Repair and Strengthening of Highway Bridges with FRP

M.M. Dawood, O.A. Rosenboom & S.H. Rizkalla

Constructed Facilities Laboratory, North Carolina State University Email: sami_rizkalla@ncsu.edu

ABSTRACT

Two different design problems are commonly encountered by highway bridge maintenance engineers: the upgrade of existing members to handle increased load demand, or the repair of bridge members which have experienced impact events due to overheight vehicles. This paper shows that through the use of carbon fiber reinforced (CFRP) materials highway bridge girders can be effectively strengthened or repaired in flexure. Two ongoing research projects are described: the first involving the strengthening or repair of prestressed concrete bridge girders, and the second involving the strengthening of steel bridge girders. Design guidelines are presented for both techniques and show innovative use of the material characteristics of CFRP to suit the type of bridge and rehabilitation requirements. Keywords: Prestressed Concrete, Steel, FRP, High Modulus, Flexural, repair, strengthening INTRODUCTION

This paper presents the application of fiber reinforced polymer (FRP) materials for strengthening and repair of concrete and steel highway bridge girders. The techniques utilize the various characteristics of carbon FRP (CFRP) to suit the type of the bridge and the needs for rehabilitation. The first application presented in this paper focuses on the use of high strength CFRP material for strengthening and repair of prestressed concrete bridge girders. The second application examines the use of high modulus CFRP for strengthening steel bridge girders. Structural design recommendations and detailing guidelines are presented for CFRP strengthening and repair of prestressed concrete bridge girders which was evaluated through the experimental testing under static and fatigue loading conditions of seventeen 9.14 m and one 16.7 m bridge girders. Four different types of FRP systems were analyzed including near surface mounted (NSM) bars and strips, and externally bonded sheets and strips. The second application studied the behavior of steel-concrete composite bridge girders strengthened with new high modulus CFRP materials. A series of guidelines were established regarding the design of high modulus CFRP strengthening systems for steel bridge girders which were evaluated through the static and fatigue testing of six scaled steel-concrete composite bridge girders. Three design criteria are presented which should be satisfied to ensure the long-term durability and safety of the strengthened girders. The paper represents also the current state-of-the-art in the use of composite materials for rehabilitation and retrofit of existing bridge infrastructure. These projects demonstrate the various performance benefits which can be achieved by using FRP materials for the repair and strengthening of concrete and steel highway bridges.

APPLICATION ONE: REPAIR AND STRENGTHENING OF PRESTRESSED CONCRETE BRIDGE GIRDERS

To strengthen deficient reinforced concrete (RC) girders a number of techniques were commonly employed including external post-tensioning or the bonding of mild steel plates to the tension side of the member. The widespread availability and competitive cost of high strength composite materials have made strengthening using carbon fiber reinforced polymer (CFRP) materials a viable alternative to the traditional methods. While strength and stiffness gains due to CFRP strengthening of RC structures are well known1,2, the bond characteristics and mechanisms are less known and are currently being investigated3,4. The effect of fatigue loading on CFRP strengthened RC members has also been investigated5-7, with the stress ratio in the reinforcing steel identified as the critical component when practical levels of strengthening are attained. Flexural strengthening of prestressed concrete with CFRP materials has also been explored under static loading conditions8,9, and fatigue loading conditions10,11. Overall results from an extensive experimental study of CFRP flexural strengthening of prestressed concrete are presented below along with guidelines which can aid in the design of such systems. An additional experimental study examining the effectiveness of CFRP repair of impact damaged prestressed concrete bridge girders is presented along with repair guidelines. CFRP Repair and Strengthening Systems

Two different types of strengthening systems were used in the experimental study: externally bonded CFRP and near surface mounted (NSM) CFRP. Externally bonded systems for flexural strengthening are affixed to the concrete substrate on the tension side of the member and have two methods of installation: wet lay-up type systems, in which the carbon fibers are delivered and saturated on-site; and precured systems where the carbon fibers are impregnated with resin and pultruded in a manufacturing facility and subsequently bonded to the surface of the concrete using a two-part structural epoxy. To enhance the bond characteristics between the concrete and the CFRP material, the NSM strengthening method has gained traction for certain applications12,13. In this method, a groove is saw-cut into the concrete, a pultruded CFRP bar or strip inserted into the groove and surrounded with epoxy material. Experimental Program – Strengthening

A total of seventeen 9.14 m prestressed concrete bridge girders were tested under static and fatigue loading conditions to evaluate the effectiveness of CFRP strengthening. The results of the static and fatigue tests on twelve of the girders can be found in an earlier paper14, along with details regarding the test setup, instrumentation and loading scheme which were common for all the tests. The five prestressed concrete bridge girders considered herein have the same cross-section and span as the earlier girders but a slightly different prestressing configuration: eight 1862 MPa strands, six of which are harped at midspan as shown in Figure 1. One of the girders was tested under fatigue loading conditions as a control girder, and the other four were strengthened with two different systems and tested in static and fatigue: two girders with three layers of normal modulus CFRP sheets (VSL V-wrap C200), and two additional girders strengthened with five layers of high modulus CFRP sheets (Mitsubishi Dialead F637400). For all of the strengthened girders, externally bonded wet lay-up U-wraps were provided at 1000 mm spacing throughout the length of the girder to prevent debonding from occurring either at the plate termination point or propagating from midspan. The material characteristics of the concrete and the CFRP laminates were determined from ancillary testing and the results are presented in Table 1.

2@11

38114

3876

38140

End strandspacing

CL strandspacing

Ø/330Ø/180

775

Sec A-A

Two-part epoxy

EB HM CFRP sheets

Prestressing strand

5 Plies/web (125 mm wide)

Sheet thickness = 1.0 mm

40

6060

40

Sheet thickness = 1.0 mm

3 Plies/web (100 mm wide)

Prestressing strand

EB CFRP sheets

Two-part epoxy

S7, F6 S8, F7

150 150150150

Externally bonded U-wraps

300750750150

7507504050 mm

A

A4572 mm

Figure 1: Cross-section and elevation of strengthened girders

Table 1: Results of tests under static loading conditions Girder Designation F5 S7 S8 Strengthening technique

-- -- EB sheets EB HM sheets

FRP details -- -- 3 plies per web

5 plies per web

FRP shape -- -- 100 mm x 1 mm

125 mm x 1 mm

AFRP mm2 -- 600 1250 Ult tensile strength of FRP

MPa -- 724 396

EFRP MPa -- 64810 132150 FRP rupture strain mm/mm -- 0.0112 0.0030 F’c MPa 50.9 55.3 46.0 Ultimate load kN 142.3 246.0 149.9 % increase in capacity

% -- 72.8 5.3

Failure mode -- Crushing of concrete

Crushing of concrete

Rupture of CFRP

Experimental Results - Strengthening

Results of the Static Tests No control girder was tested under static loading conditions with the new prestressing strand configuration, but the control girder which was tested under fatigue loading (girder F5, described in the next section) was subjected to a final test to failure and this can be used as a static control

comparison. It should be noted that the behavior of the control girders with the two different prestressing configurations were nearly identical. The load-deflection relationship for the girders tested under static loading conditions is shown in Figure 2a and summarized results are shown in Table 1. Flexural cracking of girder S7, strengthened with three layers of normal modulus CFRP sheets, occurred at a load of 61.8 kN. Failure was due to crushing of the concrete at a load of 246.0 kN. The energy released upon crushing of concrete led to peeling of the U-wraps from the webs of the girder, but did not cause the CFRP to rupture. Girder S7 achieved an increase of strength of 72.8 percent compared to the ultimate strength measured during the final static test girder F5. The measured performance of the girder exceeded the design values due to a high rupture strain measured in the CFRP during the test, which was 18.8 percent higher than the value reported by the manufacturer. Flexural cracking of girder S8 strengthened with high modulus CFRP sheets occurred at a load of 63.6 kN. At a load of 149.9 kN failure occurred due to rupture of the CFRP sheets at midspan, which provides an increase of only 5.3 percent compared to the control. The tensile strain in the CFRP achieved during this test was 13.3 percent lower than the value reported by the manufacturer. After rupture of the CFRP, the test was continued until ultimate failure occurred due to a combination of progressive CFRP rupture and concrete crushing close to the values of ultimate load and displacement of the control girder as shown in Figure 2a.

0

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8

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12

14

1000 10000 100000 1000000 10000000Number of Cycles

SR in low

er p

rest

ress

ing s

tran

d (

%)

F5: Control

F6: EB sheets

F7: EB HM sheets

F7

F5

F6

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Midspan deflection (mm)

Load

(kN

)

F5: Control

S7: EB Sheets

S8: EB HM Sheets

S7

F5S8

(a) Load-deflection behavior (b) SR versus number of cycles Figure 2: Results of static and fatigue loading

Results of the Fatigue Tests Summarized test results for the three girders tested under fatigue loading conditions are presented in Table 2. Cracking of the control girder, F5, occurred at a load level of 55.6 kN. After the initial loading cycles, the girder was cycled between the load levels 8.9 kN and 49 kN, the same level of loading which girder F0 was subjected to in an earlier study12. F5 survived 2 million cycles with very little degradation. The girder was then loaded to failure, which occurred at a load of 142 kN which was 3.6 percent less than the ultimate strength achieved in the static test of the control girder prestressed by ten 1723 MPa strands (S0). Failure of the girder was due to crushing of the concrete.

Table 2: Results of tests under fatigue loading conditions Girder Designation F5 F6 F7 Strengthening technique

-- -- EB sheets EB HM sheets

FRP details -- -- see S7 see S8 f’c MPa 50.9 46.0 56.0 # of cycles achieved -- 2,000,000 1,075,000 2,000,000 Failure mode* % C RPS then D RFRP % change in ult load from virgin girder

% -- -43.8 -11.0

* C = crushing of concrete, RPS = rupture of prestressing strands, D = debonding of FRP, RFRP = rupture of FRP A second fatigue test was performed on a girder strengthened with three layers of normal modulus CFRP sheets (F6). After the initial loading, during which the cracking load was determined to be 60.1 kN, the girder was subjected to fatigue loading with a maximum load corresponding to a 60 percent increase in live load based on AASHTO HS-15 loading15. After the first 5,000 cycles considerable flexural cracking was noticed at midspan. No new cracks or crack extensions were noticed until 500,000 cycles, at which time a flexural crack at midspan extended into the top flange of the C-Channel. At 1 million cycles new flexural cracks were noticed on both sides of the girder at midspan. The girder failed at 1,075,000 cycles due to the rupture of one (or possibly two) lower prestressing strands at midspan. A final static test was performed after completion of 1,075,000 cycles. The loss of the lower prestressing strand (or strands) during the fatigue test caused an increase in load being distributed to the CFRP system. As a result, at a load of 75.6 kN signs of debonding began appearing. Interface cracks in the concrete just above the CFRP sheets expanded at locations of the flexural cracking and joined between the flexural cracks. The debonding propagated from midspan towards the supports, peeling off a layer of cover concrete between the lower prestressing strand and the CFRP. The slow debonding of the CFRP sheets due to the presence of the transverse U-wraps provided a ductile failure and the maximum load level observed was 138.3 kN. A girder strengthened with five layers of high modulus CFRP sheets was tested under fatigue loading (F7). As a result of an increase in the stiffness obtained due to the installation of the high modulus material, it was decided to cycle the girder up to a load value corresponding to a 20 percent increase compared to an equivalent HS-15 truck loading. During the initial static test, cracking was observed at a load level of 63.2 kN, very close to the cracking load observed in girder S8. After two million cycles of fatigue loading, very little degradation was noticed and the girder was tested to failure which occurred at a load of 133.4 kN due to rupture of the high modulus CFRP. The rupture strain of the HM material observed during the test was 0.24 percent comparable to the 0.26 percent observed in the static test on girder S8. After rupture of the sheets at midspan, the girder lost 23 percent of the applied load. The applied load at which rupture of the high modulus sheets was first observed was exceeded at a higher value of displacement. This shows that the high modulus material had a significant effect on the stiffness of the girder, even when not continuous throughout its length. After progressive rupture events, the girder followed the approximate load path of the control girder and the test was terminated after crushing of the concrete in the compression zone occurred at a displacement of 193 mm. The stress ratio in the prestressing strands, SRps, for the tested girders is shown in Figure 2b versus the number of cycles and can be defined as:

10012 ×−

=pu

pspsps f

ffSR (1)

where fps1 and fps2 are the stress levels and two different levels of loading and fpu is the ultimate tensile strength of the prestressing strands. It can be clearly seen in Figure 2b that the stress ratio

induced during fatigue loading for girder F6 is severe and lead to the rupture of prestressing strands, which began around 750,000 cycles. The stable stress ratio of around three to four percent for girders F5 and F7 is much more desirable and lead to the good fatigue performance of these girders. Proposed Design Guidelines - Strengthening

Based on the findings presented in this paper and companion papers11,14, the following design guidelines are recommended for strengthening of prestressed concrete girders with CFRP:

1. Analysis should follow a non-linear cracked section analysis approach with the CFRP material modeled as linear elastic to failure.

2. The material properties of the CFRP system should be reduced using the environmental reduction factor, CE, specified in ACI 44016, “Design and Construction of Externally Bonded FRP Systems for Strengthening Concrete Structures.”

3. The induced stress ratio in the prestressing strands due to the increased live load should be kept less than 6 percent for straight prestressing strands and 3 percent for harped prestressing strands.

4. The increase in ultimate flexural strength should satisfy the ultimate strength requirement using the proper AASHTO specified load, impact and resistance factors15.

5. The ultimate flexural strength of the unstrengthened girder should exceed the unfactored summation of dead and increased live load. This criterion ensures the safety of the bridge if the CFRP strengthening system is compromised by unintentional or intentional damage.

In addition to the design criteria presented above, several detailing and specific design rules are recommended:

1. Transverse CFRP wet lay-up U-wraps should be provided at most 915 mm spacing with the appropriate length into the depth of the girder to prevent any possible debonding in externally bonded CFRP strengthened girders. In addition, U-wraps should also be provided at all termination points of longitudinal externally bonded CFRP.

2. Due to the limitations of the ultimate strain of high modulus CFRP materials, they are not recommended for use in increasing the flexural strength of prestressed concrete members. However, due to their high stiffness they are recommended to enhance girder serviceability requirements.

Experimental Program – Repair

A leading cause of bridge girder replacement is impact damage caused by overheight vehicles. The traditional methods discussed above for strengthening are commonly employed to restore the original capacity of the damaged girder, but more likely the girder itself is simply replaced. In a research program currently underway, the viability of using CFRP wet lay-up systems to restore the ultimate capacity of impact damaged prestressed concrete bridge girders has been explored through the repair and testing of one 16.7 m girder. Experimental Results – Repair

A prestressed concrete AASHTO Type II girder suffered one ruptured prestressing strand and heavy loss of concrete section near midspan due to an impact event and was obtained for testing. The girder section was restored using a polymer-based mortar, and a CFRP system was designed to restore the ultimate strength of the original girder. Three layers of CFRP wet lay-up sheets 457 mm wide were used as the main longitudinal repair system, and other various CFRP tension struts and U-wraps were used for crack control and debonding resisting reinforcement. Full details of the repair system and other test details can be found elsewhere17.

Following the repair, the girder was brought to the laboratory and subjected to some initial loading cycles to determine the initial stiffness and prestress losses. Due to the fact that the repair concrete was unprestressed, the initial loading resulted in flexural cracking in the repaired region. The girder was then tested under fatigue loading conditions designed to simulate a nominal bottom tensile stress in the concrete of '0.25 cf (MPa). After two million cycles of fatigue loading, there was very little growth of the flexural cracks induced due to the initial loading cycles and the girder was tested to failure. The maximum measured load was 605.4 kN which occurred at a midspan displacement of 146.7 mm. At a displacement of 156.8 mm, a large flexure-shear crack suddenly formed outside the longitudinal CFRP repair area and connected to the concrete crushing zone near the edge of the loading plate and caused catastrophic failure as shown in Figure 3. The load-deflection behavior of the repaired girder is shown in Figure 4. Since no control girder was tested, also shown in the figure are the results of a cracked section analysis which was performed on the tested girder and a control girder. The results show that the ultimate strength and ductility of the repaired section were superior to the control girder prediction, and that the initial stiffness and serviceability were comparable to the control prediction.

Damaged region

Figure 3: Failure due to concrete crushing during final static test after completion of 2 million cycles

of fatigue loading

0

100

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300

400

500

600

0 25 50 75 100 125 150 175

Displacement (mm)

Load

(kN

)

Initial CyclesInitial Cycles (2)at 2000k cyclesFinal Cycles (1)Final Cycles (2)Undamaged Girder PredictionDamaged and Repaired Prediction

Final Cycles (1)

Final Cycles (2)

0.25sqrt(f'c)

Initial Cycles

Undamaged Girder Prediction

Damaged and Repaired Prediction

Figure 4: Load-deflection of repaired AASHTO girder Proposed Design Guidelines – Repair

Even though the experimental program dealing with the repair of prestressed concrete using CFRP is still ongoing, some preliminary design and detailing guidelines can be proposed:

1. Previous recommendations for CFRP strengthening regarding the analysis procedures and stress ratio limits also apply for CFRP repaired prestressed concrete.

2. Prior to concrete repair, the girder can be loaded with an appropriately sized truck to induce compressive stresses in the concrete repair material to match the original magnitude.

3. The longitudinal CFRP sheets should be extended a distance away from the ruptured prestressing strand a minimum distance equal to the development length of the prestressing strand.

4. The termination points for the longitudinal CFRP sheets should be staggered at a minimum distance of 600 mm from each other.

5. At each longitudinal CFRP termination point, a CFRP U-wrap should be provided to prevent debonding failures. The sheet shall be extended up the depth of the girder, up to and including the top flange.

6. Additional transverse CFRP sheets of the same length as (4) should be provided throughout the damaged region to control crack growth.

7. To control shrinkage and flexural cracking, additional longitudinal CFRP tension struts should be affixed to the girder web on both sides.

APPLICATION TWO: STRENGTHENING OF STEEL BRIDGE GIRDERS

While the use of CFRP materials has gained widespread application for the repair and strengthening of reinforced concrete structures, due to the relatively low modulus of elasticity, their use for strengthening and repair of steel structures has been limited. Previous research has demonstrated that externally bonded conventional CFRP materials can be used to repair naturally corroded steel girders18, steel-concrete composite girders with simulated damage due to partial loss of the tension flange19,20, repair steel-concrete composite girders previously damaged due to overloading conditions21, and to strengthen undamaged steel-concrete composite bridge girders22. The results of the previous research demonstrate that externally bonded normal modulus CFRP materials can be used to increase the yield strength19,22, ultimate strength18-22 and post-elastic stiffness19-22 of steel-concrete composite girders. Additionally, externally bonded normal modulus CFRP materials can be used to restore a significant portion of the lost elastic stiffness due to partial loss of the tension flange18-20. However, due to the relatively low modulus of elasticity of the CFRP materials compared to that of steel and also due to the possible presence of shear-lag effects, the use of conventional modulus CFRP materials did not significantly increase the elastic stiffness of undamaged steel-concrete composite girders22

. The fatigue durability of the CFRP strengthening system has also been shown to be at least equal to that of conventional steel details which are currently used for common steel highway bridge construction23. Recently, high modulus CFRP (HM CFRP) materials have become commercially available which have a modulus of elasticity approximately twice that of conventional steel. A strengthening system which utilizes these HM CFRP materials was developed in a previous research project24, including selection of an appropriate adhesive and testing of large-scale strengthened members. This section of this paper presents the details of an experimental program that was conducted to investigate the fundamental behavior of steel-concrete composite bridge girders strengthened with HM CFRP materials. Further details of the research are available in other publications25. Design guidelines are also presented which can be used to design the required strengthening for a steel-concrete composite bridge girder. HM CFRP Strengthening System

The carbon fibers used in this study were the DIALEAD K63712 high modulus pitch based carbon fibers produced by Mitsubishi Chemical Functional Products Inc. The fibers were pultruded into 4mm thick, 100mm wide laminates with a fiber volume fraction of 70 percent by Epsilon Composite Inc., a French pultrusion company. The modulus of elasticity and ultimate strain of the strips were reported by the manufacturer as 460 GPa and 0.00334 respectively. The strips were bonded to the tension flange of the steel beams using the Spabond 345 two part epoxy adhesive with the fast hardener which is manufactured by SP Systems North America. The tension flange of the steel beam was grit blasted and the surface was subsequently cleaned by air blowing and solvent wiping. After the CFRP strips were installed a wooden clamping system was applied for at least 12 hours until the adhesive had thoroughly set. Experimental Program

The experimental program consisted of two phases to investigate the fundamental behavior of steel-concrete composite bridge girders strengthened with HM CFRP materials. The beams tested in both phases of the experimental program consisted of scaled steel-concrete composite beams which are typical of most highway bridge construction. The typical cross-section of the tested beams is shown in Figure 5. The beams were strengthened with different levels of HM CFRP materials and tested in a four point bending configuration as shown in Figure 6.

Figure 5: Cross-section of a typical test beam

65 mm

525 mm

102 mm

5.6 mm 6.5 mm

200 mm W200x19 steel beam

CFRP strips 2 x 4mm x 38mm or 2 x 4mm x 76mm

3050

roller support

neoprene padspreader beam

pin support

610

x

x

x

x

x - lateral brace location

xx

W200x19 steel beamconcrete slab

1220

hydraulic actuator

Figure 6: Beam test setup

The test matrix for both phases of the experimental program is presented in Table 3. Three beams were tested in the first phase to study the behavior under overloading conditions. Two of the test beams were strengthened with two different reinforcement ratios of CFRP while the third remained unstrengthened to serve as a control beam. All three of the beams were unloaded and reloaded at different load levels to simulate the effect of severe overloading conditions. In the second phase the fatigue durability of the strengthening system was investigated. Two different beams were strengthened with the same amount of CFRP materials but using different bonding techniques. The first beam was strengthened using the typical bonding procedure described previously while the second beam was strengthened using a modified bonding technique which involved increasing the final thickness of the cured adhesive and additionally including the use of a silane adhesion promoter. A third unstrengthened control beam was also tested for the fatigue study. All three beams were subjected to three million fatigue loading cycles with a frequency of 3 Hz. The minimum load in the loading cycle, Pmin, was selected as 30 percent of the calculated yield load of the unstrengthened beams to simulate the effect of the sustained dead load for a typical bridge structure. The maximum load for the unstrengthened beam , Pmin + ΔPU, was selected as 60 percent of the calculated yield load to simulate the combined effect of dead-load and live-load. The maximum load for the two strengthened beams Pmin + ΔPS, was selected as

60 percent of the calculated increased yield load of the strengthened beams to simulate the effect of a 20 percent increase of the allowable live-load level for a strengthened bridge.

Table 3: Test matrix for the three phases of the experimental program Beam ID Reinforcement

Ratio, ρ* Adhesive

Thickness, ta

Concrete Strength,

fc’

Loading

Overloading ST-CONT 0 percent N/A 44 MPa unload/reload OVL-1 4.3 percent 0.1 mm 44 MPa unload/reload OVL-2 8.6 percent 0.1 mm 44 MPa unload/reload Fatigue FAT-CONT 0 percent N/A 34 MPa fatigue: Pmin=50 kN, �PU=50 kN FAT-1 4.3 percent 0.1 mm 34 MPa fatigue: Pmin=50 kN, �PS=60 kN FAT-1b** 4.3 percent 1.0 mm 58 MPa fatigue: Pmin=50 kN, �PS=60 kN

*defined as the ratio of the cross-sectional area of the CFRP strengthening, accounting for the fiber volume fraction to the cross-sectional area of the steel beam

**included the use of a silane adhesion promoter The tensile yield strength and modulus of elasticity of the steel beams were determined by standard coupon tests as 380 MPa and 200,000 MPa respectively. The compressive strength of the concrete used for the concrete deck slabs for the seven test beams was determined from standard cylinder tests after 28 days. The measured concrete cylinder strengths are presented in Table 3. Experimental Results

Findings of the Overloading Study The typical load-deflection relationships of an unstrengthened beam, ST-CONT, and a typical strengthened beam, OVL-2, which were tested in the overloading study are presented in Figure 7.

0

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0 20 40 60

Net Mid-span Deflection (mm)

Tota

l App

lied

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(kN

)

concrete crushing

steel yield

Beam ST-CONT (ρ = 0%)

80

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0 20 40 60 8

Net Mid-span Deflection (mm)

Tota

l App

lied

Load

(kN

)

CFRP rupture

steel yield concrete

crushin

0

g

Beam OVL-1 (ρ = 8.6%)Beam OVL-2

(a) beam ST-CONT (b) beam OVL-2

Figure 7: Load-deflection behavior of the three overloading test beams The load-deflection behavior of the beams was essentially linear up to yielding of the steel. Prior to yielding of the steel, all of the tested beams exhibited minimal residual deflections upon unloading.

However, after yielding of the steel, the unstrengthened beam exhibited a significant increase of the measured residual deflection, as shown in Figure 7(a), while the strengthened beams continued to exhibit minimal residual deflections up to rupture of the CFRP. After rupture of the CFRP occurred the behavior of the strengthened beams followed a similar trend to that of the unstrengthened beam. Table 4 presents the elastic stiffness increase, the yield load, the rupture load and the crushing load for the three beams tested in the overloading study.

Table 4: Comparison of the overloading beams Beam ID Reinforcement

Ratio Stiffness Increase

Yield Load Rupture Load Crushing Load

ST-CONT 0 % - 137 kN - 222 kN OVL-1 4.3 % 27 % 181 kN 259 kN 216 kN OVL-2 8.6 % 46 % 253 kN 358 kN 216 kN

The ultimate capacity of the unstrengthened beam, ST-CONT, was governed by crushing of the concrete while the ultimate capacity of the strengthened beams was governed by rupture of the CFRP. The elastic stiffness, yield load and ultimate capacity of the beams were increased by 46 percent, 85 percent and 61 percent respectively using the higher reinforcement ratio. Inspection of Table 2 indicates that doubling the reinforcement ratio of the applied CFRP, from 4.3 percent to 8.6 percent, approximately doubled the elastic stiffness increase of the beams. Increasing the reinforcing ratio by two times also approximately tripled the increase of the measured yield load and ultimate capacity of the strengthened beams. This demonstrates that increasing the reinforcement ratio did not reduce the efficiency of utilization of the CFRP material.

Findings of the Fatigue Study Three beams were tested in the fatigue study. Beam FAT-CONT remained unstrengthened to serve as a control beam for the fatigue study while beams FAT-1 and FAT-1b were strengthened with a reinforcement ratio of CFRP of 4.3 percent. The two strengthened beams were tested with a 20 percent increase of the applied load range to simulate the effect of increasing the allowable live load for the strengthened beams. All three beams survived a three million-cycle fatigue loading course without exhibiting any indication of failure. Figure 8 (a) and (b) present the degradation of the stiffness and mean deflection respectively of the three beams, normalized with respect to the initial values at the beginning of the fatigue loading program, throughout the three million-cycle loading course.

0.85

0.90

0.95

1.00

1.05

0.0 1.0 2.0 3.0 4.0Number of Cycles (millions)

Nor

mal

ized

Stif

fnes

sFAT-1FAT-1bFAT-CONT

0.90

1.00

1.10

1.20

1.30

0.0 1.0 2.0 3.0 4.0

Number of Cycles (millions)

Nor

mal

ized

Def

lect

ion

FAT-1bFAT-1

FAT-CONT

(a)

(b)

Figure 8: Degradation of (a) stiffness and (b) mean deflection for the fatigue beams All three beams exhibited a minimal degradation of the elastic stiffness of less than 5 percent throughout the three million fatigue loading cycles as shown in Figure 5(a). However, beam FAT-CONT exhibited a nearly 30 percent increase of the mean deflection throughout the three million fatigue cycles as shown in Figure 8(b). Both of the strengthened beams exhibited superior performance with an increase of only 10 percent in the measured mean deflection. At the completion of the fatigue program, the three beams were loaded monotonically to failure. The behavior of the beams was similar to the observed behavior of the three beams that were tested in the overloading study. Proposed Design Guidelines

Based on the findings of this research program a series of design guidelines have been developed which can be used by practitioners to design HM CFRP strengthening for steel-concrete composite beams26. This section presents a summary of the proposed flexural design procedure. The flexural analysis and design of a steel-concrete composite beam strengthened with HM CFRP materials are based on a non-linear moment-curvature analysis. The analysis satisfies the requirements of equilibrium and compatibility and neglects the effect of shear-lag between the steel and the CFRP. This assumption was verified by strain measurements which were conducted throughout the experimental program. A non-linear material characteristic is used to represent the stress-strain behavior of the concrete and the steel while the CFRP is assumed to remain linear and elastic to failure. The nominal moment-curvature behavior of the strengthened section can be established by incrementally increasing the strain at the top surface of the concrete deck and iterating the neutral axis depth until horizontal force equilibrium is satisfied. The nominal moment capacity of the

strengthened section, Mn,S, will typically occur when the strain at the level of the CFRP reaches the design rupture strain of the material. The design rupture strain of the material should be reduced from that reported by the material manufacturer to account for statistical uncertainty of the material properties and possible environmental degradation as outlined in ACI 440.2R-0216. The design ultimate capacity of the strengthened beam, MU,S should be calculated as φMn,S. To account for the brittle nature of failure, a strength reduction factor, φ, of 0.75 is recommended for rupture type limit states27. After rupture of the CFRP, the analysis is similar to that of an unstrengthened steel-concrete composite beam until the crushing strain of the concrete is reached. Based on the proposed moment-curvature analysis procedure, the allowable increase of live load for a strengthened steel-concrete composite beam should be selected to satisfy three conditions. These three conditions are shown in Figure 9 with respect to the typical moment-curvature response of a strengthened steel-concrete composite beam. Due to the presence of the additional layer of HM CFRP material, the yield moment of the strengthened beam, MY,S, is greater than the yield moment of the unstrengthened beam, MY,US. Based on the findings of the fatigue study the total applied moment acting on the strengthened section under service loading conditions, including the effect of dead load, MD, and the increased live-load, ML, should not exceed 60 percent of the increased yield moment of the strengthened section. To satisfy the strength limit state, the total factored moment based on the appropriate dead-load and live-load factors, αD and αL respectively, should not exceed the ultimate moment capacity of the strengthened section, MU,S. Also, to ensure that the structure remains safe in the case of total loss of the strengthening system, the total applied moment, including the effect of dead-load and the increased live-load should not exceed the residual nominal moment capacity of the unstrengthened section, Mn,US.

Figure 9: Load levels and moment-curvature behavior for a strengthened beam

Strengthened beam

MD + ML ≤ 0.6 MYS

αD MD + αL ML ≤ MU,S

MD + ML ≤ Mn,US

Live Load, ML

Dead Load, MD

Factored Moment (αD MD + αL ML)

Strengthened Nominal Capacity, Mn,S

MY,S MY,U

Unstrengthened beam

Strengthened Ultimate Capacity, MU,S

Unstrengthened Nominal Capacity, Mn,US

Curvature

Mom

ent

ACKNOWLEDGEMENTS

The authors would like to acknowledge the support of the National Science Foundation (NSF) Industry/University Cooperative Research Center (I/UCRC) for the Repair of Buildings and Bridges with Composites (RB2C) and the North Carolina Department of Transportation through Project 2006-10. Several industry members made much appreciated donations: Akira Nakagoshi of Mitsubishi Chemical America, Peter Emmons of Structural Preservation Systems and Ed Fyfe of Fyfe Corporation. The authors would like to thank Anthony Miller and Jerry Atkinson for their invaluable help.

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