
pressure was mobilized, these distributions would still represent a level of conservatism
while at the same time provide substantial savings to NCDOT.
7.4 Impact on Numerical Models
Using the proposed simple model, the bending moments and maxi
deflections are recomputed at an excavation depth of 16 feet for the State
Monroe sites. Figures 7.7 and 7.8 show the resulting distributions of ben m
and deflection from both the SSI and FEM models using triaxial and borehole shear test
parameters, respectively. The bending moments and deflections are far b t
obtained in figure 7.1.
7.5 Impact on Retaining Structure Design
NCDOT routinely constructs gravity, pile panel, and mechanically stabi
e φ′
the t
ue e
mum
sville
ding
elow
30°, c
axial
rth
and
oment
hose
lized
88
Moment (lb*ft) Moment (lb*ft)
-10
-5
0
10
15
-60000 -40000 -2000 40000 60000
Dep
th (f
t)
5
0 0 20000-1
20
25
30
35
Statesville
Monroe
phi =30 c' = 0
-10
-5
0
5
-15 -10 -5 0 5 10 15
10
ft)
15
20
25
30
35
Dep
th (
Lateral Deflection (in)
0
5
15
30
-60000 -40000 0000 40000 60000
Dep
th (
SSI
SSI FEM
FEM
-5
0
10
-20000 0 2
ft)
20
25
35
-10
-5
0
5
-15 -10 -5 0 5 10 15
10
h (ft
)
Lateral Deflection (in)
Figure 7.7 Numerical prediction of bending moment and deflection using triaxial
15
20
25
30
35
Dep
t
parameters.
89
-10
-5
0
5
10
15
20
25
30
35
-60000 -40000 -20000 0 20000 40000 60000
Dep
th (f
t)
Moment (lb*ft)
Statesville
Monroe
phi =30 c' = 0
-10
-5
0
5
10
15
20
25
30
35
-60000 -40000 -20000 0 20000 40000 60000
Dep
th (f
t)
Moment (lb*ft)
SSI
SSI FEM
FEM
-10
-5
0
5
10
15
20
25
30
35
-15 -10 -5 0 5 10 15
Dep
th (f
t)
Lateral Deflection (in)
-10
-5
0
5
10
15
20
25
30
35
-15 -10 -5 0 5 10 15
Dep
th (f
t)
Lateral Deflection (in)
Figure 7.8 Numerical prediction of bending moment and deflection using BST parameters
90
walls in Piedmont residual soils. To demonstrate the impacts of the proposed models on
wall design, it is convenient and simpler to focus on gravity walls. As simple case study
of a gravity wall designed and built in the Piedmont to resist sliding and overturning can
best illustrate the significant change in factors of safety for the parameters used in wall
design. Consistent with the modeled cut, a wall of 16 feet designed for cases of c’ = 0
and φ′ = 30°; c’ = 250 psf and φ′ = 28°; c’ = 0.
Starting with a factor of safety (FS) of 1.5 for overturning and sliding for the case
in which the soil parameters are c’ = 0 and φ′ = 30°, the corresponding FS for overturning
and sliding are approximately 7.8 and 4.5 respectively for the case of soil parameters of
c’ = 250 psf and φ′ = 28°. Designing to the code prescribed FS will reduce the level of
conservatism of the wall and result in substantial cost savings to NCDOT.
7.6 Parameter Selection
Key to this method is the selection of soil parameters. As previously stated the
triaxial and borehole shear tests provide two parameter measurements of shear strength
and therefore would be suitable for use with the two parameter model. Two issues then
become important: 1) the number of tests needed to determine the design value for a
structure and 2) the development of parameters from the test data.
The number of triaxial tests required to determine the representative c’ and φ
values should be based upon the length and height of structure as well as the judgment of
the geologist. It is typical practice for NCDOT to perform borings at 50 to 100 foot
intervals for the design of retaining structures. These borings extend the depth of the
retaining structure including the potential foundation. After the initial borings are
comp lete, the geologist/engineer should examine the spoon samples make note of any
91
particular changes in the character of the residual soil i.e. significant parent material
changes, marked differences in particle distribution, or other geological anomalies.
It would make the most sense then to prescribe the number of tests based upon a
criteria that includes the height and length of the wall and also reflects the judgment of
the geologist/engineer. It is therefore recommended that a three point triaxial test should
be performed for every 1000 square feet of wall
face, with minimum of two tests for any
structur
answer, thus
eters, on
average
d
boring logs, should be
used to
has
he
e. Additional tests may be recommended by the geologist/engineer based upon
significant changes in stratigraphy. Additional specifications on performing triaxial tests
on residual specimens are included as Appendix E.
The BST provides an insitu measure of φ and c as well. However, questions
regarding drainage and whether there are unsaturated effects are difficult to
the parameters may or may not represent effective conditions. The param
compare well with the triaxial tests having slightly higher friction angles and
lower c. While the triaxial test appears to be the standard, the BST may be recommende
to complement the triaxial tests, or in exceptional cases, substitute for it.
Being that the tests represent discrete points within the soil mass, the SPT
blowcounts, along with the descriptive information included in the
verify that the individual tests are representative.
When interpreting the results of the tests, the strength should be determined based
upon application of the Mohr Coulomb failure criteria. Historically, the c component
been neglected by arbitrarily drawing the failure envelope through the origin. Use of a
tool such as a spreadsheet that automatically determines the angle and intercept of t
Mohr-coulomb envelope is suggested.
92
CHAPTER 8 CONCLUSION
8.1 Research Summary
n the results of the research program:
1 ont
sion intercept
l earth pressure detected during
The objective of this research project was to develop an earth pressure model for
retaining structures in Piedmont residual soils. The scope of work included literature
review, numerical models, insitu and laboratory soil testing, and full scale load testing of
cantilever sheet pile retaining structures.
8.2 Conclusions
The following statements are made based upo
) The earth pressure currently used in design of retaining structures in Piedm
soils is greater than earth pressure measured during load tests. Field
measurements from the instrumented wall load tests demonstrated that the
retained soils exerted little or no pressure on the structure.
2) The Piedmont soils that were tested in this research had significant strength.
The average friction angle was 28o and the average drained cohe
was above 300 psf. These values were consistent with those found in the
literature for similar soils.
3) Based on the soil test results as well as the minima
the load tests, the soil strength parameters, φ and c’ should be used together in
Rankine’s earth pressure equation to predict the earth pressure in Piedmont soils.
4) Triaxial tests provided the most consistent measurement of φ′ and c’. The
borehole shear tests also measure φ and c but should only be used when triaxial
testing is unavailable.
93
8.3 Recommendations
Section 2.3 listed the limitations of this study. In order to address these, two
primary recommendations for future work are put forward. First, it is suggested that a
same time provide a statistical context for the
development of a standard.
p A follow
u ars
w terials. Within the
c
p
broader database of triaxial tests in Piedmont soils be developed. This data can be
generated from historic test records as well as tests performed based on the
recommendations of this study. This database will give NCDOT engineers confidence in
the parameters, especially c’, while at the
parameters for the
Second, the long term behavior of these soils must be examined. The constraints
of this study were such that the retaining structures could not remain for a long enough
eriod to evaluate the time-dependency of the φ and c’ components of the soil.
p study using an instrumented retaining structure or cut, that can stand for several ye
ould, at the minimum, help develop trends in the behavior of the ma
study, soils would be tested at the outset of the project, and yearly to determine the
hanges in parameters, while the wall or cut is continuously monitored for changes in
ore air and water pressures, movements or failures.
94
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99
APPENDIX A STATESVILLE SOIL TEST DATA
(included as separate attachment)
100
MONROE SOIL TEST DATA
(included as separate attachment)
APPENDIX B
101
STRAIN GAGE DATA
APPENDIX C
102
Table C.1 - St a - Statesville
rain gage set dat
Pile #1Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με (set) age Frequency (Hz) Microstrain (με) Δ με (set)
1 1500.000 0.000 9 4500.000 0.0002 1500.000 0.000 10 4500.000 0.0003 1500.000 0.000 11 4500.000 0.0004 1500.000 0.000 12 4500.000 0.0005 1500.000 0.000 13 4500.000 0.0006 1500.000 0.000 14 4500.000 0.0007 1500.000 0.000 15 4500.000 0.0008 1500.000 0.000 16 4500.000 0.000
Pile #2Tension C pressionGage Frequency (Hz) Microstrain (με) Δ με (set) G e Frequency (Hz) Microstrain (με) Δ με (set)
17 1500.000 0.000 25 4500.000 0.00018 1500.000 0.000 26 4500.000 0.00019 1500.000 0.000 2720 1500.000 0.000 28 4500.000 0.00021 1500.000 0.000 29 4500.000 0.00022 1500.000 0.000 30 4500.000 0.00023 1500.000 0.000 31 4500.000 0.00024 1500.000 0.000 32 4500.000 0.000
Pile #3Compression T ionGage Frequency (Hz) Microstrain (με) Δ με (set) G e Frequency (Hz) Microstrain (με) Δ με (set)
33 4500.000 0.000 41 1500.000 0.00034 4500.000 0.000 42 1500.000 0.00035 4500.000 0.000 43 1500.000 0.00036 4500.000 0.000 44 1500.000 0.00037 4500.000 0.000 45 1500.000 0.00038 4500.000 0.000 46 1500.000 0.00039 4500.000 0.000 47 1500.000 0.00040 4500.000 0.000 48 1500.000 0.000
Pile #4 Compression T ionGage Frequency (Hz) Microstrain (με) Δ με (set) G e Frequency (Hz) Microstrain (με) Δ με (set)
49 4500.000 0.000 57 1500.000 0.00050 58 1500.000 0.00051 59 1500.000 0.00052 60 1500.000 0.00053 4500.000 0.000 61 1500.000 0.00054 4500.000 0.000 62 1500.000 0.00055 4500.000 0.000 63 1500.000 0.00056 64 1500.000 0.000
G
omag
ensag
ensag
103
Tab lle le C.2 - Post driving strain readings - Statesvi Pile #1Tension SW CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
1 602.2 1473.063 -26.937 9 1029 4301.012 -198.9882 551.1 1233.675 -266.325 10 1058 4546.857 46.8573 574.9 1342.532 -157.468 11 1056 4529.682 29.6824 563.2 1288.443 -211.557 12 1038 4376.578 -123.4225 494.3 992.479 -507.521 13 1060 4564.063 64.0636 523.9 1114.902 -385.098 14 1065 4607.222 107.2227 480.4 937.445 -562.555 15 1078 4720.385 220.3858 467.2 886.636 -613.364 16 1040 4393.459 -106.541
Pile #2Tension SW CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
17 580.1 1366.928 -133.072 25 1040 4393.459 -106.54118 580.9 1370.701 -129.299 26 1062 4581.302 81.30219 557.9 1264.307 -235.693 2720 659.2 1765.120 265.120 28 1040 4393.459 -106.54121 579.9 1365.986 -134.014 29 1058 4546.857 46.85722 487.6 965.756 -534.244 30 1064 4598.574 98.57423 531.6 1147.915 -352.085 31 1081 4746.695 246.69524 470.3 898.442 -601.558 32 1059 4555.456 55.456
Pile #3CompressiSE TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
33 1029 4301.012 -198.988 41 603.4 1478.940 -21.06034 1060 4564.063 64.063 42 581.2 1372.117 -127.88335 1028 4292.657 -207.343 43 631.2 1618.355 118.35536 1009 4135.445 -364.555 44 647.9 1705.124 205.12437 1029 4301.012 -198.988 45 591.2 1419.740 -80.26038 1031 4317.748 -182.252 46 626.5 1594.344 94.34439 1028 4292.657 -207.343 47 693.5 1953.587 453.58740 1024 4259.316 -240.684 48 611.6 1519.410 19.410
Pile #4 CompressiNE TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
49 1073 4676.698 176.698 57 583.7 1383.947 -116.05350 58 650.5 1718.836 218.83651 59 701.9 2001.200 501.20052 60 534.1 1158.738 -341.26253 1032 4326.127 -173.873 61 763.5 2367.871 867.87154 1016 4193.024 -306.976 62 733.9 2187.831 687.83155 1136 5241.995 741.995 63 593.7 1431.773 -68.22756 64 517.3 1086.988 -413.012
104
Table C.3 - Strain after 4 foot excavation - Statesville
Pile #1Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
1 602.0 1472.085 -27.915 9 1029 4301.012 -198.9882 556.6 1258.422 -241.578 10 1057 4538.265 38.2653 578.8 1360.808 -139.192 11 1056 4529.682 29.6824 564.8 1295.774 -204.226 12 1038 4376.578 -123.4225 496.1 999.720 -500.280 13 1061 4572.679 72.6796 525.4 1121.295 -378.705 14 1066 4615.878 115.8787 482.8 946.835 -553.165 15 1079 4729.147 229.1478 470.1 897.678 -602.322 16 1041 4401.912 -98.088
Pile #2Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
17 575.7 1346.271 -153.729 25 1042 4410.373 -89.62718 578.7 1360.338 -139.662 26 1064 4598.574 98.57419 561.2 1279.308 -220.692 2720 661.6 1777.997 277.997 28 1039 4385.014 -114.98621 582.0 1375.897 -124.103 29 1059 4555.456 55.45622 490.5 977.278 -522.722 30 1065 4607.222 107.22223 534.8 1161.777 -338.223 31 1083 4764.275 264.27524 475.0 916.489 -583.511 32 1060 4564.063 64.063
Pile #3Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
33 1034.0 4342.912 -157.088 41 617.5 1548.866 48.86634 1060.0 4564.063 64.063 42 584.2 1386.319 -113.68135 1026.0 4275.970 -224.030 43 639.7 1662.236 162.23636 1008.0 4127.252 -372.748 44 658.0 1758.700 258.70037 1029.0 4301.012 -198.988 45 593.9 1432.737 -67.26338 1032.0 4326.127 -173.873 46 627.3 1598.418 98.41839 1050.0 4478.355 -21.645 47 691.2 1940.651 440.65140 1025.0 4267.639 -232.361 48 612.3 1522.890 22.890
Pile #4Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
49 1077 4711.632 211.632 57 591.4 1420.701 -79.29950 58 649.9 1715.667 215.66751 59 705.4 2021.207 521.20752 60 538.8 1179.221 -320.77953 1032 4326.127 -173.873 61 764.4 2373.456 873.45654 1017 4201.282 -298.718 62 731.7 2174.733 674.73355 1136 5241.995 741.995 63 594.7 1436.600 -63.40056 64 518.2 1090.774 -409.226
105
Table C.4 - Strain after 8 foot excavation - Statesville
Pile #1Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
1 599.8 1461.345 -38.655 9 1029 4301.012 -198.9882 562.9 1287.071 -212.929 10 1053 4503.982 3.9823 608.1 1502.069 2.069 11 1050 4478.355 -21.6454 578.7 1360.338 -139.662 12 1037 4368.149 -131.8515 504.2 1032.632 -467.368 13 1062 4581.302 81.3026 528.8 1135.855 -364.145 14 1068 4633.215 133.2157 485.9 959.033 -540.967 15 1080 4737.917 237.9178 472.2 905.716 -594.284 16 1042 4410.373 -89.627
Pile #2Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
17 579.0 1361.749 -138.251 25 1040 4393.459 -106.54118 582.5 1378.262 -121.738 26 1064 4598.574 98.57419 549.5 1226.522 -273.478 2720 669.8 1822.343 322.343 28 1036 4359.728 -140.27221 591.8 1422.623 -77.377 29 1055 4521.108 21.10822 496.7 1002.140 -497.860 30 1065 4607.222 107.22223 538.4 1177.470 -322.530 31 1084 4773.077 273.07724 479.1 932.379 -567.621 32 1060 4564.063 64.063
Pile #3Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
33 1031.0 4317.748 -182.252 41 607.7 1500.094 0.09434 1060.0 4564.063 64.063 42 584.6 1388.218 -111.78235 1028.0 4292.657 -207.343 43 629.0 1607.094 107.09436 1007.0 4119.067 -380.933 44 664.8 1795.238 295.23837 1024.0 4259.316 -240.684 45 609.0 1506.519 6.51938 1030.0 4309.376 -190.624 46 632.7 1626.056 126.05639 1029.0 4301.012 -198.988 47 691.3 1941.212 441.21240 1025.0 4267.639 -232.361 48 612.2 1522.392 22.392
Pile #4Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
49 1074 4685.420 185.420 57 588.3 1405.846 -94.15450 58 648.0 1705.650 205.65051 59 697.3 1975.055 475.05552 60 552.3 1239.053 -260.94753 1026 4275.970 -224.030 61 771.5 2417.752 917.75254 1014 4176.532 -323.468 62 732.8 2181.277 681.27755 1136 5241.995 741.995 63 595.1 1438.533 -61.46756 64 518.7 1092.880 -407.120
106
Table C.5 - Strain after 12 foot excavation - Statesville
Pile #1Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
1 598.0 1452.587 -47.413 9 1032 4326.127 -173.8732 566.5 1303.586 -196.414 10 1056 4529.682 29.6823 591.4 1420.701 -79.299 11 1052 4495.432 -4.5684 610.6 1514.445 14.445 12 1025 4267.639 -232.3615 536.3 1168.303 -331.697 13 1054 4512.541 12.5416 540.9 1188.431 -311.569 14 1068 4633.215 133.2157 494.4 992.880 -507.120 15 1083 4764.275 264.2758 478.1 928.490 -571.510 16 1043 4418.843 -81.157
Pile #2Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
17 577.4 1354.233 -145.767 25 1041 4401.912 -98.08818 586.4 1396.779 -103.221 26 1064 4598.574 98.57419 554.1 1247.143 -252.857 2720 658.1 1759.234 259.234 28 1038 4376.578 -123.42221 610.6 1514.445 14.445 29 1047 4452.801 -47.19922 522.5 1108.951 -391.049 30 1055 4521.108 21.10823 554.0 1246.693 -253.307 31 1081 4746.695 246.69524 489.2 972.104 -527.896 32 1060 4564.063 64.063
Pile #3Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
33 1030.0 4309.376 -190.624 41 605.4 1488.760 -11.24034 1059.0 4555.456 55.456 42 585.9 1394.399 -105.60135 1031.0 4317.748 -182.252 43 627.0 1596.890 96.89036 1017.0 4201.282 -298.718 44 639.7 1662.236 162.23637 1023.0 4251.001 -248.999 45 609.9 1510.975 10.97538 1018.0 4209.548 -290.452 46 668.1 1813.105 313.10539 1022.0 4242.694 -257.306 47 706.6 2028.090 528.09040 1024.0 4259.316 -240.684 48 612.9 1525.876 25.876
Pile #4Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
49 1073 4676.698 176.698 57 582.8 1379.682 -120.31850 58 646.5 1697.763 197.76351 59 704.5 2016.053 516.05352 60 549.4 1226.076 -273.92453 1020 4226.105 -273.895 61 780.7 2475.758 975.75854 999 4053.880 -446.120 62 746.5 2263.599 763.59955 1129 5177.592 677.592 63 601.8 1471.107 -28.89356 64 519.9 1097.942 -402.058
107
Table C.6 - Strain after 16 foot excavation - Statesville
Pile #1Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
1 594.4 1435.151 -64.849 9 1035 4351.316 -148.6842 562.4 1284.785 -215.215 10 1061 4572.679 72.6793 584.5 1387.743 -112.257 11 1057 4538.265 38.2654 603.5 1479.430 -20.570 12 1027 4284.309 -215.6915 578.7 1360.338 -139.662 13 1041 4401.912 -98.0886 567.6 1308.654 -191.346 14 1062 4581.302 81.3027 513.9 1072.747 -427.253 15 1083 4764.275 264.2758 487.0 963.380 -536.620 16 1046 4444.299 -55.701
Pile #2Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
17 573.9 1337.865 -162.135 25 1043 4418.843 -81.15718 581.2 1372.117 -127.883 26 1065 4607.222 107.22219 548.9 1223.845 -276.155 2720 658.2 1759.769 259.769 28 1038 4376.578 -123.42221 595.8 1441.919 -58.081 29 1048 4461.311 -38.68922 556.6 1258.422 -241.578 30 1042 4410.373 -89.62723 595.5 1440.467 -59.533 31 1065 4607.222 107.22224 507.8 1047.431 -452.569 32 1054 4512.541 12.541
Pile #3Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
33 1031.0 4317.748 -182.252 41 604.7 1485.319 -14.68134 1059.0 4555.456 55.456 42 585.3 1391.544 -108.45635 1030.0 4309.376 -190.624 43 626.4 1593.835 93.83536 1019.0 4217.822 -282.178 44 634.9 1637.384 137.38437 1035.0 4351.316 -148.684 45 591.1 1419.260 -80.74038 1031.0 4317.748 -182.252 46 659.4 1766.192 266.19239 1022.0 4242.694 -257.306 47 736.0 2200.369 700.36940 1019.0 4217.822 -282.178 48 628.0 1601.988 101.988
Pile #4Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
49 1076 4702.886 202.886 57 578.8 1360.808 -139.19250 58 645.7 1693.564 193.56451 59 701.0 1996.071 496.07152 60 548.7 1222.953 -277.04753 1030 4309.376 -190.624 61 769.1 2402.733 902.73354 1002 4078.264 -421.736 62 742.7 2240.613 740.61355 1119 5086.278 586.278 63 613.2 1527.370 27.37056 64 530.4 1142.739 -357.261
108
Table C.7 - Strain after 20 foot excavation - Statesville
Pile #1Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
1 595.5 1440.467 -59.533 9 1041 4401.912 -98.0882 565.4 1298.529 -201.471 10 1068 4633.215 133.2153 581.8 1374.951 -125.049 11 1065 4607.222 107.2224 601.3 1468.664 -31.336 12 1033 4334.516 -165.4845 581.4 1373.061 -126.939 13 1042 4410.373 -89.6276 596.5 1445.309 -54.691 14 1047 4452.801 -47.1997 544.3 1203.418 -296.582 15 1173 5589.024 1089.0248 500.1 1015.906 -484.094 16 1045 4435.806 -64.194
Pile #2Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
17 570.4 1321.597 -178.403 25 1045 4435.806 -64.19418 578.3 1358.458 -141.542 26 1069 4641.895 141.89519 545.9 1210.504 -289.496 2720 651.6 1724.654 224.654 28 1043 4418.843 -81.15721 591.9 1423.104 -76.896 29 1052 4495.432 -4.56822 543.7 1200.767 -299.233 30 1046 4444.299 -55.70123 612.9 1525.876 25.876 31 1057 4538.265 38.26524 531.9 1149.211 -350.789 32 1047 4452.801 -47.199
Pile #3Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
33 1034.0 4342.912 -157.088 41 615.1 1536.850 36.85034 1061.0 4572.679 72.679 42 585.9 1394.399 -105.60135 1032.0 4326.127 -173.873 43 628.8 1606.072 106.07236 1023.0 4251.001 -248.999 44 629.1 1607.605 107.60537 1045.0 4435.806 -64.194 45 578.4 1358.928 -141.07238 1056.0 4529.682 29.682 46 626.9 1596.381 96.38139 1047.0 4452.801 -47.199 47 700.9 1995.501 495.50140 1022.0 4242.694 -257.306 48 645.6 1693.039 193.039
Pile #4Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
49 1078 4720.385 220.385 57 581.4 1373.061 -126.93950 58 644.0 1684.658 184.65851 59 699.5 1987.538 487.53852 60 546.9 1214.943 -285.05753 1039 4385.014 -114.986 61 761.3 2354.245 854.24554 1030 4309.376 -190.624 62 708.0 2036.134 536.13455 1140 5278.975 778.975 63 576.6 1350.483 -149.51756 64 534.0 1158.304 -341.696
109
Table C.8 - Strain gage set data - Monroe Pile #1 (SE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με (set) Gage Frequency (Hz) Microstrain (με) Δ με (set)
1 1500.000 0.000 9 4500.000 0.00065 1500.000 0.000 722 1500.000 0.000 10 4500.000 0.000
66 1500.000 0.000 73 4500.000 0.0003 1500.000 0.000 11 4500.000 0.000
67 1500.000 0.000 74 4500.000 0.0004 1500.000 0.000 12 4500.000 0.000
68 1500.000 0.000 75 4500.000 0.0005 13 4500.000 0.000
69 1500.000 0.000 76 4500.000 0.0006 1500.000 0.000 14 4500.000 0.000
70 77 4500.000 0.0007 1500.000 0.000 15 4500.000 0.000
71 1500.000 0.000 78 4500.000 0.0008 1500.000 0.000 16 4500.000 0.000
Pile #2 (NE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με (set) Gage Frequency (Hz) Microstrain (με) Δ με (set)
17 1500.000 0.000 25 4500.000 0.00079 1500.000 0.000 86 4500.000 0.00018 1500.000 0.000 26 4500.000 0.00080 1500.000 0.000 87 4500.000 0.00019 1500.000 0.000 27 4500.000 0.00081 1500.000 0.000 88 4500.000 0.00020 1500.000 0.000 28 4500.000 0.00082 89 4500.000 0.00021 2983 1500.000 0.000 90 4500.000 0.00022 1500.000 0.000 30 4500.000 0.00084 1500.000 0.000 91 4500.000 0.00023 1500.000 0.000 31 4500.000 0.00085 1500.000 0.000 92 4500.000 0.00024 1500.000 0.000 32 4500.000 0.000
Pile #3 (SW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με (set) Gage Frequency (Hz) Microstrain (με) Δ με (set)
33 4500.000 0.000 41 1500.000 0.00093 4500.000 0.000 100 1500.000 0.00034 4500.000 0.000 4294 4500.000 0.000 101 1500.000 0.00035 4500.000 0.000 43 1500.000 0.00095 4500.000 0.000 102 1500.000 0.00036 4500.000 0.000 44 1500.000 0.00096 4500.000 0.000 103 1500.000 0.00037 4500.000 0.000 45 1500.000 0.00097 4500.000 0.000 104 1500.000 0.00038 4500.000 0.000 46 1500.000 0.00098 4500.000 0.000 105 1500.000 0.00039 4500.000 0.000 47 1500.000 0.00099 4500.000 0.000 106 1500.000 0.00040 48 1500.000 0.000
Pile #4 (NW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με (set) Gage Frequency (Hz) Microstrain (με) Δ με (set)
49 4500.000 0.000 57 1500.000 0.000107 11450 58 1500.000 0.000
108 4500.000 0.000 11551 59 1500.000 0.000
109 4500.000 0.000 116 1500.000 0.00052 60 1500.000 0.000
110 4500.000 0.000 117 1500.000 0.00053 4500.000 0.000 61
111 4500.000 0.000 118 1500.000 0.00054 4500.000 0.000 62
112 4500.000 0.000 119 1500.000 0.00055 4500.000 0.000 63 1500.000 0.000
113 4500.000 0.000 120 1500.000 0.00056 4500.000 0.000 64 1500.000 0.000
110
Tab oe le C.9 - Post driving strain readings - MonrPile #1 (SE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με (set) Gage Frequency (Hz) Microstrain (με) Δ με (set)
1 605.9 1491.220 -8.780 9 1022 4242.694 -257.30665 610.2 1512.461 12.461 722 585.4 1392.020 -107.980 10 1046 4444.299 -55.701
66 650.6 1719.365 219.365 73 1055 4521.108 21.1083 588.2 1405.368 -94.632 11 1041 4401.912 -98.088
67 659.4 1766.192 266.192 74 1036 4359.728 -140.2724 542.2 1194.150 -305.850 12 1027 4284.309 -215.691
68 579.1 1362.219 -137.781 75 1047 4452.801 -47.1995 13 1026 4275.970 -224.030
69 581.2 1372.117 -127.883 76 1053 4503.982 3.9826 585.9 1394.399 -105.601 14 1033 4334.516 -165.484
70 77 1044 4427.320 -72.6807 608.4 1503.552 3.552 15 1048 4461.311 -38.689
71 677.6 1865.034 365.034 78 1040 4393.459 -106.5418 624.0 1581.645 81.645 16 1032 4326.127 -173.873
Pile #2 (NE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με (set) Gage Frequency (Hz) Microstrain (με) Δ με (set)
17 528.3 1133.708 -366.292 25 1015 4184.774 -315.22679 620.7 1564.961 64.961 86 1070 4650.584 150.58418 616.4 1543.353 43.353 26 1064 4598.574 98.57480 598.9 1456.963 -43.037 87 1041 4401.912 -98.08819 613.4 1528.366 28.366 27 1002 4078.264 -421.73681 522.2 1107.678 -392.322 88 1061 4572.679 72.67920 521.0 1102.593 -397.407 28 1001 4070.128 -429.87282 89 1019 4217.822 -282.17821 2983 613.1 1526.872 26.872 90 1018 4209.548 -290.45222 536.8 1170.483 -329.517 30 1023 4251.001 -248.99984 674.0 1845.269 345.269 91 1060 4564.063 64.06323 637.6 1651.340 151.340 31 1079 4729.147 229.14785 663.8 1789.841 289.841 92 1038 4376.578 -123.42224 584.3 1386.793 -113.207 32 1026 4275.970 -224.030
Pile #3 (SW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με (set) Gage Frequency (Hz) Microstrain (με) Δ με (set)
33 1034 4342.912 -157.088 41 601.4 1469.152 -30.84893 1037 4368.149 -131.851 100 636.5 1645.647 145.64734 1053 4503.982 3.982 4294 998 4045.768 -454.232 101 611.5 1518.913 18.91335 999 4053.880 -446.120 43 662.8 1784.452 284.45295 1010 4143.646 -356.354 102 649.1 1711.446 211.44636 1015 4184.774 -315.226 44 599.0 1457.450 -42.55096 1024 4259.316 -240.684 103 690.7 1937.844 437.84437 1039 4385.014 -114.986 45 583.7 1383.947 -116.05397 1070 4650.584 150.584 104 662.4 1782.299 282.29938 1057 4538.265 38.265 46 571.1 1324.842 -175.15898 1043 4418.843 -81.157 105 641.3 1670.561 170.56139 1077 4711.632 211.632 47 699.2 1985.833 485.83399 1052 4495.432 -4.568 106 656.4 1750.157 250.15740 48 626.0 1591.800 91.800
Pile #4 (NW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με (set) Gage Frequency (Hz) Microstrain (με) Δ με (set)
49 1054 4512.541 12.541 57 562.9 1287.071 -212.929107 11450 58 673.3 1841.438 341.438
108 1017 4201.282 -298.718 11551 59 722.4 2119.802 619.802
109 999 4053.880 -446.120 116 693.2 1951.898 451.89852 60 723.8 2128.027 628.027
110 998 4045.768 -454.232 117 710.6 2051.116 551.11653 1010 4143.646 -356.354 61
111 1005 4102.722 -397.278 118 717.9 2093.475 593.47554 1015 4184.774 -315.226 62
112 1019 4217.822 -282.178 119 700.9 1995.501 495.50155 1066 4615.878 115.878 63 683.9 1899.875 399.875
113 1064 4598.574 98.574 120 662.2 1781.223 281.22356 1042 4410.373 -89.627 64 721.6 2115.110 615.110
111
T able C.10 - Strain after 4 foot excavation - MonroePile #1 (SE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
1 609.7 1509.984 9.984 9 1024 4259.316 -240.68465 608.6 1504.540 4.540 722 584.6 1388.218 -111.782 10 1044 4427.320 -72.680
66 652.9 1731.543 231.543 73 1054 4512.541 12.5413 590.0 1413.982 -86.018 11 1042 4410.373 -89.627
67 658.9 1763.514 263.514 74 1037 4368.149 -131.8514 542.4 1195.031 -304.969 12 1027 4284.309 -215.691
68 579.6 1364.573 -135.427 75 1048 4461.311 -38.6895 13 1027 4284.309 -215.691
69 582.3 1377.316 -122.684 76 1054 4512.541 12.5416 586.6 1397.732 -102.268 14 1034 4342.912 -157.088
70 77 1045 4435.806 -64.1947 608.3 1503.057 3.057 15 1049 4469.829 -30.171
71 677.8 1866.135 366.135 78 1041 4401.912 -98.0888 624.3 1583.166 83.166 16 1033 4334.516 -165.484
Pile #2 (NE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
17 527.6 1130.705 -369.295 25 1014 4176.532 -323.46879 624.3 1583.166 83.166 86 1069 4641.895 141.89518 618.1 1551.877 51.877 26 1065 4607.222 107.22280 599.4 1459.397 -40.603 87 1042 4410.373 -89.62719 614.2 1532.356 32.356 27 1004 4094.561 -405.43981 523.2 1111.925 -388.075 88 1062 4581.302 81.30220 521.1 1103.017 -396.983 28 1002 4078.264 -421.73682 89 1020 4226.105 -273.89521 2983 614.4 1533.354 33.354 90 1019 4217.822 -282.17822 538.7 1178.783 -321.217 30 1024 4259.316 -240.68484 675.4 1852.943 352.943 91 1060 4564.063 64.06323 639.4 1660.677 160.677 31 1080 4737.917 237.91785 665.1 1796.858 296.858 92 1039 4385.014 -114.98624 585.7 1393.447 -106.553 32 1026 4275.970 -224.030
Pile #3 (SW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
33 1034.0 4342.912 -157.088 41 604.4 1483.846 -16.15493 1037.0 4368.149 -131.851 100 637.5 1650.822 150.82234 1052.0 4495.432 -4.568 4294 998.0 4045.768 -454.232 101 610.9 1515.934 15.93435 999.0 4053.880 -446.120 43 662.4 1782.299 282.29995 1011.0 4151.856 -348.144 102 648.1 1706.177 206.17736 1017.0 4201.282 -298.718 44 598.8 1456.477 -43.52396 1025 4267.639 -232.361 103 691.0 1939.528 439.52837 1040 4393.459 -106.541 45 583.6 1383.472 -116.52897 1071 4659.281 159.281 104 663.1 1786.068 286.06838 1057 4538.265 38.265 46 572.4 1330.881 -169.11998 1044 4427.320 -72.680 105 642.5 1676.819 176.81939 1078 4720.385 220.385 47 703.6 2010.905 510.90599 1053 4503.982 3.982 106 657.8 1757.631 257.63140 48 627.1 1597.399 97.399
Pile #4 (NW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
49 1055 4521.108 21.108 57 564.4 1293.939 -206.061107 11450 58 675.0 1850.749 350.749
108 1016 4193.024 -306.976 11551 59 723.2 2124.500 624.500
109 999 4053.880 -446.120 116 693.9 1955.842 455.84252 60 724.6 2132.733 632.733
110 999 4053.880 -446.120 117 711.8 2058.050 558.05053 1011 4151.856 -348.144 61
111 1007 4119.067 -380.933 118 719.3 2101.648 601.64854 1016 4193.024 -306.976 62
112 1020 4226.105 -273.895 119 702.6 2005.193 505.19355 1067 4624.542 124.542 63 685.7 1909.889 409.889
113 1065 4607.222 107.222 120 663.8 1789.841 289.84156 1043 4418.843 -81.157 64 722.9 2122.738 622.738
112
Table C.11 - Strain after 8 foot excavation - Monroe Pile #1 (SE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
1 610.3 1512.957 12.957 9 1022 4242.694 -257.30665 610.8 1515.437 15.437 722 581.6 1374.006 -125.994 10 1047 4452.801 -47.199
66 638.2 1654.450 154.450 73 1057 4538.265 38.2653 570.6 1322.524 -177.476 11 1045 4435.806 -64.194
67 631.5 1619.894 119.894 74 1036 4359.728 -140.2724 536.0 1166.996 -333.004 12 1022 4242.694 -257.306
68 580.1 1366.928 -133.072 75 1044 4427.320 -72.6805 13 1024 4259.316 -240.684
69 581.1 1371.645 -128.355 76 1052 4495.432 -4.5686 584.2 1386.319 -113.681 14 1032 4326.127 -173.873
70 77 1044 4427.320 -72.6807 605.3 1488.268 -11.732 15 1047 4452.801 -47.199
71 676.0 1856.237 356.237 78 1040 4393.459 -106.5418 622.9 1576.074 76.074 16 1032 4326.127 -173.873
Pile #2 (NE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
17 531.9 1149.211 -350.789 25 1013 4168.298 -331.70279 628.2 1603.008 103.008 86 1068 4633.215 133.21518 620.5 1563.952 63.952 26 1066 4615.878 115.87880 600.5 1464.758 -35.242 87 1040 4393.459 -106.54119 617.1 1546.860 46.860 27 1001 4070.128 -429.87281 532.0 1149.643 -350.357 88 1058 4546.857 46.85720 523.1 1111.500 -388.500 28 1001 4070.128 -429.87282 89 1021 4234.395 -265.60521 2983 613.8 1530.360 30.360 90 1020 4226.105 -273.89522 537.0 1171.355 -328.645 30 1024 4259.316 -240.68484 673.5 1842.532 342.532 91 1060 4564.063 64.06323 638.5 1656.005 156.005 31 1080 4737.917 237.91785 664.9 1795.778 295.778 92 1039 4385.014 -114.98624 585.6 1392.971 -107.029 32 1026 4275.970 -224.030
Pile #3 (SW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
33 1034 4342.912 -157.088 41 603.2 1477.960 -22.04093 1037 4368.149 -131.851 100 635.4 1639.964 139.96434 1056 4529.682 29.682 4294 1002 4078.264 -421.736 101 594.9 1437.566 -62.43435 1002 4078.264 -421.736 43 654.6 1740.572 240.57295 1010 4143.646 -356.354 102 645.6 1693.039 193.03936 1015 4184.774 -315.226 44 599.8 1461.345 -38.65596 1024 4259.316 -240.684 103 689.6 1931.677 431.67737 1039 4385.014 -114.986 45 578.9 1361.279 -138.72197 1070 4650.584 150.584 104 661.2 1775.847 275.84738 1057 4538.265 38.265 46 571.1 1324.842 -175.15898 1043 4418.843 -81.157 105 641.5 1671.603 171.60339 1078 4720.385 220.385 47 700.3 1992.086 492.08699 1053 4503.982 3.982 106 657.5 1756.028 256.02840 48 627.0 1596.890 96.890
Pile #4 (NW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
49 1054 4512.541 12.541 57 566.0 1301.286 -198.714107 11450 58 672.9 1839.251 339.251
108 1014 4176.532 -323.468 11551 59 729.5 2161.676 661.676
109 993 4005.331 -494.669 116 698.5 1981.859 481.85952 60 726.6 2144.523 644.523
110 997 4037.665 -462.335 117 712.3 2060.942 560.94253 1010 4143.646 -356.354 61
111 1007 4119.067 -380.933 118 718.9 2099.312 599.31254 1017 4201.282 -298.718 62
112 1020 4226.105 -273.895 119 702.5 2004.622 504.62255 1068 4633.215 133.215 63 686.0 1911.561 411.561
113 1066 4615.878 115.878 120 664.1 1791.459 291.45956 1043 4418.843 -81.157 64 723.4 2125.675 625.675
113
Table C.12 - Strain after 12 foot excavation - Monroe Pile #1 (SE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
1 605.7 1490.236 -9.764 9 1019 4217.822 -282.17865 611.4 1518.416 18.416 722 582.4 1377.789 -122.211 10 1045 4435.806 -64.194
66 639.2 1659.638 159.638 73 1057 4538.265 38.2653 570.6 1322.524 -177.476 11 1046 4444.299 -55.701
67 631.3 1618.868 118.868 74 1036 4359.728 -140.2724 536.2 1167.867 -332.133 12 1022 4242.694 -257.306
68 580.4 1368.342 -131.658 75 1044 4427.320 -72.6805 13 1024 4259.316 -240.684
69 581.0 1371.173 -128.827 76 1052 4495.432 -4.5686 584.1 1385.844 -114.156 14 1032 4326.127 -173.873
70 77 1044 4427.320 -72.6807 605.1 1487.285 -12.715 15 1047 4452.801 -47.199
71 675.9 1855.687 355.687 78 1040 4393.459 -106.5418 622.8 1575.568 75.568 16 1032 4326.127 -173.873
Pile #2 (NE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
17 532.5 1151.805 -348.195 25 1012 4160.073 -339.92779 629.3 1608.627 108.627 86 1069 4641.895 141.89518 621.4 1568.492 68.492 26 1066 4615.878 115.87880 602.7 1475.510 -24.490 87 1038 4376.578 -123.42219 614.1 1531.857 31.857 27 998 4045.768 -454.23281 538.6 1178.345 -321.655 88 1054 4512.541 12.54120 525.9 1123.431 -376.569 28 998 4045.768 -454.23282 89 1018 4209.548 -290.45221 2983 613.2 1527.370 27.370 90 1019 4217.822 -282.17822 535.5 1164.820 -335.180 30 1024 4259.316 -240.68484 671.5 1831.606 331.606 91 1060 4564.063 64.06323 637.4 1650.304 150.304 31 1080 4737.917 237.91785 664.3 1792.538 292.538 92 1039 4385.014 -114.98624 585.3 1391.544 -108.456 32 1026 4275.970 -224.030
Pile #3 (SW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
33 1032 4326.127 -173.873 41 603.0 1476.980 -23.02093 1037 4368.149 -131.851 100 635.9 1642.546 142.54634 1056 4529.682 29.682 4294 1003 4086.409 -413.591 101 593.4 1430.326 -69.67435 1003 4086.409 -413.591 43 653.5 1734.727 234.72795 1010 4143.646 -356.354 102 645.2 1690.942 190.94236 1015 4184.774 -315.226 44 599.8 1461.345 -38.65596 1023 4251.001 -248.999 103 689.3 1929.996 429.99637 1038 4376.578 -123.422 45 578.2 1357.989 -142.01197 1070 4650.584 150.584 104 660.6 1772.626 272.62638 1056 4529.682 29.682 46 570.5 1322.060 -177.94098 1043 4418.843 -81.157 105 640.9 1668.478 168.47839 1078 4720.385 220.385 47 699.8 1989.243 489.24399 1053 4503.982 3.982 106 657.2 1754.426 254.42640 48 626.8 1595.871 95.871
Pile #4 (NW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
49 1052 4495.432 -4.568 57 563.2 1288.443 -211.557107 11450 58 671.6 1832.151 332.151
108 1012 4160.073 -339.927 11551 59 733.7 2186.638 686.638
109 987 3957.074 -542.926 116 715.0 2076.596 576.59652 60 739.3 2220.145 720.145
110 991 3989.213 -510.787 117 715.7 2080.664 580.66453 1007 4119.067 -380.933 61
111 1005 4102.722 -397.278 118 717.0 2088.230 588.23054 1016 4193.024 -306.976 62
112 1019 4217.822 -282.178 119 699.9 1989.811 489.81155 1067 4624.542 124.542 63 683.9 1899.875 399.875
113 1065 4607.222 107.222 120 662.7 1783.914 283.91456 1043 4418.843 -81.157 64 722.7 2121.563 621.563
114
Table C.13 - Strain after 16 foot excavation - Monroe Pile #1 (SE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
1 611.6 1519.410 19.410 9 1022 4242.694 -257.30665 612.2 1522.392 22.392 722 583.6 1383.472 -116.528 10 1048 4461.311 -38.689
66 638.3 1654.968 154.968 73 1061 4572.679 72.6793 567.3 1307.271 -192.729 11 1051 4486.889 -13.111
67 609.8 1510.479 10.479 74 1044 4427.320 -72.6804 508.9 1051.974 -448.026 12 1023 4251.001 -248.999
68 538.3 1177.033 -322.967 75 1043 4418.843 -81.1575 13 1022 4242.694 -257.306
69 580.5 1368.814 -131.186 76 1048 4461.311 -38.6896 575.9 1347.206 -152.794 14 1027 4284.309 -215.691
70 77 1039 4385.014 -114.9867 601.2 1468.175 -31.825 15 1044 4427.320 -72.680
71 672.8 1838.704 338.704 78 1037 4368.149 -131.8518 620.1 1561.937 61.937 16 1030 4309.376 -190.624
Pile #2 (NE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
17 535.6 1165.255 -334.745 25 1014 4176.532 -323.46879 628.3 1603.519 103.519 86 1070 4650.584 150.58418 618.4 1553.384 53.384 26 1068 4633.215 133.21580 602.6 1475.021 -24.979 87 1041 4401.912 -98.08819 612.0 1521.398 21.398 27 1001 4070.128 -429.87281 533.2 1154.836 -345.164 88 1054 4512.541 12.54120 524.2 1116.179 -383.821 28 993 4005.331 -494.66982 89 1010 4143.646 -356.35421 2983 614.7 1534.851 34.851 90 1014 4176.532 -323.46822 531.6 1147.915 -352.085 30 1021 4234.395 -265.60584 669.6 1821.255 321.255 91 1058 4546.857 46.85723 635.7 1641.513 141.513 31 1078 4720.385 220.38585 663.2 1786.607 286.607 92 1038 4376.578 -123.42224 584.3 1386.793 -113.207 32 1025 4267.639 -232.361
Pile #3 (SW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
33 1034 4342.912 -157.088 41 605.7 1490.236 -9.76493 1039 4385.014 -114.986 100 635.8 1642.030 142.03034 1057 4538.265 38.265 4294 1007 4119.067 -380.933 101 588.6 1407.280 -92.72035 1008 4127.252 -372.748 43 648.7 1709.337 209.33795 1022 4242.694 -257.306 102 621.3 1567.988 67.98836 1022 4242.694 -257.306 44 585.1 1390.593 -109.40796 1019 4217.822 -282.178 103 698.5 1981.859 481.85937 1032 4326.127 -173.873 45 575.7 1346.271 -153.72997 1066 4615.878 115.878 104 654.6 1740.572 240.57238 1053 4503.982 3.982 46 565.0 1296.692 -203.30898 1040 4393.459 -106.541 105 636.7 1646.682 146.68239 1076 4702.886 202.886 47 697.9 1978.456 478.45699 1052 4495.432 -4.568 106 654.1 1737.914 237.91440 48 624.3 1583.166 83.166
Pile #4 (NW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
49 1054 4512.541 12.541 57 557.7 1263.401 -236.599107 11450 58 665.9 1801.183 301.183
108 1017 4201.282 -298.718 11551 59 729.9 2164.047 664.047
109 986 3949.060 -550.940 116 710.1 2048.231 548.23152 60 753.1 2303.802 803.802
110 985 3941.054 -558.946 117 747.2 2267.846 767.84653 1004 4094.561 -405.439 61
111 1003 4086.409 -413.591 118 719.0 2099.896 599.89654 1015 4184.774 -315.226 62
112 1019 4217.822 -282.178 119 699.6 1988.106 488.10655 1068 4633.215 133.215 63 683.0 1894.878 394.878
113 1066 4615.878 115.878 120 661.8 1779.072 279.07256 1044 4427.320 -72.680 64 722.3 2119.216 619.216
115
Table C.14 - Strain after 20 foot excavation - Monroe Pile #1 (SE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
1 611.1 1516.926 16.926 9 1021 4234.395 -265.60565 614.1 1531.857 31.857 722 585.1 1390.593 -109.407 10 1049 4469.829 -30.171
66 638.1 1653.931 153.931 73 1061 4572.679 72.6793 565.7 1299.907 -200.093 11 1053 4503.982 3.982
67 610.3 1512.957 12.957 74 1047 4452.801 -47.1994 504.4 1033.451 -466.549 12 1027 4284.309 -215.691
68 530.4 1142.739 -357.261 75 1047 4452.801 -47.1995 13 1028 4292.657 -207.343
69 556.3 1257.066 -242.934 76 1051 4486.889 -13.1116 567.9 1310.037 -189.963 14 1028 4292.657 -207.343
70 77 1037 4368.149 -131.8517 598.4 1454.531 -45.469 15 1040 4393.459 -106.541
71 671.1 1829.424 329.424 78 1034 4342.912 -157.0888 618.8 1555.394 55.394 16 1028 4292.657 -207.343
Pile #2 (NE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
17 532.6 1152.238 -347.762 25 1014 4176.532 -323.46879 626.6 1594.853 94.853 86 1070 4650.584 150.58418 616.2 1542.351 42.351 26 1068 4633.215 133.21580 600.5 1464.758 -35.242 87 1040 4393.459 -106.54119 617.7 1549.869 49.869 27 1004 4094.561 -405.43981 534.6 1160.908 -339.092 88 1056 4529.682 29.68220 519.4 1095.832 -404.168 28 990 3981.166 -518.83482 89 1011 4151.856 -348.14421 2983 633.1 1628.113 128.113 90 1008 4127.252 -372.74822 542.9 1197.236 -302.764 30 1018 4209.548 -290.45284 667.1 1807.681 307.681 91 1057 4538.265 38.26523 634.7 1636.353 136.353 31 1078 4720.385 220.38585 662.9 1784.991 284.991 92 1037 4368.149 -131.85124 584.1 1385.844 -114.156 32 1025 4267.639 -232.361
Pile #3 (SW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
33 1034 4342.912 -157.088 41 603.0 1476.980 -23.02093 1039 4385.014 -114.986 100 633.9 1632.230 132.23034 1059 4555.456 55.456 4294 1008 4127.252 -372.748 101 585.8 1393.923 -106.07735 1010 4143.646 -356.354 43 636.8 1647.199 147.19995 1027 4284.309 -215.691 102 618.0 1551.375 51.37536 1029 4301.012 -198.988 44 566.7 1304.507 -195.49396 1030 4309.376 -190.624 103 682.3 1890.996 390.99637 1031 4317.748 -182.252 45 571.1 1324.842 -175.15897 1059 4555.456 55.456 104 660.0 1769.407 269.40738 1048 4461.311 -38.689 46 560.2 1274.753 -225.24798 1038 4376.578 -123.422 105 629.9 1611.696 111.69639 1074 4685.420 185.420 47 691.6 1942.897 442.89799 1050 4478.355 -21.645 106 650.4 1718.308 218.30840 48 621.7 1570.007 70.007
Pile #4 (NW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
49 1055 4521.108 21.108 57 552.4 1239.502 -260.498107 11450 58 660.2 1770.480 270.480
108 1021 4234.395 -265.605 11551 59 721.8 2116.283 616.283
109 992 3997.268 -502.732 116 698.2 1980.157 480.15752 60 753.2 2304.414 804.414
110 981 3909.110 -590.890 117 710.7 2051.694 551.69453 1000 4062.000 -438.000 61
111 999 4053.880 -446.120 118 719.7 2103.986 603.98654 1013 4168.298 -331.702 62
112 1019 4217.822 -282.178 119 696.0 1967.698 467.69855 1067 4624.542 124.542 63 680.3 1879.926 379.926
113 1066 4615.878 115.878 120 659.8 1768.335 268.33556 1044 4427.320 -72.680 64 721.0 2111.594 611.594
116
Table C.1 Monroe 5 - Strain after 20 foot excavation and 8 foot surcharge -Pile #1 (SE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
1 611.1 1516.926 16.926 9 1021 4234.395 -265.60565 614.1 1531.857 31.857 722 585.1 1390.593 -109.407 10 1049 4469.829 -30.171
66 638.1 1653.931 153.931 73 1061 4572.679 72.6793 565.7 1299.907 -200.093 11 1053 4503.982 3.982
67 610.3 1512.957 12.957 74 1047 4452.801 -47.1994 504.4 1033.451 -466.549 12 1027 4284.309 -215.691
68 530.4 1142.739 -357.261 75 1047 4452.801 -47.1995 13 1028 4292.657 -207.343
69 556.3 1257.066 -242.934 76 1051 4486.889 -13.1116 567.9 1310.037 -189.963 14 1028 4292.657 -207.343
70 77 1037 4368.149 -131.8517 598.4 1454.531 -45.469 15 1040 4393.459 -106.541
71 671.1 1829.424 329.424 78 1034 4342.912 -157.0888 618.8 1555.394 55.394 16 1028 4292.657 -207.343
Pile #2 (NE)Tension CompressionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
17 532.6 1152.238 -347.762 25 1014 4176.532 -323.46879 626.6 1594.853 94.853 86 1070 4650.584 150.58418 616.2 1542.351 42.351 26 1068 4633.215 133.21580 600.5 1464.758 -35.242 87 1040 4393.459 -106.54119 617.7 1549.869 49.869 27 1004 4094.561 -405.43981 534.6 1160.908 -339.092 88 1056 4529.682 29.68220 519.4 1095.832 -404.168 28 990 3981.166 -518.83482 89 1011 4151.856 -348.14421 2983 633.1 1628.113 128.113 90 1008 4127.252 -372.74822 542.9 1197.236 -302.764 30 1018 4209.548 -290.45284 667.1 1807.681 307.681 91 1057 4538.265 38.26523 634.7 1636.353 136.353 31 1078 4720.385 220.38585 662.9 1784.991 284.991 92 1037 4368.149 -131.85124 584.1 1385.844 -114.156 32 1025 4267.639 -232.361
Pile #3 (SW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
33 1034 4342.912 -157.088 41 602.5 1474.531 -25.46993 1039 4385.014 -114.986 100 633.6 1630.686 130.68634 1058 4546.857 46.857 4294 1008 4127.252 -372.748 101 585.5 1392.495 -107.50535 1010 4143.646 -356.354 43 636.8 1647.199 147.19995 1026 4275.970 -224.030 102 618.1 1551.877 51.87736 1028 4292.657 -207.343 44 566.7 1304.507 -195.49396 1029 4301.012 -198.988 103 682.2 1890.442 390.44237 1030 4309.376 -190.624 45 570.8 1323.451 -176.54997 1058 4546.857 46.857 104 659.5 1766.727 266.72738 1048 4461.311 -38.689 46 559.4 1271.115 -228.88598 1037 4368.149 -131.851 105 629.2 1608.116 108.11639 1074 4685.420 185.420 47 691.3 1941.212 441.21299 1049 4469.829 -30.171 106 649.6 1714.083 214.08340 48 621.0 1566.474 66.474
Pile #4 (NW)Compression TensionGage Frequency (Hz) Microstrain (με) Δ με Gage Frequency (Hz) Microstrain (με) Δ με
49 1055 4521.108 21.108 57 552.7 1240.849 -259.151107 11450 58 660.6 1772.626 272.626
108 1021 4234.395 -265.605 11551 59 721.4 2113.938 613.938
109 992 3997.268 -502.732 116 698.0 1979.023 479.02352 60 753.1 2303.802 803.802
110 981 3909.110 -590.890 117 709.9 2047.077 547.07753 999 4053.880 -446.120 61
111 998 4045.768 -454.232 118 719.3 2101.648 601.64854 1013 4168.298 -331.702 62
112 1018 4209.548 -290.452 119 695.2 1963.177 463.17755 1067 4624.542 124.542 63 679.3 1874.404 374.404
113 1065 4607.222 107.222 120 659.0 1764.049 264.04956 1043 4418.843 -81.157 64 720.3 2107.496 607.496
117
118
INCLINOMETER DATA
APPENDIX D
119
Plot Coordinates for Inclinometer NWRelative to initial reading on 6 Oct2005 at 1206
UNC Charlotte - J. Brian Anderson, Ph.D., P.E.
Printed 11-09-2005 at 18:02:53
skew = 135deg
18 Oct-05 19 Oct-05 20 Oct-05 21-Oct-05 25-Oct-05 27-Oct-05Depth Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum
Table D. 1 - NW inclinometer Statesville
------- ------------ ---------- ----------- --------- ------------ -------- ---------- ------------ -------- ------------ -------- ------1 0.042 -0.033 0.174 0.02 0.401 0.035 0.643 -0.003 0.729 0.003 0.9423 0.045 -0.035 0.16 0.016 0.365 0.032 0.59 -0.006 0.668 0.001 0.8945 0.041 -0.028 0.141 0.008 0.317 0.026 0.522 -0.007 0.587 0 0.8187 0.036 -0.022 0.12 0.006 0.266 0.02 0.446 -0.01 0.501 -0.003 0.7269 0.028 -0.02 0.093 0.003 0.219 0.017 0.381 -0.006 0.431 0.002 0.647
11 0.023 -0.016 0.065 0.001 0.169 0.011 0.313 -0.005 0.359 0 0.56413 0.022 -0.014 0.044 -0.002 0.13 0.006 0.265 -0.002 0.307 0 0.49615 0.021 -0.014 0.029 -0.012 0.087 -0.006 0.215 -0.012 0.249 -0.012 0.431 717 0.02 -0.014 0.022 -0.011 0.058 -0.008 0.172 -0.016 0.198 -0.016 0.374 219 0.022 -0.013 0.019 -0.008 0.041 -0.006 0.121 -0.014 0.142 -0.014 0.317 2121 0.019 -0.011 0.014 -0.005 0.027 -0.003 0.074 -0.011 0.091 -0.011 0.246 -0.02123 0.017 -0.011 0.011 -0.004 0.019 -0.002 0.042 -0.011 0.053 -0.01 0.157 -0.01425 0.014 -0.008 0.008 -0.003 0.01 0 0.022 -0.01 0.028 -0.008 0.083 -0.00927 0.008 -0.003 0.005 0 0.002 0.003 0.007 -0.006 0.014 -0.005 0.038 -0.00629 0.005 -0.003 0.003 -0.001 -0.001 0.004 0.001 -0.002 0.006 -0.003 0.013 -0.00131 0 0 0 0 0 0 0 0 0 0 0 0
Y----
0.0-0.0-0.
-0.0-0.
-0.0-0.0-0.01
-0.0-0.0
05110224021711
Table D. 2 - SW inclinometer Statesville
120
Plot Coordinates for Inclinometer SWRelative to initial reading on 4 Oct2005 at 1443
UNC Charlotte - J. Brian Anderson, Ph.D., P.E.
Printed 11-09-2005 at 18:19:29
skew = 135deg
10/18/2005 10/18/2005 10/19/2005 10/19/2005 10/20/2005 10/20/2005 10/21/2005 10/21/2005 10/21/2005 10/21/2005 10/25/2005 10/25/2005 10/27/2005 10/27/2005Depth Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y
----(ft)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)-----3 0.02 0.007 0.081 0.027 0.182 0.027 0.247 0.04 0.303 0.02 0.398 0.035 0.588 -0.0055 0.014 0.006 0.071 0.022 0.165 0.026 0.231 0.042 0.283 0.024 0.367 0.038 0.55 0.0197 0.01 0.006 0.054 0.017 0.143 0.024 0.209 0.045 0.261 0.028 0.336 0.041 0.518 0.0349 0.008 0.006 0.034 0.016 0.119 0.022 0.188 0.044 0.241 0.029 0.307 0.039 0.496 0.0411 0.007 0.006 0.021 0.012 0.093 0.016 0.169 0.04 0.22 0.025 0.28 0.03 0.484 0.03413 0.008 0.006 0.015 0.008 0.064 0.011 0.151 0.032 0.2 0.017 0.256 0.018 0.475 0.02515 0.008 0.003 0.011 0.006 0.043 0.01 0.123 0.026 0.172 0.008 0.223 0.01 0.458 0.01517 0.007 0.003 0.01 0.005 0.027 0.007 0.081 0.028 0.119 0.009 0.165 0.01 0.411 0.01419 0.005 0.004 0.009 0.004 0.018 0.006 0.047 0.025 0.067 0.011 0.108 0.013 0.333 0.01621 0.004 0.005 0.006 0.003 0.012 0.003 0.029 0.022 0.037 0.008 0.075 0.008 0.247 0.01923 0.004 0.007 0.009 0.001 0.011 0.003 0.02 0.016 0.021 0.008 0.059 0.001 0.17 0.01425 0.003 0.006 0.008 0.001 0.008 0.003 0.012 0.015 0.015 0.004 0.045 -0.003 0.114 0.00727 0.004 0.006 0.01 0 0.01 0.001 0.01 0.012 0.01 0.001 0.031 0 0.074 -0.00129 0 0.006 0.007 0 0.008 0.002 0.009 0.007 0.004 -0.002 0.021 -0.002 0.043 -0.00331 0 0.003 0.001 0.001 0.006 -0.001 0.005 0.003 -0.003 -0.003 0.011 -0.002 0.019 -0.00233 0 0 0 0 0 0 0 0 0 0 0 0 0 0
Table D. 3 - NE inclinometer Statesville
121
Plot Coordinates for Inclinometer NERelative to initial reading on 4 Oct2005 at 1215
UNC Charlotte - J. Brian Anderson, Ph.D., P.E.
Printed 11-09-2005 at 18:11:55
skew = -45deg
10/18/2005 10/18/2005 10/19/2005 10/19/2005 10/20/2005 10/20/2005 10/21/2005 10/21/2005 10/25/2005 10/25/2005 10/27/2005 10/27/2005Depth Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y
----(ft)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)-----1 0.062 0.003 0.319 -0.038 0.901 -0.118 1.628 -0.153 1.944 -0.155 3.288 -0.2153 0.048 -0.004 0.28 -0.027 0.823 -0.105 1.485 -0.147 1.767 -0.15 3.019 -0.2165 0.033 -0.006 0.227 -0.027 0.72 -0.098 1.318 -0.138 1.565 -0.141 2.731 -0.2027 0.019 -0.006 0.177 -0.036 0.613 -0.106 1.143 -0.14 1.358 -0.139 2.434 -0.1939 -0.001 -0.008 0.132 -0.038 0.504 -0.106 0.969 -0.136 1.152 -0.132 2.14 -0.1811 -0.014 -0.007 0.079 -0.028 0.399 -0.092 0.808 -0.122 0.963 -0.119 1.867 -0.1613 -0.02 -0.005 0.036 -0.023 0.327 -0.076 0.678 -0.111 0.807 -0.107 1.627 -0.13915 -0.02 -0.002 -0.004 -0.007 0.241 -0.038 0.551 -0.065 0.656 -0.061 1.391 -0.08617 -0.022 0.002 -0.019 0 0.148 -0.013 0.426 -0.028 0.505 -0.02 1.156 -0.04219 -0.011 -0.004 -0.019 0.001 0.067 -0.008 0.305 -0.014 0.364 -0.008 0.928 -0.02421 -0.007 -0.005 -0.015 0 0.023 -0.003 0.185 -0.014 0.226 -0.008 0.706 -0.01823 -0.005 -0.003 -0.01 0.001 0.005 0.005 0.087 -0.011 0.115 -0.008 0.507 -0.01525 -0.004 -0.002 -0.008 -0.001 0 0.007 0.035 -0.003 0.051 -0.002 0.332 -0.01927 -0.002 -0.001 -0.006 0 -0.005 0.003 0.012 0.003 0.018 0.004 0.188 -0.01629 -0.002 -0.001 -0.003 0 -0.004 0.002 0.003 0.006 0.004 0.006 0.087 -0.01131 -0.001 0 -0.001 0 -0.002 0.001 0 0.004 0 0.005 0.031 -0.00333 0 0 0 0 0 0 0 0 0 0 0 0
122
Plot Coordinates for Inclinometer SERelative to initial reading on 4 Oct2005 at 1454
UNC Charlotte - J. Brian Anderson, Ph.D., P.E.
Printed 11-11-2005 at 14:19:33
skew = -45deg
10/18/2005 10/18/2005 10/19/2005 10/19/2005 10/20/2005 10/20/2005 10/20/2005 10/20/2005 10/21/2005 10/21/2005 10/21/2005 10/21/2005 10/25/2005 10/25/2005 10/27/2005 10/27/2005Depth Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y
----(ft)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)------8.2 0.163 0.063 0.37 0.057 1.056 0.484 1.045 0.372 1.456 0.548 1.745 0.664 1.985 0.79 3.805 1.191-6.2 0.117 0.032 0.324 0.051 0.93 0.423 0.94 0.341 1.306 0.5 1.585 0.612 1.819 0.728 3.525 1.073-4.2 0.073 0.022 0.279 0.046 0.803 0.365 0.838 0.299 1.165 0.445 1.441 0.547 1.655 0.663 3.241 0.966-2.2 0.042 0.013 0.232 0.038 0.68 0.303 0.727 0.248 1.024 0.384 1.289 0.48 1.485 0.593 2.963 0.86-0.2 0.024 0.013 0.192 0.029 0.556 0.239 0.613 0.192 0.878 0.311 1.135 0.403 1.302 0.51 2.679 0.7511.8 0.015 0.012 0.148 0.022 0.435 0.182 0.497 0.135 0.735 0.241 0.98 0.325 1.127 0.429 2.395 0.6393.8 0.003 0.011 0.112 0.012 0.343 0.124 0.399 0.087 0.608 0.178 0.833 0.252 0.964 0.331 2.14 0.5365.8 -0.004 0.01 0.081 0.011 0.274 0.086 0.324 0.052 0.507 0.126 0.702 0.19 0.825 0.261 1.903 0.4497.8 -0.005 0.003 0.043 0.003 0.22 0.063 0.262 0.035 0.419 0.087 0.583 0.139 0.69 0.199 1.667 0.3649.8 -0.006 -0.001 0.012 -0.006 0.177 0.049 0.208 0.027 0.344 0.063 0.473 0.1 0.566 0.144 1.437 0.29111.8 -0.005 -0.002 -0.006 -0.005 0.125 0.047 0.148 0.029 0.281 0.055 0.377 0.08 0.449 0.112 1.213 0.2313.8 -0.006 -0.002 -0.013 -0.005 0.062 0.034 0.076 0.022 0.211 0.048 0.281 0.065 0.333 0.084 0.985 0.17115.8 -0.006 -0.003 -0.013 -0.005 0.014 0.02 0.02 0.011 0.132 0.031 0.182 0.045 0.216 0.052 0.757 0.10617.8 -0.006 -0.003 -0.011 -0.005 -0.003 0.012 -0.003 0.008 0.061 0.016 0.09 0.024 0.111 0.023 0.538 0.05419.8 -0.003 -0.002 -0.007 -0.004 -0.007 0.007 -0.006 0.006 0.023 0.011 0.032 0.01 0.044 0.008 0.339 0.02221.8 -0.002 -0.002 -0.003 -0.001 -0.005 0.004 -0.005 0.005 0.006 0.005 0.01 0.005 0.014 0.004 0.156 0.00523.8 0 0 0 0 0 0 0 0 0 0 0 0 0 0 0 0
Table D. 4 - SE inclinometer Statesville
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0 -1 -0.5 0 0.5 1
Depth(ft)
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0
-1 -0.5 0 0.5 1 32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0 -1 -0.5 0 0.5 1LEGEND
Initial 6 Oct2005
Depth(ft)
0
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
-1 -0.5 0 0.5 1
18 Oct200519 Oct200520 Oct200521 Oct200525 Oct200527 Oct2005
Cumulative DeflectionDirection X
Deflection (in)
Cumulative DeflectionDirection Y
Deflection (in)
skew = 135deg
Retaining Structures Research, Inclinometer NWUNCC/NCDOT
J. Brian Anderson, UNC Charlotte
Figure D.1 - NW inclinometer Statesville
123
34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0
Depth(ft)
-1 -0.5 0 0.5 1
34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0
-1 -0.5 0 0.5 1 34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0
Depth(ft)
-1 -0.5 0 0.5 1
34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0
-1 -0.5 0 0.5 1
LEGENDInitial 4 Oct2005
18 Oct200519 Oct200520 Oct200521 Oct200521 Oct200525 Oct200527 Oct2005
Cumulative DeflectionDirection X
Deflection (in)
Cumulative DeflectionDirection Y
Deflection (in)
skew = 135deg
Retinaing Structures Research, Inclinometer SWUNCC/NCDOT
J. Brian Anderson, UNC Charlotte
Figure D.2 - SW inclinometer Statesville
124
34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0
Depth(ft)
-4 -2 0 2 4
34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0
-4 -2 0 2 4 34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0
Depth(ft)
-4 -2 0 2 4
34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0
-4 -2 0 2 4
LEGENDInitial 4 Oct2005
18 Oct200519 Oct200520 Oct200521 Oct200525 Oct200527 Oct2005
Cumulative DeflectionDirection X
Deflection (in)
Cumulative DeflectionDirection Y
Deflection (in)
skew = -45deg
Retaining Structures Research, Inclinometer NEUNCC/NCDOT
J. Brian Anderson, UNC Charlotte
Figure D.3 - NE inclinometer Statesville
125
126
24
22
20
18
16
14
12
10
8
6
4
2
0
-2
-4
-6
-8
-10
Depth(ft)
-5 -2.5 0 2.5 5
24
22
20
18
16
14
12
10
8
6
4
2
0
-2
-4
-6
-8
-10
-5 -2.5 0 2.5 5 24
22
20
18
16
14
12
10
8
6
4
2
0
-2
-4
-6
-8
-10
Depth(ft)
-5 -2.5 0 2.5 5
24
22
20
18
16
14
12
10
8
6
4
2
0
-2
-4
-6
-8
-10
-5 -2.5 0 2.5 5
LEGENDInitial 4 Oct2005
18 Oct200519 Oct200520 Oct200520 Oct200521 Oct200521 Oct200525 Oct200527 Oct2005
Cumulative DeflectionDirection X
Deflection (in)
Cumulative DeflectionDirection Y
Deflection (in)
Ref. Elevation 0 ftskew = -45deg
Retaining Structures Research, Inclinometer SEUNCC/NCDOT
J. Brian Anderson, UNC Charlotte
Figure D.4 - SE inclinometer Statesville
127
Plot Coordinates for Inclinometer NWRelative to initial reading on 7 Dec2006 at 1421
J. Brian Anderson, UNC Charlotte
Printed 09-03-2007 at 17:01:18
skew = 135deg
12/11/2006 1
Table D. 5 - NW inclinometer Monroe
2006 12/12/2006 12/12/2006 12/12/2006 12/12/2006 12/13/2006 12/13/2006 12/14/2006 12/14/2006 12/14/2006 12/14/2006 12/17/2006 12/17/2006Depth Cum X Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y
----(ft)----- -----(in)----- - ----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)-----5 0 4 0.015 0.039 0.18 -0.059 0.309 -0.141 0.607 -0.263 0.656 -0.248 0.651 -0.2657 0.001 3 0.004 0.043 0.133 -0.028 0.244 -0.108 0.493 -0.213 0.534 -0.195 0.533 -0.2139 0.001 7 -0.002 0.043 0.092 -0.013 0.185 -0.093 0.391 -0.188 0.425 -0.166 0.426 -0.18511 0.003 25 -0.008 0.044 0.048 0.003 0.12 -0.069 0.278 -0.141 0.306 -0.118 0.309 -0.13513 0.003 24 -0.011 0.043 0.019 0.009 0.071 -0.053 0.184 -0.104 0.205 -0.078 0.21 -0.09415 0.003 24 -0.012 0.044 0 0.014 0.03 -0.043 0.101 -0.074 0.116 -0.046 0.12 -0.06117 -0.001 5 -0.015 0.043 0 0.015 0.01 -0.035 0.045 -0.048 0.056 -0.015 0.057 -0.03119 -0.005 4 -0.017 0.044 -0.001 0.016 0.011 -0.027 0.019 -0.023 0.023 0.014 0.018 021 -0.003 6 -0.014 0.046 -0.005 0.023 0.018 -0.012 0.015 0 0.018 0.039 0.01 0.02823 0 27 -0.009 0.048 -0.007 0.03 0.017 -0.005 0.027 0.009 0.009 0.051 0.003 0.04125 -0.001 9 -0.001 0.052 -0.006 0.039 0.02 0.001 0.021 0.014 0.009 0.053 0.006 0.04427 -0.003 9 -0.001 0.056 -0.011 0.047 0.013 0.008 0.015 0.014 0.003 0.052 0 0.04329 0.008 31 0.008 0.045 -0.001 0.041 0.009 0.016 0.019 0.008 0.001 0.051 0.006 0.03231 -0.006 5 -0.004 0.038 -0.008 0.038 -0.003 0.018 -0.002 0.021 -0.011 0.042 -0.006 0.02533 0 0 0 0 0 0 0 0 0 0 0 0 0 0
2/11/Cum
----(in)0.030.0
0.020.00.00.00.020.020.020.00.030.030.00.02
Table D. 6- SW inclinometer Monroe
128
Plot Coordinates for Inclinometer SWRelative to initial reading on 7 Dec2006 at 1412
J. Brian Anderson, UNC Charlotte
Printed 09-03-2007 at 17:01:42
skew = 135deg
12/11/2006 12/11/2006 12/12/2006 12/12/2006 12/12/2006 12/12/2006 12/13/2006 12/13/2006 12/14/2006 12/14/2006 12/14/2006 12/14/2006 12/17/2006 12/17/2006Depth Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y
----(ft)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)-----1 -0.087 0.022 -0.063 -0.034 -0.061 0.012 -0.041 0.003 -0.074 0.083 -0.035 0.088 -0.022 0.0743 -0.088 0.017 -0.058 -0.039 -0.046 -0.001 -0.023 0.001 0.013 0.071 0.037 0.081 0.054 0.0715 -0.079 0.012 -0.05 -0.041 -0.034 -0.004 0.021 0.003 0.09 0.073 0.112 0.084 0.13 0.0747 -0.074 0.012 -0.045 -0.043 -0.03 -0.003 0.06 0.013 0.162 0.084 0.18 0.098 0.202 0.0859 -0.068 0.011 -0.048 -0.043 -0.032 0 0.099 0.032 0.232 0.104 0.25 0.118 0.272 0.10511 -0.065 0.011 -0.052 -0.042 -0.036 0.008 0.147 0.066 0.307 0.14 0.323 0.154 0.346 0.14313 -0.063 0.01 -0.053 -0.04 -0.04 0.016 0.179 0.101 0.355 0.18 0.373 0.193 0.394 0.18315 -0.062 0.007 -0.053 -0.042 -0.053 0.01 0.158 0.095 0.345 0.176 0.36 0.189 0.378 0.18117 -0.057 0.004 -0.05 -0.043 -0.056 0.002 0.075 0.045 0.254 0.124 0.268 0.138 0.28 0.12919 -0.054 0.008 -0.048 -0.04 -0.053 0.001 -0.014 -0.004 0.121 0.05 0.131 0.067 0.14 0.05921 -0.051 0.003 -0.043 -0.043 -0.051 -0.002 -0.044 -0.027 0.014 -0.011 0.02 0.008 0.027 -0.00223 -0.045 -0.005 -0.038 -0.043 -0.044 -0.011 -0.039 -0.036 -0.033 -0.03 -0.028 -0.015 -0.024 -0.02125 -0.029 -0.017 -0.034 -0.037 -0.032 -0.017 -0.034 -0.032 -0.039 -0.025 -0.028 -0.018 -0.031 -0.02227 -0.029 -0.021 -0.034 -0.036 -0.025 -0.032 -0.031 -0.032 -0.037 -0.025 -0.025 -0.025 -0.026 -0.02929 -0.025 -0.033 -0.031 -0.041 -0.023 -0.045 -0.025 -0.04 -0.033 -0.03 -0.022 -0.032 -0.022 -0.03531 -0.026 -0.031 -0.027 -0.042 -0.026 -0.041 -0.024 -0.04 -0.026 -0.036 -0.022 -0.037 -0.025 -0.03833 0 0 0 0 0 0 0 0 0 0 0 0 0 0
Table D. 7 - NE inclinometer Monroe
129
Plot Coordinates for Inclinometer NERelative to initial reading on 7 Dec2006 at 1439
J. Brian Anderson, UNC Charlotte
Printed 09-03-2007 at 16:58:30
skew = -45deg
12/11/2006 12/11/2006 12/12/2006 12/12/2006 12/12/2006 12/12/2006 12/13/2006 12/13/2006 12/14/2006 12/14/2006 12/14/2006 12/14/2006 12/17/2006 12/17/2006Depth Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y
----(ft)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)-----1 0.036 -0.021 0.258 -0.081 0.466 -0.196 1.17 -0.414 1.936 -0.72 1.977 -0.695 1.951 -0.6863 0.043 -0.026 0.241 -0.09 0.43 -0.198 1.041 -0.406 1.732 -0.702 1.779 -0.68 1.759 -0.6745 0.042 -0.028 0.207 -0.088 0.376 -0.192 0.894 -0.385 1.51 -0.665 1.557 -0.644 1.542 -0.647 0.031 -0.022 0.148 -0.068 0.276 -0.158 0.703 -0.329 1.24 -0.586 1.288 -0.566 1.279 -0.5679 0.024 -0.018 0.07 -0.033 0.147 -0.105 0.476 -0.244 0.925 -0.466 0.977 -0.45 0.972 -0.45411 0.025 -0.019 0.028 -0.011 0.043 -0.06 0.266 -0.157 0.627 -0.339 0.681 -0.327 0.678 -0.3313 0.026 -0.017 0.021 -0.008 -0.014 -0.034 0.116 -0.095 0.403 -0.236 0.451 -0.235 0.446 -0.23515 0.028 -0.015 0.018 -0.008 -0.038 -0.022 0.02 -0.056 0.243 -0.168 0.291 -0.172 0.284 -0.17117 0.029 -0.016 0.013 -0.009 -0.04 -0.021 -0.018 -0.039 0.144 -0.125 0.188 -0.129 0.176 -0.12519 0.031 -0.019 0.014 -0.011 -0.024 -0.022 -0.041 -0.024 0.085 -0.088 0.117 -0.09 0.102 -0.08621 0.032 -0.017 0.017 -0.01 -0.011 -0.023 -0.034 -0.016 0.034 -0.051 0.051 -0.053 0.038 -0.04823 0.021 -0.012 0.016 -0.011 -0.009 -0.02 -0.014 -0.014 -0.001 -0.024 0.015 -0.025 -0.004 -0.0225 0.014 -0.011 0.008 -0.01 -0.009 -0.013 -0.014 -0.015 -0.004 -0.018 0.008 -0.017 -0.013 -0.01227 0.012 -0.01 0.003 -0.008 -0.002 -0.008 0.004 -0.013 0.006 -0.014 0.008 -0.014 -0.006 -0.01129 0.011 -0.008 0.01 -0.007 0.008 -0.008 0.008 -0.008 0.009 -0.009 0.008 -0.009 0.01 -0.00831 0 0 0 0 0 0 0 0 0 0 0 0 0 0
130
Plot Coordinates for Inclinometer SERelative to initial reading on 7 Dec2006 at 1431
J. Brian Anderson, UNC Charlotte
Printed 09-03-2007 at 16:59:47
skew = -45deg
12/11/2006 12/11/2006 12/12/2006 12/12/2006 12/12/2006 12/12/2006 12/13/2006 12/13/2006 12/14/2006 12/14/2006 12/14/2006 12/14/2006 12/17/2006 12/17/2006Depth Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y Cum X Cum Y
----(ft)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)----- -----(in)-----1 0 -0.042 0.032 -0.084 0.064 -0.063 0.228 -0.025 0.709 0.261 0.633 0.241 0.594 0.2133 0.009 -0.041 0.021 -0.078 0.062 -0.05 0.201 -0.026 0.66 0.242 0.594 0.229 0.56 0.2075 0.02 -0.045 0.014 -0.071 0.073 -0.046 0.185 -0.021 0.608 0.224 0.556 0.217 0.521 0.1987 0.023 -0.042 0.006 -0.062 0.067 -0.036 0.164 -0.01 0.559 0.218 0.522 0.213 0.487 0.1989 0.024 -0.033 0.002 -0.049 0.059 -0.023 0.148 0.004 0.52 0.216 0.494 0.212 0.465 0.203
11 0.027 -0.025 -0.002 -0.03 0.053 -0.008 0.14 0.019 0.484 0.213 0.467 0.213 0.449 0.20913 0.025 -0.026 -0.006 -0.024 0.046 0 0.14 0.032 0.46 0.21 0.453 0.215 0.443 0.21515 0.028 -0.028 -0.008 -0.026 0.04 -0.002 0.146 0.04 0.451 0.208 0.445 0.215 0.446 0.21817 0.026 -0.026 -0.007 -0.026 0.019 -0.004 0.137 0.042 0.417 0.196 0.416 0.202 0.42 0.20719 0.028 -0.026 0.001 -0.027 0.011 -0.009 0.098 0.025 0.359 0.165 0.363 0.168 0.365 0.17721 0.029 -0.028 0.003 -0.025 0.013 -0.021 0.053 -0.003 0.271 0.111 0.279 0.115 0.28 0.12723 0.024 -0.021 0.005 -0.015 0.009 -0.014 0.024 -0.008 0.156 0.064 0.16 0.067 0.161 0.07325 0.02 -0.012 0.005 -0.005 0.008 -0.01 0.012 -0.006 0.057 0.021 0.059 0.021 0.061 0.02327 0.011 -0.005 0.001 -0.001 0.003 -0.003 0.004 -0.001 0.011 0.004 0.01 0.004 0.012 0.00429 0 0 0 0 0 0 0 0 0 0 0 0 0 0
Table D. 8 - SE inclinometer Monroe
34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0 -1 -0.5 0 0.5 1
Depth(ft)
34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0
-1 -0.5 0 0.5 1 34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0 -1 -0.5 0 0.5 1LEGEND
Initial 7 Dec2006
Depth(ft)
0
34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2 11 Dec200612 Dec200612 Dec2006
-1 -0.5 0 0.5 1
13 Dec200614 Dec200614 Dec200617 Dec2006
Cumulative DeflectionDirection X
Deflection (in)
Cumulative DeflectionDirection Y
Deflection (in)
skew = 135deg
, Inclinometer NW
UNC Charlotte - J. Brian Anderson, Ph.D., P.E.
Figure D.5 - NW inclinometer Monroe
131
34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0
Depth(ft)
-0.5 -0.25 0 0.25 0.5
34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0
-0.5 -0.25 0 0.25 0.5 34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0
Depth(ft)
-0.5 -0.25 0 0.25 0.5
34
32
30
28
26
24
22
20
18
16
14
12
10
8
6
4
2
0
-0.5 -0.25 0 0.25 0.5
LEGENDInitial 7 Dec2006
11 Dec200612 Dec200612 Dec200613 Dec200614 Dec200614 Dec200617 Dec2006
Cumulative DeflectionDirection X
Deflection (in)
Cumulative DeflectionDirection Y
Deflection (in)
Ref. Elevation ftskew = 135deg
, Inclinometer SW
UNC Charlotte - J. Brian Anderson, Ph.D., P.E.
Figure D.6 - SW inclinometer Monroe
132
32
30
28
26
24
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16
14
12
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8
6
4
2
0
Depth(ft)
-2 -1 0 1 2
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-2 -1 0 1 2 32
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Depth(ft)
-2 -1 0 1 2
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0
-2 -1 0 1 2
LEGENDInitial 7 Dec2006
11 Dec200612 Dec200612 Dec200613 Dec200614 Dec200614 Dec200617 Dec2006
Cumulative DeflectionDirection X
Deflection (in)
Cumulative DeflectionDirection Y
Deflection (in)
Ref. Elevation ftskew = -45deg
, Inclinometer NE
UNC Charlotte - J. Brian Anderson, Ph.D., P.E.
Figure D.7 - NE inclinometer Monroe
133
30
28
26
24
22
20
18
16
14
12
10
8
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2
0
Depth(ft)
-1 -0.5 0 0.5 1
30
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-1 -0.5 0 0.5 1 30
28
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Depth(ft)
-1 -0.5 0 0.5 1
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-1 -0.5 0 0.5 1
LEGENDInitial 7 Dec2006
11 Dec200612 Dec200612 Dec200613 Dec200614 Dec200614 Dec200617 Dec2006
Cumulative DeflectionDirection X
Deflection (in)
Cumulative DeflectionDirection Y
Deflection (in)
skew = -45deg
, Inclinometer SE
UNC Charlotte - J. Brian Anderson, Ph.D., P.E.
Figure D.8 - SE inclinometer Monroe
134
APPENDIX E RECOMMENDATIONS FOR TRIAXIAL TESTS IN PIEDMONT RESIDUAL SOILS
135
Residual soils should be samp e with AASHTO T207 or ASTM
D1587 specifications. Unlike clays that are typically sampled using Shelby tubes,
residual soils require delicate and deliberate handling to insure the viability of test
specimens. The procedure currently used by the Materials and Tests Unit of NCDOT to
cut and extract specimens from Shelby Tubes utilizes a custom manufactured Shelby tube
vise and large pipe cutter, shown in figure E.1, has proven to effective in creation of
triaxial specimens.
The method is as follows:
1) If this is the first specimen from the Shelby tube, cut one inch from the
beveled end with a 3 inch pipe cutter to remove disturbed/contaminated soil.
2) Mark a six inch length from the cut end of the tube and note the direction the
orientation of the tube.
3) Place the Shelby tube in the tube vise and use a 3 inch pipe cutter to cut the
Shelby tube at the six inch mark. Tighten the clamp slowly to insure even
cutting and limit deformation of the tube. It should take more than 25 turns to
cut the tube.
4) When cutter penetrates the tube and the entire circumference is separated, use
a wire saw or knife edge to cut the soil inside.
5) Reseal the remaining tube length with a protective cap.
6) Using a vertically oriented jack extractor with a retaining ring appropriate for
Shelby tubes, extrude the soil specimen insuring the soil moves in the same
direction as when it was sampled.
led in accordanc
136
Each triaxial test should be conducted using AASHTO T297 or ASTM D4767.
Strain rates, and time to failure, can be determined based upon those specified by A
D4767 (2004), equation E.1, and recommendations by Bishop and Henkel (1957),
equation E.2.
STM
50
4%'10t
ε = (E.1)
Whe
t50 isre ‘ε is the strain rate
the time to 50% consolidation of the triaxial specimen
220f
v
ht = (cη
Ε.2)
Where t is time to failure
η r radial and drainage at both ends f consolidation
for increasing values of t
quickest practical tim inutes for 15% strain. On the other hand, tests
consolidation is 10 minutes or less (120 if filter strips are used), a drained test is
guide. Generally, the addition of
tests on the specim
fh is half height of the triaxial specimen
= 3 drainage at both ends or 35 focv is the coefficient o
Using 15% strain as the arbitrary failure point, the time to failure and strain rate
50 for a 3 x 6 inch specimen are shown in table E - 1. The
e to failure is 90 m
with a duration longer than 6 hours are impractical as well. Therefore if t50 during
recommended. For specimens with longer t50 values, undrained tests should be
conducted. The strain rates and times to failure shown in table E.1 can be used as a
filter strips will enhance the ability to conduct drained
ens.
137
138
Figu
Table E.1 – Time to failure and strain rate recommendations for drained tests
ASTM CU
t50 cv f '
(min) (in2/min) (min) (in/min) (min) (in/min) (min) (in/min) (%/min)
0.1
re E.1 – NCDOT Shelby Tube Vise and 3 inch pipe cutter
Bishop & Henkel CD Recommendation
End Drainage End and Radial Drainage
ε' t ε' ε tf ε' tf
17.64 3.4 0.2646 0.3 3.087 3.75 0.24 4.00
0.2
0 0.80
0.40
0 0.04
0.02
0.01
0.00
0.00
8.82 6.8 0.1323 0.6 1.5435 7.50 0.12 2.00
0.5 3.528 17. 0.05292 1.5 0.6174 18.75 0.048
1 1.764 34.0 0.02646 2.9 0.3087 37.50 0.024
2 0.882 68.0 0.01323 5.8 0.15435 75.00 0.012 0.20
5 0.3528 170.1 0.005292 14.6 0.06174 187.50 0.0048 0.08
10 0.1764 340.1 0.002646 29.2 0.03087 375.00 .0024
20 0.0882 680.3 0.001323 58.3 0.015435 750.00 0.0012
50 0.03528 1700.7 0.0005292 145.8 0.006174 1875.00 0.00048
100 0.01764 3401.4 0.0002646 291.5 0.003087 3750.00 0.00024
120 0.0147 4081.6 0.0002205 349.9 0.0025725 4500.00 0.0002