paper two: design loads and distribution of loads

STRALIAV

NAASRA BRIDGE DESIGN SEMINAR

Paper Two: Design Loads and Distribution of Loads

N.A.A.S.R.A. BRIDGE DESIGN SPECIFICATION

SECTION 2 : DESIGN LOADS

SECTION 3 : DISTRIBUTION OF LOADS

COMMENTARY ON THE REVISED SPECIFICATION

J.E. WHEELER

SENIOR ENGINEER BRIDGE DESIGN

MAIN ROADS DEPARTMENT

WESTERN AUSTRALIA

SU4RY

In 1972 the National Association of Australian State Road Authorities decided that the Highway Bridge Design Speci-fication should be reviewed and, where necessary, revised. Accordingly the Bridge Engineering Committee established a number of sub-committees each with the task of assessing a particular aspect of the Specification. This paper deals with the re-drafting of those Sections relating to Design Loads and Load Distribution, over the period 1972 to 1975, and is a commentary on the changes made therein.

Most significant amongst these was an increase in live loading with the development of new design vehicles and

.

overload provisions. Technical changes were introduced in a number of other areas and overall, an attempt was made to clarify rules and improve format.

o

0 NAASRA BRIDGE DESIGN SPECIFICATION 1.

INTRODUCTION

This paper is by way of a commentary on the revisions made to Sections 2 and 3 by the Sub-Committee. While many portions of the document remain techni-cally unchanged, every effort was made to improve the presentation and format and to remedy the lack of specific directives and coverage of subject matter. Most would agree that the previous specification contained weaknesses exhibited in the form of misplaced rules; dispersion of over-riding require-ments and, occasionally, interpretative problems.

It is the opinion of the author that every design code should fulfil two basic requirements :-

(a) It should clearly state the design conditions to be investigated and the design requirements to be fulfilled.

. (b) It should permit the designer to use rational methods of analysis and design,of any degree of complexity and sophistication he considers to be necessary for his purpose. Additionally it should, wherever possible, provide empirical rules for the inexperienced designer such that they, by the nature of their inbuilt conservatism, ensure the structural integrity of the end product.

It is to be hoped that this new specifTcation goes some way towards satisfying these aims. No code can ever be perfect and most will never even go part way in that direction without a period of trial followed by construc-tive criticism from the users. The obvious constraints of time and money precluded exhaustive research into design philosophies but, where possible, investigative exercises were undertaken as part of the review and re-drafting.

The topics dealt with in this report deal only with technical changes or where new rules and specific directives have been added. To widen the sources from which designers may derive additional information relating to the subjects, further references have been included.

LIVE LOAD

2.1 WHY INCREASE THE LIVE LOAD ?

Throughout the world there is constant pressure from the transport industry to raise both the legal weight limits and those allowed under permit. These overtures can only be construed as part of a search to reduce costs and, while Road Authorities do have limits to finance, it is worthwhile remember-ing that their prime function is to provide a service to the community.

The transport industry, at least in Western Australia, had taken every advantage of loopholes in regulations and relaxation of regulations in particular areas to use multi-axle vehicles in order to carry heavier loads. The implication was that there was a considerable economic benefit to the haulier and it seemed reasonable to assume that these benefits could continue to accrue with the increase in vehicle size. For example, one cannot imagine large mining companies operating uneconomically. The general size of their off-road quarry haul trucks seems to be around 250 tonnes.

2. NAASRA BRIDGE DESIGN SPECIFICATION

0 The general consensus was that there was a need to increase bridge

design loadings. Bridge strengths comprise the most specific constraint to the allowable vehicle loads on highways. Surveys of design loadings throughout the world showed the AASHTO standards to be significantly lower than the majority of industrialist countries'. The pressure of increasing vehicle loadings associated with major industries had already led some State Road Authorities to raise their design loadings* and the increasing use of interstate road transport would continually raise the need for overall standardisation.

Obviously, there is a limit to the amount one can afford to increase design loading at any one time. The greater the increase the greater the number of existing bridges designed to a lesser standard that become affected and may need strengthening or replacement. On the basis that most bridges in our primary road system had been upgraded to HS20, it was felt that the new a

loading should be set at a level that HS20 bridges could take at a reasonable level of overstress. This led to the proposal of approximately 1/3 more than HS20. It could also be argued that a possible heavy degree of overstress might reduce bridge life but this should be acceptable for older bridges.

The Sub-Committee's initial deliberations on the subject, in late 1972 to early 1973, were inspired more by philosophical arguments of this nature rather than by extensive calculation**. The recent NAASRA ERVL Study2 , however, has confirmed that significant increases in axle load and gross mass limits would be justified in economic terms, that these benefits accrue to the whole community, and that load-benefit increase can only continue to go hand-in-hand if they are preceded at a respectable distance by design strength increase.

* See, for example, MRD W.A. design loadings introduced in 1971 (Table i).

** The Committee did undertake estimates of the cost impact of new design load which indicated a national average of the order of 6%. 0

NAASRA BRIDGE DESIGN SPECIFICATION 3.

2.2 THE NEW DESIGN VEHICLES (ARTICLE 2.3.2)

From the outset it was agreed that the loading would remain an AASHTO type vehicle. That is, a tractor truck with semi trailer of variable axle spacing with the same wheel spacing and clearance limits. Amongst other things, this would enable the empirical distribution factors of Section 3 to remain viable. Additionally, the familiar Lane Loading configuration was to be retained*.

The result was that the HS20 main axle load of 144 kN then increased to 192 kN, or a single wheel load of 96 kN. Clearly a problem existed here since the individual wheel load of the HS20 truck had controlled local design of decks and slabs.

& Bridge and culvert spans up to 4 m span were also involved as their design is basically controlled by local wheel load design. The existing 72 kN wheel load provided a design condition which adequately catered for present conditions and for any conceivable increases in individual wheel loads. In order that the new loading would not impose an unwarranted penalty on deck slabs a logical modification was introduced, viz, the main axles of the design vehicle (T44) were converted to tandems. With the increase from HS20 to T44 the B.M., shear and reaction diagrams for various span arrange-ments, as well as the transition point from truck to lane load were, of course, merely scaled up. The introduction of tandem axles made negligible change (for example, see Figure 1).

A comparison of HS2O and T44 stress resultants for some representative span arrangements (Figures 2, 3, 4) shows the regime of each to be clearly defined below 4 m and above 5 m span lengths. This led to the concept of the A14 vehicle, which is no more than the ubusiness en& of the HS20. In fact, one axle controls for all effects below 4 m spans except for hogging moments and the second axle (Article 2.3.2.2 paragraph 2) need only be investigated for such unusual structures as continuous culverts in the 2.125 to 4 m span range. The transition from A14 to T44 dominance takes place in a matter of 1 m increase in span. For designers of lesser experience, perhaps trapped in fear and trepidation at the 4.5 m mark, we have proffered a straight-forward interpolation method.

* Earlier, in Western Australia, we had carried out an exercise which showed that the design vehicle could be converted to an equivalent U.D.L. and concentrated load (after the style of the British H.A.) which was a function of loaded length. This was abandoned as there were difficulties in matching negative B.M. without the use of two concentrated loads and as it could not be extended below 5 in spans. The concept, however, has the merit of simplicity in application and permits easy extension to loadings more appropriate to spans in excess of 120 m. At any future review of the code it should be re-examined as it is a logical means of providing a live load spectrum for all ranges of spans (but needs to be accompanied by

01

a local design requiiement).


2.3 LOCAL EFFECTS OF THE A14 AND T44 VEHICLES (ARTICLE 3.3.4)

For local effects (slabs, slabs between beams, cantilever slabs) A14 has been prescribed as the design vehicle and the empirical formulae of Article 3.3.4 can still be applied*. However, it was considered worthwhile to investigate the local effects of the T44 tandem wheels and, in the light of recent reports 3 ', the conservatism or otherwise of the cantilever slab formulae.

Figure 5 shows the results of an'investigation into the local effects on slabs spanning between beams compared with the effects calculated by the formula of Article 3.3.4.2 (b). Within the range prescribed the T44 effects are well approximated to by the formula. It is perhaps initially disturbing to note that the HS20 itself falls outside the curve below S = 4.27 m (14 feet). However, this is explained by Newmark's contention in his 1949 paper, that the theory is over-conservative in this range of 'S' and should be reduced by about 30%. Some checks were carried out on slabs continuous over beams of representative bridges and beam stiffnesses (using a rigorous analysis). They indicated the formula, with a continuity factor of 0.8, to be adequate and the hogging moments so calculated to be conservative.

The question of cantilever slabs is perhaps more interesting. Our investigations set out to examine three questions

Is the code formula conservative or otherwise ?

Some analyses suggest that sagging B.M. can be significant in the cantilever. Should the code cater for this ?

Does the A14 vehicle cover the effects of the T44 vehicle ?

Appendix A is a dissertation by the author made at. the time in which the predictions of several theories are discussed. The important conclusions to be derived therefrom are that the empirical AASHTO formula is adequately conservative and that a design sagging B.M. of sufficient accuracy could be derived from it.

Although the rules specifically state that the A14 vehicle is intended for local design the question has recently arisen as to what might be defined as the geometric limits for local effects in the transverse direction and, to this end, the experience in a current design in Western Australia serves as an example. The effects of all the Standard Vehicle Loadings were analysed on a 28.5 m, wide single box structure having spans between webs of 7.7 m and cantilevers in excess of 5 m., The results are shown in Figures 6 and 7, clearly indicating the predominance of the T44 truck. The empirical formulae of Article 3.3.4.2 are still conservative but, with structures of this size and importance, designers would invariably employ more exact methods of analysis, particularly when transverse prestres-sing is involved. Ultimately the T44 effects were employed for design, the reasoning being that one was no longer dealing with local effects but rather with major components of the structure in which loading of this order was a practical possibility.

* Note, however, that for slabs with main reinforcement parallel to traffic (Article 3.3.4.1) the appropriate vehicle must be used, according to the span length.


Perhaps some future amendment of the rules would be advisable or necessary. In the interim the author would recommend that, for slabs with reinforcement perpendicular to traffic where the spans between beams exceeded 4 m and/or the cantilevers exceeded 2 m, the transverse effects of the T44 may govern and, if so should be used for design. Clearance from the kerb to the outside wheel should be as prescribed in Figures 2.1 and 2.2 of the rules.

2.4 ABNORMAL LOADINGS (ARTICLE 2.2.3)

AASHTO and NAASRA bridge specifications had always required a design check on the overload capacity of bridges as a provision for infrequent heavy loads. The inadequacy of using a double HS20 vehicle had been recognised in Western Australia for some time and the Main Roads Department had taken steps to improve this situation (see Table 1). Furthermore, the occurrence of special load vehicles nowadays is by no means infrequent, especially on main and industrial routes. The Sub-Committee proposed the use of an Abnormal Vehicle,

S so chosen as to ensure that major structural components in a bridge all exhibit comparable overload capacity under the effect of the type of vehicle likely to be used and to ensure their integrity against damage and unservice-ability by the occasional passage of such a heavy vehicle.

Appendix B is a report, prepared at the time of drafting and this, we believe, clearly demonstrates the need for an overload check (Abnormal Vehicle) of practical configuration and the shortcomings of the double HS20 concept.

It is only fair to point out that the total mass given in Article 2.3.3 is, to a degree, tentative. Various States may have different concepts of what the mass or even the configuration should be and are therefore urged and encouraged to carry out their own investigations into Special Abnormal Vehicles. Table 2, byway of example, is the current MRD WA system of design live load. However, we have recognised that any increase in Standard Vehicle Loading must be accompanied by a reassessment of the Abnormal Vehicle. Appendix C shows the MRD WA proposal currently in . force.

In Western Australia, except on special heavy load routes, we have endeavoured to match the selected Abnormal Vehicle mass to the strength available in the main structural components designed to the Standard Vehicle Loading.

2.5 MINIMUM BRIDGE LOADINGS (ARTICLE 2.3.4)

It was appropriate that the allocation of full design live load or a reduced level (for "other's bridges) be aligned with the NAASRA functional road classification and this has been done. Table 2 shows the further sub-classification of this in Western Australia.

2.6 STANDARD DESIGN LANES (ARTICLE 2.3.5)

The notion of design traffic lanes and load lanes in the previous rules was often a source of confusion to new designers and, undoubtedly, was over complicated. The application of two lines of vehicles, for example, each confined to its own design traffic lane which can stretch, as the width between kerbs increased from 6.1 m (20 ft) to 9.2 m (30 ft),does not seem


logical. Although we have not dispensed with the idea of design traffic lanes it is now easier to apply*.

The transverse positioning of design vehicles when calculating load distribution is important and the former Load Lane constraints were difficult to accept as practical concepts for design.

Presently, with availability of computer programs and the speed of calculation it is becoming more common to analyse the transverse distribution of stress resultants, rather than use the empirical factors in Section 3**•

2.7 IMPACT AND DYNAMIC BEHAVIOUR UNDER LIVE LOADS (ARTICLES 2.3.7 and 2.19)

The study of the dynamic response of bridges to traffic is a complex subject and depends upon the vibration characteristics of the structure, the relative mass of the vehicle and structure, the vibration characteristics of the axle suspension system and the roughness of the approach and deck surfaces***. Obviously the effects of all but the first parameter tend to be random in nature while the last would almost certainly vary over the life of the structure.

Factors which are of major concern in the response of bridges are' :-

The dynamic contribution to peak static stresses for considerations of strength.

The dynamic contribution to stress ranges for considerations of fatigue life.

The dynamic contribution to the deflection-time history for consideration of human reaction to bridge movement.

The vibration of bridges as a phenomenon is a question of matching the natural frequencies of the superstructure and the vehicle. The impact (i.e. the ratio of the dynamic and static amplitudes of deflection) is not really dependent on the mass of the live load but rather of the frequency match between vehicle and structure and the state of excitation and velocity of the vehicle when entering the bridge.

Considerations of dynamic behaviour must be directed along two paths. Firstly, adequate provisions for the integrity and life of the structure (a, b) and, secondly, an examination of the dynamic system in terms of pedestrian (and perhaps stationary vehicle passenger) reaction (c). On the face of it, it is difficult to accept a formula which is simply a function of

* In the formula given, the denominator was originally 3 to coincide with the concept of a 3 m standard design lane but, during the review period, it was requested that this be changed to 3.1, thereby ensuring that the commonly used 9.2 m (30 ft) between kerbs structure did not ttcop?? an extra lane.

** In retrospect it is felt thatwe could have dropped all reference to "lanes" in favour of numbers of vehicle lines in a given width between kerbs couple'd with a statement of minimum clearances.

The research literature seems to indicate that surface and joint irregularities are the most important factor, which stresses the need for good joint design and maintenance.

0 NAASRA BRIDGE DESIGN SPECIFICATION 7.

span length to cater for the first consideration. In fact, some researches6 have shown the impact factor to increase with span over a range and then decrease, to be different for continuous and simple spans, and generally to be well in excess of the 30% maximum used by AASHTO. Recent investigations 7 also indicate that peak dynamic forces occurred mostly at the start and end of outer spans with a predominance of higher values at the first support.

To date, no alternative is offered (at least not a simple one). However, we may take comfort in the thought that the present provision for impact is probably not unconservative on two scores

(a) The vehicles which induced the large impact factors measured were single vehicles or two vehicles running in phase. It could be expected that, normally, heavy vehicles would occur in a random sequence and, in traversing out of phase, there is a tendency to negate each other's energy input.

. (b) The impact factor prescribed by the rules is applied to the total design live load and likely to produce a dynamic increment larger than that of a single heavy vehicle. Abnormally heavy loads would, in practice, be travelling at very low speeds and the 10% factor prescribed should be quite adequate.

The investigation of the structural effects of wind on such exotic structures as suspension bridges, and others that may be aero-dynamically sensitive, is a specialist subject. Reduction of wind excitation can be achieved by increasing the structural stiffness and structural damping. Additionally, aero-dynamic design is used to reduce aero-dynamic forcing. It is unlikely that these effects would ever need to be investigated for the usual types of bridges but should designers wish to pursue the subject they are referred to References 8, 9, 10.

The question of human reaction to vibrations in bridges covers both road bridges with footwalks and pedestrian footbridges. The complete reso-lution of the problem depends upon being able to solve the equations of motion in structures, leading to values of peak acceleration and displacement, and then to relate these to various criteria* for human tolerance to such effects. In the case of footbridges the exciting force (the pedestrians themselves) is not so easy to quantify but the results can be just as disturbing. It has been said (of people and structures) that "when a motion is unexpected and suspected to be a symptom of structural inadequacy, it's perception alone is disturbing".

A survey of the literature is interesting but not over helpful to the designer's needs. There are, however, a few empirical criteria which can be applied and these are to be found in References 12, 13, 15, 16, 17 and 18. Reference 6 provides some of the bases for the requirement in Article 2.19 relating to natural frequencies.

There are several formulae in the references for calculating natural frequency of structures but, for the more interested, and for haunched struc-tures, the Rayleigh-Ritz method is straightforward in application 19 '20

* There is a fairly extensive literature on the subject but, for example, see References 11, 12, 13, 114.

NAASRA BRIDGE DESIGN SPECIFICATION

0 2.8 LONGITUDINAL TRAFFIC FORCES (ARTICLE 2.5)

For some time there has been general agreement that the allowance of 0.05g was too low and that braking accelerations could even be as high as ig in some vehicles. A NAASRA survey was carried out some years ago but did not result in any specific recommendation. World standards, with the exception of AASHTO, were of the order O.2g to 0.25g. While on the one hand 0.05g was unrealistically low, a compromise of O.lg was settled on, since there had been little evidence of damage to bridges from braking forces.

It was agreed that relating the proposed horizontal accelerations of O.lg to the Lane Load would be appropriate but that there should be limits so that short bridges were not under-designed and long bridges were not unduly penalised. A lower limit of 70 kN (approximately 20% of H520) and an upper limit of 270 kN was set. This latter corresponds to a 200 tonne vehicle coming to rest from 19.3 km/hour (12 mph) in 9 m (30 ft).

The idea of this force being applied at 2 m above deck level was dropped because all such forces are frictionally applied at deck level and the height of the vehicle C.G. merely produces local differences in the tyre reactions. The same reasoning could be applied to centrifugal forces but, in this instance, it was retained since in single lane - single pier struc-tures a torsional effect can be induced which may be significant.

2.9 FOOTWAY LOADING (ARTICLE 2.3.8)

The essential distinction to be made here is between footway floors and floor members (local design) and footbridges or main footway members (global design). For the former the 5 kPa loading is considered necessary, as well as for the occasional footbridge, so located that on some occasions ('regal' or 'rock') it could be jammed with excited onlookers.

For the remainder (footbridges and sidewalks on bridges) the loading is a function of span and width, although it is considered rather remote a possibility that a suburban footbridge would receive its full design live load*.

3. KERB AND BARRIER LOADING (ARTICLE 2.6)

The specification, in the past, has been aligned with the AASHTO loading for various types of barrier. It was therefore logical to bring it up to date with the latest and more comprehensive AASHTO standards. Design by this standard (elastic method) results in members not very different from those of the British standards (plastic method). An extensive bibliography and survey on the subject is contained in Reference 21.

* A fruitful area in which to use partial prestress design.

NAASRA BRIDGE DESIGN SPECIFICATION 9. . 4. WIND FORCES (ARTICLE 2.7)

During the review period AS 1170-Part 2, a new Australian Standard on wind speeds for structural design, was issued, which provided a much more rational basis for establishing a design wind speed in any given area. In keeping with general policy to make maximum possible use of such standards the Committee decided that it should be adopted. However, the combination of conservative wind pressures* and, in some areas, increased wind speeds proved to be unacceptable.

It was therefore decided to give designers an option of using quick conservative values or of using more exact methods where required22 .

In the light of the recommendations of the Australian Standard for cyclonic areas (and of recent events) the relevant design wind speed in Table 2.3 was also increased.

5. THERMAL EFFECTS (ARTICLE 2.8)

In the last decade or so much more attention has been given to the effects induced in bridges by changes of temperature, and especially where gradients can exist throughout structural members. Review of this Article was along two lines. First, to see if the average bridge temperatures, which relate mostly to overall movements and allowances, could be refined. Secondly, to introduce the necessary requirements to cater for induced stresses due to temperature differential. These stresses have been demonstrated to be as high as live load stresses and, although they are unlikely to be an important influence on the ultimate load capacity, they are worthy of consideration for serviceability checks into concrete cracking, bearings and joint loads.

The classification of climatic conditions in the existing rules was rather too vague and some research was put into determining area classi-fications and temperature ranges. The temperature gradient diagrams have been based largely upon the British requirements. Appendix D is a copy of the detailed commentary by Mr. G. B,ott issued at the time. Additional to the references contained therein, References 22, 23 and 24 give methods of calculating the stresses due to temperature differential.

It is recognised that the new rules have relied to a fair degree on the work of Mary Emerson of the Road Research Laboratories, that the regional classifications still relate to fairly large areas within which variations may occur, and that the relationship between air shade temperature to structure temperature is still the subject of investigation for Australian conditions. Recent measurements on a concrete box girder bridge in Western Australia have indicated that differentials given in the new rules are reasonably correct. Also that, while bituminous surfacing might be regarded as an insulant, it absorbs radiation to a greater degree and the two effects tend to cancel. On the other hand, diurnal temperature records for the

*

The AASHTO wind pressures, based on 100 mph, imply a drag coefficient of approximately 2.0.


S locality, taken over 70 years and coupled with shade temperature-deck temperature readings show the range to be 75% of the value given. The designer should be prepared to take advantage of such records, where they are available, as a means of incorporating economies in the expansion joints for long bridges.

6. FRICTION FORCES (ARTICLE 2.12

The major change here is in the values for friction coefficient listed in Table 2.5. In his commentary (Appendix E), Mr. J.K. Lloyd records that proposals were strongly influenced by Reference 25 which stresses the need to use friction coefficient values that allow for change in characteristics with age and state of maintenance. A further survey of PTFE bearings is contained in Reference 26.

7. EARTHQUAKE FORCES (ARTICLE 2.13)

Fortunately, Australia does not lie on a tectonic plate boundary and is not therefore liable to the frequent and often disastrous earthquakes typical of those regions. However, 5% of the world's earthquakes are of the intra-plate type (shallow depth with high local intensity) and Australia has experienced six of intensity 5.5 to 6.8 (M.M. Scale) between 1941 and 1973. While damage to bridges resulting from these has either not been detected or of great consequence, there is certainly a good case for assessing the risk of occur-rence of significant earthquakes in any particular area and for taking adequate precautions in detail design. Furthermore, in the period 1969 to the present, the National Committee on Earthquake Engineering has progressed a considerable way towards establishing an Australian Standard on the subject27 '28 . In view of this we felt that it was appropriate that the rules be brought in line with current approaches to seismic engineering with some amplification on important detail design aspects*.

The question of seismic design rules is divided into two aspects :-

The general problem of identifying those areas in which earthquake risk tends to indicate earthquake resistant design as a prudent or necessary undertaking.

The engineering problem of providing structures with earthquake resist-ance of a sufficiency to prevent collapse (but usually allowing some damage, depending upon the severity of the earthquake).

* In view of the leading part that U.S.A. and, in particular, California, has played in developing rational approaches to seismic design (UBS and SEAOC) it seems surprising that the AASHTO rules were not amended

29 until 1975.


0 In the first instance we have been guided entirely by the draft SAA

Code. At the time of drafting Article 2.13, South Australia and Western Australia were the only areas in which some confidence was expressed in the assessment.

However, the SAA Draft has now progressed to the stage where bridge designers in other States would now be recommended to consult Appendix F of Reference 28 as a guide to seismic risk areas.

It is interesting to note that both the SAA and AASHTO now recognise the effect that the type and depth of soil over bed rock have in modifying the motion. Table 3 is a calculation of design horizontal seismic force using Article 2.13, AASHTO, and the SAA draft. This shows comparable results, the AASHTO being higher (since it stems from a code applicable to a plate boundary area).

On the subject of earthquake loading and design there is an ever- growing literature but we feel that the new specification gives an adequate coverage in the references supplied. While the risk in Australia is not high by other standards, it is of sufficient order in some areas for designers to pay due attention to detailing in bearings, hinges, joints, connections, etc*. It is interesting to note that not all codes prescribe vertical seismic forces and, in Reference 19 of Section 2 in the new Specification, the absence of such a requirement is deemed an important deficiency. We felt it was worthy of inclusion and in line with the SAA Draft Code.

8. SHRINKAGE AND CREEP (ARTICLE 2.14) DIFFERENTIAL SETTLEMENT (ARTICLE 2.15) CONSTRUCTION FORCES AND EFFECTS (ARTICLE 2.16)

The essential aim with these three items was to bring them to the notice of the designer by means of a general description of the effects. The calcula-tion of shrinkage and creep in concrete is adequately dealt with in Section 6 and we merely point out the various situations under which stresses or deflections could be significant. Differential settlement may or may not be important and it is up to the designer to satisfy himself on this point. The same is even more true of construction forces and effects. In addition to loads of a temporary nature the trend towards methods such as launched or cantilever construction introduces stages of loadings which can be quite significant.

* References 24, 30 and 31 contain valuable contributions to this subject. The commentary in Reference 24 (pages 24, 25) makes some important points as to the risk factor for different types of bridges. Reference 33 evaluates bridge types which are amenable to the simplified analysis set out in the rules compared with those for which a more complex dynamic analysis should be carried out.


9. ALLOWABLE STRESSES AND COMBINATION OF LOADING (ARTICLES 2.1 AND 2.17)

Bridges can be subjected to a much wider variety of loadings in their useful lifetime than most other structures. Obviously more than one effect can occur at any one time and the examination of possible combinations must be carried out with the following criteria in mind :-

The probability of occurrence and the probability of simultaneous occurrence.

The duration of each loading.

The significance of each load on various parts of the structure.

The accuracy with which each load can be calculated.

Depending upon the probability of concurrent application how closely the limits of serviceability and ultimate capacity may be encroached upon. This implies a determination of specific design limitations and a range within which stresses may vary (the lower bound being the Basic Allowable Stress).

It was decided that there was no justification for departing from the basic AASHTO arrangement* of group loadings except for the modifications which are listed below

The permanent effects from construction sequences and methods and from differential settlement are to be included in Group I.

A note is made that differential temperature effects be considered for inclusion in Group I and X for prestressed design. It is, however, the author's opinion that Group IV adequately covers the combination of live load and temperature effects.

Group X, the Abnormal Vehicle overload check, has been added with an allowable overstress of 140%. 0 In Group III, braking force and centrifugal forces are alternatives as they would not be likely to occur simultaneously at peak values in the design vehicle.

The item for elastic (rib) shortening, R, has been dropped. The only elastic shortening of any significance in modern structures is catered for in prestress effects.

* Ideally, in the author's opinion, the individual effects should be categorised into permanent and transient types. Design cases would then consist of any.logical combination of permanent and transient loading that could occur. Each effect could be weighted according to its probability of occurrence and, for any combination, an overall probability (and threby the allowable stress) determined. New

23 '32 Zealand has progressed some way towards this aim.


Group VII now constitutes a check on the serviceability of the structure components under earthquake loading and is, more or less, in line with AASHTO.

An inconsistency existed in the previous rules, between Articles 2.2.2, 2.24 and 3.3, in cases where high fill on culverts permits the designer to ignore live load and he is invited to invoke a 30% overstress allow-ance. This should have been 25% as Group II would have been the relevant case, not Group VII.

Article 2.1 now lists the specific exceptions to the rules for combining loads, over and above theincrease in basic allowable stresses which takes into account the improbability of simultaneous occurrence and (for steel) safeguards against flexural-torsional buckling. These exceptions are

Prestressed design where stress limitations are based on other considerations.

Elastomeric bearings where an ultimate condition (total shear strain) can be over-riding.

Elastic instability of any structural component. While this possibility is catered for in steel and prestressed concrete design, a general state-ment is considered necessary to cover design in general. Formulae,which are intended to preclude buckling, generally have an in-built factor of safety of the order of 2.

Overall stability of structures against overturning, uplift and sliding.

OVERTURNING. UPLIFT AND SLIDING (ARTICLE 2.18

The previous rules covering this consideration related only to attachments (Article 2.22) and abutments (Article 4.8.1). The new rules are set out as a general statement for all structures aswell as for individual parts. The classification of stabilising and overturning forces is now divided into permanent loads, live (traffic loads) and other transient loads with factors which are intended to reflect accuracy of calculation and importance. The designer is left to select rational combinations of these in checking over-turning or sliding and safety of connections. In Appendix F, Mr. J.K. Lloyd's commentary indicates how the provisions of this Article were derived.

HIGHWAY DESIGN SUPPORT AND LIGHTING STRUCTURES - (ARTICLE 2.20)

While this Specification is mainly concerned with bridges and major structures, it should be capable of being used for the design of other highway "hardware". In freeway situations this is particularly true of large signs, overhead gantries, and high mast lights. The most comprehensive and relevant specifi-cation is undoubtedly that issued by AASHTO and this has been adopted.


S The rule states that designs "shall comply with the relevant* pro-

visions" of the AASHTO standard. In this regard there are some aspects which are worthy of mention

As would be expected, wind forces constitute the major loading on signs and light standards and the design wind speed represents a 3 second gust which has a return period of 50 years. The usual approach** with wind loading is to assume that this can be represented by Bernoulli's equation for steady flow, F = C.A.p.V2 , where C is a coefficient depending on body shape and height above ground. In the AASHTO formula the wind speed is found from a map of isotachs and multiplied by 1.3 to represent a gust. In the Australian Standard*** the wind speed so determined is, in fact, the 3 second gust speed for 50 year return interval and should therefore replace the term (1.3V).

In high mast lighting and some overhead signs resonant vibration of the structure can coincide with the periodic forces associated with vortex shedding. These vibrations are undesirable in signs and have caused electrical equipment damage in high masts. The so-called "critical wind speed" at which this occurs is usually much lower than the design wind speed. On the Narrows Bridge, Perth, this speed is about 30 mph and causes a continuing maintenance problem with luminaires.

Estimation of the critical wind speed and the forces involved can usually only be carried out by making rather sweeping simplifying assumptions. Based on past experience some "rules of thumb" have been used to check the adequacy of structural stiffness in preventing vibrations in signs and lights. For overhead sign structures (span type) the AASHTO recommends dead load vertical deflection be limited to d2 /122 (where d is the sign depth in metres). For high mast lights "good engineering judgement" combined with experi,ence is suggested. For large post-mounted signs (i.e. cantilever) the M.R.D. W.A. limits the deflection at the centre of the sign face (h above ground level) under maximum design wind speed to h/200. For some time in Western Australia, we have prescribed that the deflection at the luminaires under the maximum design wind speed be no greater than 1/20 of the height and it will be noted that this is mentioned in the latest (1975) issue of the AASHTO specification.

These criteria, in effect, impose minimum stiffness limits. However, in some instances, the solution to wind-induced vibration problems is to reduce stiffness and some overseas high mast manufacturers employ a flexible insert in the foundation bolts. The basis of this approach is to reduce critical wind speed and induced cyclic forces to a minimum, thereby also reducing fatigue risk.

* That is, loading and structural design.

** A more exact approach is to calculate the wind loading taking into account the structural response to the gust (dependent upon natural frequency and damping) and this forms the basis of a draft British Standard for high mast lighting.

It is implied that design gust speed (1.3V) should be determined from AS 1110-Part 2, Table 2, and that terrain and height corrections are catered for in Table 1.2.5B of the AASHTO specification.


0 12. DISTRIBUTION OF LOADS

12.1 GENERAL

In the discussion on Standard Design Lanes, it was suggested that, with the facilities now available to the designer, it was becoming more common to carry out an analysis of the transverse distribution of vehicle effects rather than use the empirical factors of Section 3. In a current survey of bridge design engineers in the M.R.D. WA., it was found that no-one ever used the empirical factors for final design. Certainly they are employed for quick preliminary designs and comparative estimates, and additionally the local design formulae (Article 3.3.4.2) are used for final design in some structure types. It is suggested that, with the passage of time, Section 3 will be relegated to the role of quick estimate aids. It is for this reason that general directives and a description of the effects to be investigated assume more importance. The distribution factors, however, remain a viable means of design provided the limits of application are made clear.

12.2 LIMITATIONS TO ARTICLE 3.3.2

It had always been suspected, and was later confirmed, that the AASHTO factors assumed that wheels could not fall outside the exterior stringer. This is implied in the statement (AASHTO 1.3.1 (B) 2 (a) ) that "live load bending moment for outside roadway stringers or beam shall be determined

.......assuming the flooring to act as a simple span between stringers or beams". It can be demonstrated by rigorous analysis that in structures where the wheel does fall outside the exterior beam the distribu-tion factor can be quite unconservative* (naturally depending on the distance involved). In the interests of safety and uniformity the rule is now quite specific. The exception to this is the multi-cell concrete box girder which achieves considerably better distribution from the torsional stiffness of the cells. In this case the limit is set at half average beam spacing.

Lateral distribution of wheel loads in calculating end shear and reactions is now disallowed over a specific distance rather than the pre- . viously used "adjacent". In M.R.D. W.A., this rule is extended to allow a

451 dispersion angle outside that distance and up to ¼ of the span and uniform distribution beyond ¼ span. This irule is somewhat arbitrary but on the conservative side.

The footnote to Article 3.3.2 may need some elaboration. For beam and slab bridges the original work on which the AASHTO factors are based did not include intermediate diaphragms and therefore, if the factors are used, there is no advantage to be gained in using diaphragms. However, if the designer intends to improve distribution by the use of diaphragms then they must be properly analysed and designed.

In Reference 34 a cautionary note is sounded for concrete box girders where the AASHTO factors are shown to be slightly unconservative towards the shorter span range. Analyses of composite box girder bridges (Article 3.3.2.1) carried out in Western Australia have also shown the factors to be quite unconservative when wheel loads can fall outside the centreline of the exterior web of the outside girder.

* Particularly in composite steel girder bridges.

16. NAASRA BRIDGE DESIGN SPECIFICATION . The factors are not applicable to wheel configurations other than the

standard Vehicle Loading (e.g. the Abnormal Vehicle). Also the provisions of Article 2.3.6 do not apply when using the factors for multi-lane bridges.

12.3 SLABS AND MULTI-BEAM DECKS (ARTICLE 3.3.4)

The applicability of the empirical factors for locat slab design (Article 3.3.4.2) has been dealt with earlier in this paper where it was demonstrated that slabs between beams and cantilever slabs (perpendicular reinforcement) could be designed with confidence under the new Standard Vehicle Loading.

We felt it was important to include the note on multi-beam bridges. If the elements (beam units) are transversely prestressed so that there is moment as well as shear transfer, then the deck is approaching the case of an integral slab and could be treated as such. However, in the case of trans-verse bolts and bars or shear keys, then a rational analysis is necessary. Additional to the references in the Specification, References 34, 35, 36 and 37 are all applicable to the subject and set out methods of analysis (Refer-ences 34 and 37 are suitable for hand calculations).

The formulae for distributive reinforcement (Article 3.3.4.3) have been removed and transferred to Section 5 where they logically belong. It is important to remember that these expressions applied basically to right bridges and it is generally recognised that skew slabs need to be looked at carefully to ensure that serviceability as well as ultimate requirements are met38 .

13. CONCLUSION

This major review and re-drafting of the Bridge Design Specification must represent something of a milestone in the life of NAASRA. The first mile-stone was, of course, the initiation and issue of a uniform Australian Speci-fication between 1949 and 1953. Not only did this later exercise establish personal contact between the practising level of Bridge Engineers throughout the authorities, but it permitted the exchange of a wealth of information and points of view on many other aspects of bridge and road engineering. This interchange of ideas must be encouraged and maintained and it is hoped that the NAASRA Bridge Seminars will serve this cause. It is important too that the Review Sub-Committees will not be disbanded. Regular review and amend-ment is essential if the Specification is to be maintained as a working document.

In some respects, we must look on the Specification as an interim one because it is still largely tied to the elastic design approach. The trend is towards Limit State design and, to this end, a start has already been made with a committee to study the concept and for an ARRB research project into live load spectra. Undoubtedly the advent of Limit State design will have to be preceded by the establishment of more representative Standard Vehicle live loading and by a re-appraisal of loading combinations and overstress allowances.

0

NAASRA BRIDGE DESIGN SPECIFICATION

17.

ACKNOWLEDGEMENTS

The author gratefully acknowledges the co-operation and assistance of his fellow committee members :-

Mr. N.C. Flaylock, Country Roads Board, Victoria Mr. J.K. Lloyd, Department of Construction, Melbourne Mr. G. Bott, Department of Main Roads, New South Wales

Thanks are due to all of the Bridge Engineers in the State Road Authorities who reviewed and constructively criticised the draft.

The author thanks Mr. J.G. Marsh and Dr. K.C. Michael for their interest and for many fruitful discussions on the subject.

The permission of Mr. D.H. Aitken, Commissioner of Main Roads, Western Australia, to present this paper is acknowledged.

FIGURE 1

I --

50 100 150 (1524) (3048) (45-7)

COMPARISON OF SINGLE AXLE & TANDEM - AXLE VEHICLE FOR INCREASED LOADING -

T44 LANE /

z 7

LL

cd ii

- 0 MS.18 x 133

SPAN 'C- FT(m

4000 (2950

3000- (2213)

2000- (1475)

1000- (738)

FIGURE 2

19.

COMPARISON I I I I I I

OF MAXIMUM EFFECTS DUE• TO T.L4 & A.11 (SIMPLE SPANS

E

I-

m CD

T.'

< , z

//III

10

SPAN'L'-

5 15 20Ft

150- (110 6)

100- (738)

50- (36.9)

0

FIGURE 3

20.

-COMPARISON TO

I I I I I I I OF MAXIMUM EFFECTS DUE

T44 & A.11. (CONTINUOUS SPANS) SAGGING BENDING MOMENT

A

- E-

z

U- I

cc All.

I______

• _ _ _ _ _ / 1 _ _ _ _

/.

/

12 I SPAN'1 /-'

2Ft I I I

2 3 4 5 6 m

150- (1106)

100- (738)

50 (361.)

0 -

C

FIGURE 4

I I I I I

COMPARISON OF MAXIMUM EFFECTS DUE—TO T.4 & A.14 (CONTINUOUS SPANS)

HOGGING BENDING MOMENT __

A A

T.i4- E

LL

I / /

,( /

m /

(D z CD 0

_21Z

SPAN 'L

20Ft

100 (738)

50 (369)

:si

150- .

fl 21.

14 (1033

12 (885

10' (738)

8 (590)

6 (41.3)

4 (295)

2 (148)

0

32

r - 6 6

. A 1

- AA (AlL.)

-E CUR'EA'.-CURVEB-__CURVE C- - -. - - - z

..J

M S.18 (4.14)

T •-- =T.L./. LL

---- ---- - -. -. - - - -

(A. -144

TNOTE: Calculated by Westergaard Theor

(S+2)16 32

'S'Ft (m)

2 4 6 8 10 12 14 16 18 20 22 -n

24 - (061) (120) (183) (21.4) (3-05) (366) (427) (488) (549) (610) (671) (731) 0

C

COMPARISON OFT.4L & MS.18 EFFECTS m

MAXIMUM TRAVERSE SLAB B.M. ( UNDER CRITICAL WHEEL) U,

SLAB SIMPLY SUPPORTED OVER BEAM SPACING S

r

E 23•

FIGURE 6

MAXIMUM TRANSVERSE B.M. DUE TO STANDARD

\ VEHICLE LOADINGS- \ SLAB BETWEEN BEAMS

WITH LARGE SPAN

04

00

I Hf

__

(WN) 1N31610V4 ONION38

24.

FIGURE 7

MAXIMUM TRANSVERSE B.M.

DUE TO STANDARD VEHICLE LOADINGS- CANTILEVER SLAB WITH

1

Fl LARGE SPAN

19

~-A

CLASS LOCATION OF BRIDGES STANDARD LOADING ABNORMAL VEHICLE SPECIAL LOADING

1 Freeways, Main Roads, Important HS20 + Impact in all lanes as 25 units of load + 10% Secondary Roads and Metropolitan per NAASRA Bridge Design Speci- Impact centrally placed Arterial Roads other than those fication at working stress at 25% overstress and - in Class 2 3V6 tension in pre-

_______ stressed_concrete

Main Industrial Roads in HS25 + Impact in all lanes as 30 units of load + 10% Metropolitan area per NAASRA Bridge Design Speci- Impact centrally placed

cation at working stress 2 at 25% overstress and - 3/F'c tension in pre-

_______ stressed concrete 2

3 Main and Important Seconday Roads HS25 + Impact in all lanes as 25 units of load 4 + 10% Road train of six 16 ton north of Geraldton and roads in per NAASRA Bridge Design Speci- Impact centrally placed tandem axles at 14 ft. mining areas subject to haulage fication at working stress at 25% overstress and centres + Impact applied of ore 3v7F'c tension in pre- anywhere on the structure

stressed concrete without concurrent loading at working stres S3

4 Developmental Roads, Forestry HS20 + Impact in all lanes as Roads and Residential Roads per NAASRA Bridge Design Speci-

cation_at_working_stress

5 Tourist Roads, National Park Each case on its merits - usually Roads and similar roads HiS as per NAASRA Bridge Design - -

Specification_at_ working _stress

NOTES. 1. The HS25 loading is 25% greater than the HS20 loading given in the NAASRA Bridge Design Specification with the exception that the wheel load is unchanged when local wheel efects are being investigated. F'c is the 28 day concrete compressive strength based on a standa-d cylinder test. A tandem axle in this instance is defined as two axles spaced 4 f:. apart. 200 ton total load 240 ton total load

m

BRIDGE DESIGN LOADING - M.R.D.. W.A. 1971 - 1975

N) C'

NAASRA STANDARD

ROAD DESCRIPTION CLASSIFICATION HIGHWAY ABNORMAL VEHICLE SPECIAL LOADING

LOADING (S.H.L.)

A. Freeways, Main Roads, 1, 2, 3, 6, 7 A14, T44 plus Impact in Standard Abnormal Vehicle Secondary Roads Standard Design Lanes, of total mass 204 tonnes

Nil. Loading Groups I to IX (250 kNper axle) plus inclusive Impact. Group X loading.

Centrally placed.

B. Main Industrial Roads 7 Standard Abnormal Vehicle in the Metropolitan of total mass 245 tonnes Nil area (300 kM per axle) plus

Impact. Group X loading. Centrally placed.

C. Main Roads and Second- 1, 2, 3 Standard Abnormal Vehicle Road train of six 165kN ary Roads north of of total mass 204 tonnes tandem axles at 4.25 m Geraldton and roads in (250 kN per axle) plus centres. Impact as for mining areas subject to Impact. Group X loading. S.H.L. Applied anywhere haulage of ore Centrally placed. on structure without con-

current loading. Loading Groups I to IX inclusive *

D. Main Roads and Second- 1, 2, 3 Road train of four 165kN ary Roads south of tandem axles at 4.25 m Geraldton subject to centres. Impact as for regular haulage associ- S.H.L. Applied anywhere ated with specific on a structure without industries, concurrent loading. Loading

Groups I to IX inclusive *

E. Developmental Road, 4, 8 A14 and 75% T44 plus Forestry Roads and Un- Impact in Standard Nil Nil classified Roads other Design Lanes. Groups than in (F) I to IX inclusive

F. Tourist Roads, National 9 Each case assessed on Park and similar roads its merits but generally Nil Nil

not less than 60% of A14_ and _144

* A tandem axle in this instance is defined as two axles 1.2 m apart.

BRIDGE DESIGN LOADING - M.R.D., W.A. 1976

. S . 0 1 S

792

--H

27.

W] TABLE 3

HORIZONTAL SEISMIC FORCES

STRUCTURE :- R.C. Slab - two spans of 6.7 m (22 ft.) Overall Width - 7.92 (26 ft.) Slab - 300 (12 in.) Piers - three 360 x 360 P/S piles, estimated

fixity 5.18 m (17 ft.) below deck level. Tops of piles integral with deck slab

Total weight 1205 kN (271 kip)

Horizontal Stiffness - 7.99 kM/mm (longitudinal) 6.53 kM/mm (transverse)

EARTHQUAKE RISK :- Zone 2.

NAASRA 1976 - ARTICLE 2.13

I = 0.065 = 05

0.798 secs; C = = 0.054

EQ= K.C.D. = 0.67 x 0.054 x 1205 = 43.6 kN

AASHTO INT. 1975

ZONE II, A = 0.22g

Combined Response Coefficient 'C' = 0.06 (assume minimum)

F = 0.08

EQ = C.F.W. = 0.06 x 0.8 x 1205 = 57.84 kN

S.A.A. DRAFT 1976

ZONE 2, Z = 0.375; I = 1.0 K = 0.67

cs = 0.14 (Assume maximum)

H = Z.I.K. CS.W = 0.375 x 1.0 x 0.67 x 0.14 x 1205 = 42.39 kN


APPENDIX A

CANTILEVER SLABS WITH REINFORCEMENT PERPENDICULAR TO TRAFFIC

The existing code (Article 3.2.8(a) ) prescribes that, for each wheel line, the B.M. per unit width of slab is

M = PX/E kNm

where E = 0.8X + 1.1, metres

X = distance of load to cantilever root, metres.

(The code states that this provides for other preceding and trailing wheels in the line).

The values of maximum root B.M. for a point 1ud, as indicated by the various theories, are shown in non-dimensional form in the attached graph (Figure A.1).

From a design point of view all theories imply that, with the load over the cantilever root, a B.M. of n/il can exist, and with the load at the edge of the cantilever, a root B.M. of about 0.5P can exist even with a canti-lever length of minute proportions. Engineering intuition suggests that neither can be correct and that, in the first instance, the root B.M. must be zero and, in the second instance, the root B.M. would probably a:ymptote to a constant value when the cantilever moment arm reached proportions at least greater than the thickness of the slab. Further, practical loads are patch loads rather than point loads, and additionally because of the necessity to have kerbs and parapets and respect a 1 ft clearance from the kerb, the canti-lever is never loaded at its extremity. The theories also consider the cantilever as encastre, a condition which would rarely occur in practice. Sawko and Mills (in replies to discussion at the Conference** which is published separately), admit that objections along the above lines are valid, especially that the root B.M. tends to zero as the load eccentricity tends to zero. The fact that the root B.M., for small eccentricities tends to a value P/il, need not have any practical significance as this peak value would exist over such a small length as to be instantaneous stress concentration rather than a design moment.

Using classical plate theory (M.R.D. "Tee Beams" programme) the effects due to A14 and T44 have been studied for cantilevers up to 4 m with exterior wheel line placed at 1/6, 2/6, 3/6, 4/6, 5/6, 6/6 of cantilever length 'a' * . With values of '' and 'a' less than about 0.5 metres the convergence of the solution was very slow but gave indications of tending towards P/n .

* See Figure A.2. ** See Reference 3.

NAASRA BRIDGE DESIGN SPECIFICATION 29. . The second graph (Figure A.3) shows a plot of the results, which are summar-ised as follows :-

Beyond a = 1.5 m the T44 provided greater root B.M. than the A14.

Beyond a = 1 m the formula PX/E is sufficiently conservative.

Below a = 1 m, the maximum root B.M. dropped to approximately 0.5P and then below a = 0.5 m gave indications of tending to P/ri.

Below a = 0.5 m the disparity between the formula and the theory is evident but, because the theoretical maximum value is spread over such short lengths and because practical loads are patch loads the formula should be adequate (the patch load for 16K is 500 mm and below a = 0.5 m, it cannot completely fall over the cantilever*).

The maximum sagging B.M.s due to T44 were no more than 50% of those due to A14 (alleviating effect from other wheels).

The maximum sagging B.M. due to a patch load of 500 mm length is less than 50% of that due to the point load.

The envelope of maximum sagging moments due to point load is almost exactly predicted by a curve equal to 25% of the AASHTO formula for one wheel line.

It is therefore recommended that the existing formula is adequate but that designers be advised to cater for sagging moments which are 25% of the value of PX/E for one wheel.

Results for shear have not been studied. However, it is doubtful if shear could ever predominate over flexural considerations except for very short thin cantilevers. Presumably, if a shear is calculated, the design approach would be to treat it as a punching problem.

J.E. Wheeler

June 1974

11

(a * Except for the extreme position, with C.G. at 0.25 m.

x

-1•00

075

0•6

PED P

11

0-4 SAWKO & MILLS

L/u = 015

JARAMILLO ,tL: 03 ,03 / II ---- WESTERGAARD

/ I /I

/ II /0.2

/ / I

/ 0•1 / - / /

/ -

61 C, C

m I I I I I I I

- 20 -15 -10 -05 0 05 10 15 20 Y,V0 .

DISTRIBUTION OF MOMENT PER UNIT LENGTH (M) AT THE ROOT OF A CANTILEVER SLAB DUE TO A CONCENTRATED LOAD (P)

E 3'.

FIGURE A.2

CANTILEVER LOADING

cl

/ 70K N

70 KN

83 /77

k

A.14 LOADING

24 KN

21. KN

48KN

48KN

1.8 KN

1.8KN

48KN 1.8F(N

48KN

48KN

k

T.44 LOADING

0

120

100

80

60

z

m

F- 0

-o

/ /

/ 1/

(BOTH WHEEL LINES)

T. /4 (BOTH WHEEL LINES) a

A.16 (BOTH WHEEL LINES)

M = (OUTER WHEEL LINES )

- --A.14 (1 WHEEL LINE)

WESTERGAARD (2 WHEELS)

WE S TER GA ARD

1 POSITION OF OUTER WHEEL LINE (m) (IWHEEL)

2 3 4 1 5

-n

CANTILEVER BENDING MOMENTS DUE TO WHEEL LOADS NOTES:(1) HOGGING MOMENT -

(2) SAGGING MOMENT - 0< <a (o= 25% ofPX —ONE WHEEL LINE)

G) C xi m

(A)

NAASRA BRIDGE DESIGN SPECIFICATION 33•

APPENDIX B

NAASRA "HIGHWAY BRIDGE DESIGN SPECIFICATION" REVISION TO SECTIONS 2 and 3

ARTICLE 2.5 "OVERLOAD PROVISION"

In reviewing and re-drafting this Section the Sub Committee has proposed that the substantiation of a bridge to carry infrequent heavy loads be carried out under the application of an Abnormal Vehicle rather than the method currently prescribed by Article 2.5.

In effect, the H.B.D.S. presently requires the structure to be designed for 100% of the basic allowable stress using the MS vehicle (with as many vehicles as are necessary to produce maximum effects, up to and including a number equal to the number of design lanes). Design for local effects is by use of the individual 72 kN wheel loads in one axle. It further requires for overload check that the loading of a single MS vehicle be doubled and be checked against 150% of basic allowable stress.

The first two requirements above are unchanged in the new draft, except that the mass of the MS vehicle will be increased by 1/3 and local effects are still to be designed for single axle with 72 kN wheel loads (since it is unlikely that future legislation will permit any increase in individual wheel loading). However, the proposed overload assessment will be carried out under Load Group X wherein an Abnormal Vehicle, plus 10% impact, will replace the live loading of Group I (stream flow omitted), with allowable overstress 25%*.

Because of anticipated differences in requirements between S.R.A. 's the new draft currently allows the folloWing :-

A Standard Abnormal Vehicle - configuration and mass given (but S.R.A. may vary total mass)

or, alternatively

A Special Abnormal Vehicle - configuration and total mass to be chosen by the S.R.A.

The 1974 B.E.C. directed that the Sub Committee should investigate and justify an Abnormal Vehicle configuration and total mass. The calcula-tions reported herein have been undertaken with this aim in view. It is presumed that if such configuration and mass is determined and accepted it would form the basis of alternative (a) above.

Basically, the objective of an overload assessment is to ensure that the structure is capable of accepting occasional heavy indivisible loads, such as those required by power generating authorities, without the structure suffering permanent damage or becoming unserviceable. It is, therefore, not enough to ensure that the main flexural members should exhibit this capability

0 * Later amended by B.E.C. to 43.

34 NAASRA BRIDGE DESIGN SPECIFICATION

S but rather that each and every component should have a reserve strength which is comparable and at least capable of accepting a practical overload without exceeding the permissible overstress. In other words, it is possible to multiply the MS vehicle effect by some factor (greater than unity) so that the main flexural members achieve a permissible overstress; and it is likewise possible to apply a vehicle, having a practical heavy load configuration and weight, whose effect results in the same permissible overstress in the main flexural members. Assuming the in-span sagging bending moment as the main index of the bridge load capacity, we thereby ensure that in both cases the overload capacity of the main in-span flexural members is the same. However, in the first instance we are applying a six-wheeled vehicle with a factored load to the structure. In the second instance we are applying a mul .1-wheeled, multi-axled vehicle* to the structure. To assume that the first approach would ensure overload reserve strengths in other components of the structure that are comparable or adequate is not considered to be valid or perhaps even logical.

The investigation was carried out for several types of bridges which are typical of those currently being constructed in Western Australia, namely prestressed concrete and universal steel beam composite bridges, R.C. box beams and R.C. slab bridges, all of which were amenable to distributive analysis by the Guyon-Massonet method. The stress resultants calculated were the longitudinal hogging and sagging bending moments in beams, beam shears, pier reactions and deck transverse bending moments. The loadings were MS24 type vehicle (for design against basic allowable stresses the mass was 45 tonnes** ) and an Abnormal Vehicle of the Standard configuration.

The stress resultants calculated were

The design value for a 45 tonne (MS24) vehicle using as many (up to and including the number of design lanes) as necessary to give maximum effect.

The mass of one MS vehicle required to give 25% increase*** in stress over design value (mass = M1 tonnes)

The mass of one Standard Abnormal Vehicle required to give 25% increase in stress over design value (mass = M2 tonnes; 196 tonnes being the mass of a W.A. Abnormal VehIcle).

* It is suggested here that the Standard Abnormal Vehicle chosen, having two groups of multi-wheeled axles with variable spacing, is representative of the type used to convey large heavy indivisible components and is therefore a practical configuration. In longer spans the maximum effect is more a function of the load than of the configuration but it can be readily demonstrated that, for the shorter spans which form the bulk of road bridges, the configuration chosen produces the most severe overall effects.

** Later amended to 11.14 tonnes.

The allowance was amended to 140% in the new rules.

NAASRA BRIDGE DESIGN SPECIFICATION 35

The attached tables show the maximum effects in several important points in a bridge, namely longitudinal B.M. in-span and over supports (kNm per beam), shear (kN per beam), reactions (kN per column), transverse B.M. in deck (kNm per unit width).

Table B1 shows the design value and the unfactored values due to a single MS vehicle and the Abnormal Vehicle.

Table B2 taking each part of the structure separately, we calculate the mass of the MS vehicle and of the Abnormal Vehicle to give a 25% increase above the basic allowable stress at that part, assuming that the stress resulting from the design effect (plus D.L.) equals the basic allowable stress.

Table B3 using the in-span sagging B.M. as the main structural index for the bridge, take the MS vehicle and the Abnormal Vehicle with masses calculated for that point to give 25% overstress and apply it to

is

all other parts. Calculate the resulting overstress above design value.

The usual procedure in bridge design based on working stresses is to proportion the bridge members using the maximum effects derived from the Standard Highway Loading and the specified allowable stresses. The maximum effects due to a vehicle, intended to check the design for occasional heavy luads, are then calculated and each component is assessed for ovcrstrcs. Where the stress exceeds the overstress value permitted then the component is re-proportioned. This should ensure that, under a heavy loading, the structure has no "weak" points and the degree of overstress does not vary markedly throughout the bridge. However, such an assessment for overload capacity does depend entirely on the choice of overload check vehicle, especially the configuration.

In Figure B1 the comparisons have been made on the assumption that an overload check vehicle mass has been established using in-span sagging B.M. as the basis (as in Table B3). If that vehicle were a heavy load of practical configuration (the Abnormal Vehicle) then the component is pro-

S

portioned for the design effect and is increased in proportion if the overstress exceeds 125%. On the other hand, had the overload check been made using an MS type vehicle the component would have resulted in a deficit or an unwarranted excess in overstress capacity. The table shows that in 37% of cases the result is equality of effect, in 43% of cases the result is a deficit and in 6% of cases the result is an excess.

Table B4 has been calculated in exactly the same manner as Table B3, except that, whereas the design effects in Table B3 were calculated by Guyon-Massonnet distribution methods, the design effects in Table B4 were distri-buted using the empirical factors given in H.B.D.S. Section 3. Generally, the result shows the conservatism in the design values of the latter method.

CONCLUSIONS

The exercise shows that use of an Abnormal Vehicle of practical con-figuration as an overload check is a more logical procedure than the method currently advocated in the H.B.D.S. The use of a "factored up" MS vehicle could result in unacceptable variations in component overload capacity.

The Standard Abnormal Vehicle configuration consisting of two groups of multi-wheeled axles with variable spacing is a practical one. There are


other overload configurations which generally produce less severe effects. Under the terms of the proposed draft the S.R.A. can, if need be, opt for an alternative or additional Abnormal Vehicle.

3. The magnitude of the Abnormal Vehicle should, it is recommended, be established by various S.R.A.'s calculating the mass of such a vehicle as to produce 125% overstress under Group X Load combination on the following :-

The assessment is at the point of maximum in-span sagging bending moment

The assessment is carried out on that type of bridge most commonly built (or to be built) by the S.R.A.

J.E. Wheeler

September 1974

11

MS24 VEHICLE (45 TONNE), IMPACT AS PER H.B.D.S. LIVE LOAD PLUS IMPACT

- ABNORMAL VEHICLE (196 TONNE). 10% IMPACT iIMax. Sagging B.M. Max. Hogging B.M. Max. Shear Max. Column Max. Trans. B.M.

BRIDGE TYPE kNm/Beam kNm/Beam kN Reaction kN kN/Unit Width

I II III I II III I II III I II III I II III

Comp. U.B. - 7 Beams 485 314 430 393 254 555 265 171 247 414 318 842 10.8 10.8 25.8 Spans 37'-47'-37' (2) (2) (2) (2) (1) 3 Design Lanes

Comp. U.B. - 4 Beams 800 770 1 050 660 630 1 230 325 255 441 535 365 868 9.2 9.2 21.9 Spans 50'-60'-60'-50' (2) (2) (2) (2) (1) 2 Design Lanes

Comp. U.B. - 9 Beams 700 635 605 595 540 583 215 196 213 346 294 690 9.4 9.4 31.6 Spans 60'-80'-80'-60' (2) (2) (2) (2) (1) 2 Design Lanes

Comp. P/S. - 7 Beams 680 620 540 560 510 660 281 267 236 535 365 868 11.5 11.5 25.7 Spans 50'-60'-60'-50' (2) (2) (2) (2) (1) 2 Design Lanes

R.C. Slab 27.5 21 23.5 31.5 24 20 66 50 21 337 166 306 5.1 5.1 12.4 Spans 20'-20'-20' (2) (2) (2) (2) (1) 2 Design Lanes + Footwal k

R.C. Box Girders 1 065 375 770 1 025 374 640 253 94 183 985 428 980 86.5 104 132.1 Spans 80-100-80' (5) (5) (2) (5) (3) 6 Design Lanes

NOTES:- (a) I - MS24 Vehicles (No. to give nax. effect in brackets)

II - One MS24 Vehicle

III - One Abnonnal Vehicle (Central Movement)

(b) Effects for Bridge (D) are per unit width

MASS OF MS VEHICLE (M ) AND MASS OF ABNORMAL VEHICLE (M ) REQUIRED TO PRODUCE 25% 1INCREASE IN STRESS ABOVE DESIGN 2VALUE

(TONNES)

BRIDGE Sagging B.M. Hogging B.M. Shear Reaction Trans. B.M.

M1 M2 M1 M2 M1 M2 M1 M2 M1 M2

A 90 276 106 196 93 260 93 265 57 206

B 63 200 68 106 78 260 99 276 57 206

C 67 304 72 244 67 267 88 285 57 146

D 68 337 59 224 64 267 107 288 57 218

E 78 296 81 428 77 253 123 246 57 200

F 194 410 270 660 202 356 141 350 57 265

I, .

.. . OVERSTRESS PRODUCED* BY M.S. VEHICLE (M1) AND ABNORMAL VEHICLE (M2) REQUIRED TO

GIVE 125% OVERSTRESS ABOVE MID-SPAN SAGGING MOMENT DESIGN VALUES (PERCENTAGE)

BRIDGE & Sagging B.M. Hogging B.M. Shear Reaction Trans. B.M. LOADING

MS AV MS AV MS AV MS AV MS AV

MS = got 125 110 124 128 200 AV = 276t 125 159 125 147 142

MS = 63t 125 118 106 98 140 AV = 200t 125 150 127 140 103

MS = 67t 125 217 124 110 147 AV = 304t 125 240 136 115 154

D.MS = 68t 125 127 145 100 148 AV - 337t 125 156 150 172 172

E. MS 78t 125 122 125 90 173 AV = 296t 125 98 65 128 152

F. MS = 194t 125 112 128 145 430 AV = 410t 125 100 120 134 210

* LOAD DISTRIBUTION CALCULATED BY GUYON-MASSONNET METHOD

0

OVERSTRESS PRODUCED BY MS VEHICLE (M1) AND ABNORMAL VEHICLE (M ) REQUIRED TO GIVE 125% OVERSTRESS ABOVE MID-SPAN SAGGING MOMENT DESIGN VAUES

(PERCENTAGE)

BRIDGE Sagging B.M. Hogging B.M. Shear

MS AV MS AV MS AV LOADING

A. MS = 66t 125 119 99 AV = 202t 125 158 100

B. MS = 66t 125 120 108 AV 206t 125 153 126

C. MS = 50t 125 119 128 AV = 228t 125 129 135

D. MS = 48t 125 115 129 AV = 235t 125 140 129

E. MS = 63t 125 116 118 AV = 294t 125 110 61

F. MS 180t 125 115 121 AV = 392t 125 100 118

LOAD CALCULATED BY NAASRA DISTRIBUTION FACTORS

FIGURE B.1

41

70 60I 50I . I I

40 30 20I I

10 0 10 20I I 30I 40I 50 60 70

- % DEFICIT I % EXCESS

+ J IN-SPAN I 1MAIN BEAMJ

a 110.

I MAIN BEAMS OVER SUPPORTS

I +Dox

ii

X0O4 'MAIN BEAMS I lob SHEAR J

PIER SUPPORT SY + ox i STEM

D E C K T R!!~~_ 0 X 1111110. 0 +

NOTES: 1. ZERO 0/0 LINE REPRESENTS COMPONENT PROPORTIONED FOR:—

(a)STRESSES DUE TO MS DESIGN VEHICLE ALLOWABLE STRESS

(b)STRESSES DUE TO ABNORMAL VEHICLE 1'25 xALLOWABLE STRESS

2. BRIDGES:—A + B0 CX D0 E F4

EXCESS OR DEFICIT IN BRIDGE STRUCTURAL

COMPONENT WHICH WOULD RESULT FROM

USE OF MS VEHICLE CONFIGERATION AS OVERLOAD

CHECK COMPARED WITH OVERLOAD CHECK

BASED ON ABNORMAL VEHICLE WITH PRACTICAL

WHEEL CONFIGURATION.


APPENDIX C

BRIDGE OVERLOAD CAPACITY

TO: ALL ENGINEERS, BRIDGE SECTION

Up to the present the H.B.D.S. (1970 Edition) has required that structures should be checked for overload capacity using a single H520 truck, increased by 100% and without concurrent loading of any other lane. The combined DL + LL + I stresses under this load should not exceed 150% of the allowable.

In recent years the M.R.D. has not used this criteria to check the overload capacity of bridges. Based on road classification, an abnormal vehicle of 200 or 240 tons was applied and stresses were required to be with-in 125% of the allowable.

During the review and re-draft of Section 2 investigations showed that use of the design vehicle as an overload check was not satisfactory. The wheel and axle configuration is not a practical one (for heavy loads) and results in considerable divergence of structural component capacity throughout the bridge. Any investigation into overload capacity, especially by use of the so-called "abnormal vehicle", is intended to ensure the integrity of the structure against damage and unserviceability by the occasional passage of a heavy vehicle (large indivisible loads transported from time to time by organisations such as power generating authorities). Not only should the main flexural members be capable of sustaining these loads without damage or cracking, but all the structural components should, ideally, exhibit comparable overload capacity. By this is meant sagging and hogging flexurál members, shear in webs and cell walls, transverse bending and local effects in slabs and decks, reactions in bearings and support members, and stress resultants in cross heads, diaphragms, half caps, etc.

The new draft code proposes an increase in Standard Highway Loading of approximately 33%. It proposes the use of an abnormal vehicle check. It specifies the configuration and the mass but permits State Road Authorities to vary the mass and also to use other (special) abnormal vehicles if desired. A new load group is introduced (Group X = D + AL (abnormal live load) + I (10%) + E + PS and the overstress limit is' 140% of the allowable, except for prestress concrete (where overstress is given as a function of F'c) and elastomeric bearings (where design is based total shear strain not exceeding ,a percentage of ultimate tensile strain).

Obviously, if an abnormal vehicle is to be used as an overload check, then it must bear some relationship to the design loading. If the design vehicle increases from 33 tonne mass (MS18) to 44 tonne mass (T44) then the capacity of a structure for overload must logically increase and a new abnormal vehicle mass must be determined (remembering of course that not only the overstress limitation on the structural members should be preserved but that all structural members should have comparable overload capacity). In the investigation it was found that if the old system of increasing the HS vehicles was employed, some components would be highly overstressed and others over-conservative in design when a heavy load of practical configur-ation was applied.

The process of establishing the abnormal vehicle (AL) mass can be described as follows. Take a selection of bridges, all designed for the Standard Highway Loading (SHL) and analyse the load distribution under the


0 passage of an AL of unit total mass. The vehicle should generally be placed centrally between kerbs except in the case of pavements with a median. On the assumption that the flexural members resisting sagging B.M. represent the strength index of the structure, the maximum sagging B.M. per beam or per unit width of slab under the AL is calculated. The unit mass of the AL is then factored up until the overstress limit is reached (e.g. 140% or see Article 6.4.1 (f) for P/S concrete).

The resultant AL mass is then the limiting value for that structure. On a graph, va'ues of mass would be plotted against span length for various types of structure. The degree of spread in the points on the graph would, in some measure, reflect consistency in design. It is also possible that some materials of construction would show up unfavourably with low values of mass and no doubt, this deficiency in overload capacity is reflected in construction cost (i.e. the apparent lower cost of construction is a false indication if the overload capacity is low).

From the spectrum of results one could establish upper and lower bounds to the AL mass, as well as a mean value. It would then involve a policy decision as to which level of mass to adopt throughout, or whether more than one vehicle is desirable depending on projected usage. This latter can be noted in the present M.R.D. classification of bridges by road type and location.

With the advent of new design loadings, the position of abnormal vehicle load levels has to be re-assessed. It is therefore proposed that all new designs should be treated as follows :-

Analyse and carry out a preliminary design for the prescribed Standard Highway Loading.

Analyse the structure under the application of an abnormal vehicle of unit total mass as shown in Figure 2.3 (new draft of Section 2).

The stress resultants should be calculated for at least the following:-

3.1 Maximum sagging B.M.

3.2 Maximum hogging B.M.

3.3 Maximum shear

3.4 Maximum reaction

3.5 Maximum transverse B.M.

Calculate, in each case, the AL mass which would cause the stress resultant to reach the maximum permissible overstress for Group X.

Summarise these results and refer them, through your Senior Engineer, to Principal Engineer (Design) for assessment and a directive as to how to proceed with the design.

J.E. Wheeler SENIOR ENGINEER BRIDGE DESIGN

September 29 1975

44 NAASRA BRIDGE DESIGN SPECIFICATION

S APPENDIX D

THERMAL EFFECTS - COMMENTARY

The division of climatic conditions of Table 2.3 into "cold", "moderate" and "hot" is vague. Although different regions of Australia have varying mean temperatures, the maximum and minimum temperatures are often similar.

AS 1170 Pt.1 seems to be one step better in that specific areas have been defined. However AS 1170 does not give values for mean temperature (which we require for setting bearings, expansion joints, etc.) but orly gives ranges of temperature. In our code we also need to give bridge temperatures rather than air temperatures of AS 1170.

In an attempt to put better labels on the climatic regions, I divided Australia into a number of regions and considered the mean and extreme temperatures in each. The regions are shown in Tables D.1 and D.2. 5

I then tried to combine groups with similar temperature ranges into three or four groups (something like AS 1170). I think those given in Table D.3 are reasonable.

The next step is to relate air shade temperatures to bridge tempera-tures. Unfortunately, not much information on bridge temperatures in Australia is available and the most useful references I have found are two by Mary Emerson of the Road Research Laboratories (see References).

She concludes :-

For steel box-section bridges -

1.1 On cold days during winter, the minimum mean bridge temperature will fall 3 to 4°C below the minimum shade temperature.

1.2 On hot days during summer, the maximum mean bridge temperature will be about 1.5 times the corresponding maximum shade temperature (expressed in degrees Celcius).

For a concrete bridge -

2.1 Minimum about 6°C above minimum shade temperature.

2.2 Maximum about equal to maximum shade temperature.

The groups adopted by BS 116 seem a logical choice for types of structure,

viz. (a) Steel decks on steel box girders

(b) Steel decks on steel truss or place girders*

* Omitted from draft as being unlikely to be required.


0 Concrete decks on steel box, truss or place girders

Concrete slab, and concrete deck on concrete beams or box girders

BS 116 gives values of mean temperature for both surfaced and unsurfaced decks. Different values have been given for categories (a) and (b) (steel decks). In view of all previous assumptions I suggest one temperature for both cases.

For steel decks, factoring the maximum shade temperature of say 52°C (1250 F) by 1.5 gives a temperature of 78°C (1720 F). This seems high and I question whether the factor of 1.5 can be applied to Australian conditions. However, an increase of temperature of about 30°F (17°C) over the shade temperature seems possible and I have selected a maximum mean temperature of 70°C for steel structures.

The values for the other types of structures I have arrived at by considering Mary Emerson's conclusions for concrete structures and BS 116 values for various types of structure.

DIFFERENTIAL TEMPERATURES

I suggest adoption of the values of differential temperature given in Draft British Standard BS 116. The only figures for comparison are for concrete box girder structures and the differential temperatures assumed were not exceeded. I have suggested omitting the reverse gradient from the bottom of concrete slabs and concrete box girders - this distribution would then model closely that obtained in the Springvale Road structure. The omission of the reverse gradient gives greater stresses in continuous structures.

(Note: The values adopted in the draft were based on the British draft but amended in the light of measurements made by CRB (results to be published).

S REFERENCES

Bureau of Meteorology "Climatic Averages - Australia"

Mary Emerson "Temperature Effects in Bridges" Highways Design and Construction, July 1972.

Bryant A.H., Buckle, I.G., Lanigan, A.G. "Prediction of Temperatures in Box Girder Bridges" 7th ARRB Conference Adelaide 1974.

4

G. Bott

D.M.R. N.S.W. 1974

0

TABLE Dl

Description Latitude Distance from

Coast Height above Sea Level

Comments

1 Far North-coast North of 200S Less than 200 km All Note 1

2 North-coastal Between 20 & 250S All

(East coast)

3 North-coastal All

(West coast)

4 Central-coastal ,, , 1 000 ft

(East coast)

4A " " " " between 1 000' Note 2 & 3 000'

5 Central-coastal , All

(West coast)

6 Southern coastal South of 300S " 1 000 ft

6A " II " " " Between 1 000 & 3 000'

7 Over 3 000' " " " " 3 000 ft

8 North-Interior North of 250S 200 km All

9 South-Interior South of 250S 200 km All

10 Tasmania - - All Note 3

NOTES

Only six meteorology stations in this region were at a height above sea level of more than 1 000 ft. The mean temperatures were about the same or very slightly lower than stations below 1 000 ft., the maximum temperatures the same and the minimum about the same or slightly lower. I thus decided a further subgroup here was necessary.

In this region there were 18 stations between 1 000 and 3 000 ft. above sea level. Combining all the stations in this region below 3 000 ft. the following values were obtained.

Number of Stations - 39

Mean S.D. Maximum Minimum

Mean Temperature 65.28 3.53 70.0 58.2

Maximum Temperature 108.13 4.97 114.0 95.0

Minimum Temperature 23.95 7.42 42.0 13.6

There was only one station in Tasmania over 3 000 ft., and four between 2 000 and 3 000. All stations in Tasmania were thus included in the same group.

'6.

S

0

I . S

Group No of MEAN TEMP. MAXIMUM TEMP. MINIMUM TEMP. a ions Mean S.D. Max. Mm. Mean S.D. I Max. Mm. Mean S.D. Max. Mm.

1 36 77.61 3.54 84.4 67.6 108.16 5.02 118.0 98.2 40.69 8.53 63.5* 25.2

2 17 71.23 1.24 73.2 68.5 105.82 8.24 119.0 90.0** 31.50 8.74 46.3 19.0

3 6 78.30 2.44 81.9 76.1 118.22 2.24 121.0 115.0 33.72 4.93 40.0 27.0

4 21 67.96 1.21 70.0 65.0 108.76 5.89 114.0 95.0 28.78 6.04 42.0 21.2

4A 18 62.15 2.62 66.5 58.2 107.39 3.65 114.0 101.5 18.31 4.20 28.2 13.6

5 8 68.48 1.73 71.3 66.6 115.61 1.12 117.2 114.0 29.48 3.08 33.4 25.0

6 155 60.26 2.81 66.3 55.5 112.48 4.36 123.2 100.0 26.91 4.86 40.4 17.3

6A 52 56.00 2.79 63.4 49.3 108.45 3.51 114.9 100.5 19.43 4.55 27.8 10.0

7 19 50.72 4.74 56.8 40.4 99.04 5.97 109.0 82.0 14.78 6.01 26.5 5.0

8 28 75.07 3.04 80.9 69.1 115.76 3.70 127.5 108.6 26.87 4.49 39.0 19.0

9 159 63.00 4.40 72.4 52.5 115.01 3.88 125.0 101.0 22.32 4.07 42.0 12.0

10 23 51.96 3.40 56.0 43.0 97.37 5.49 105.2 84.6 22.28 6.48 31.0 9.0

* Thursday Island

Max. excluding this is 57.6 Cape York

** Lady Elliot Island Min excluding this is 95.0 Cape Capricorn

TABLE D3

REGION I NORTH OF 250S

Combines groups 1, 2, 3 and 8

MEAN TEMPERATURE MAXIMUM TEMPERATURE MINIMUM TEMPERATURE

Mean Max. Mm. Mean Max. Mm. Mean Max. Mm.

1 77.6 84.4 67.6 108.2 118.0 98.2 40.7 63.5 25.2

2 71.2 73.2 68.5 105.8 119.0 90.0 31.5 46.3 19.0

3 78.3 81.9 76.1 118.2 121.0 115.0 33.7 40.0 27.0

8 75.1 80.9 69.1 115.8 127.5 108.6 26.9 39.0 19.0

REGION II SOUTH OF 250S

excluding areas above 3 000' and excluding Tasmania, i.e. combines groups 4, 5, 6 and 9



4 68.0 70.0 65.0 108.8 114.0 95.0 28.8 42.0 21.2

4A 62.2 66.5 58.2 107.4 114.0 101.5 18.3 28.2 13.6

5 68.5 71.3 66.6 115.6 117.2 114.0 29.5 33.4 25.0

6 60.3 66.3 55.5 112.5 123.2 100.0 26.9 40.4 17.3

6A 56.0 63.4 49.3 108.5 114.9 100.5 19.4 27.8 10.0

9 63.0 72.4 52.5 115.0 125.0 101.0 22.3 42.0 12.0

REGION III TASMANIA

and areas south of 250S at a height of more than 3 000' i,e. combines groups 7 and 10



7

10 11

50.7

52.0

56.8

56.0

40.4

43.0

99.0

97.4

109.0

105.2

82.0

84.6

14.8

22.3

26.5

31.0

5.0

9.0

ME

fl

NAASRA BRIDGE DESIGN SPECIFICATION 49• . APPENDIX E

FRICTION FORCES

The values in Table 2.4 are generally set as an upper bound of friction coefficients and thus to be used say in checking anchorage of bearings and restraining forces likely to be imposed on substructures due to movement or attempted movement of structures. The choices of values have been strongly influenced by DOE, RRL Report LR 382 "Notes on Bridge Bearings" by W. Black and are penal for bearings or surfaces involving sliding steel plates to discourage their use.

The values to be adopted for PTFE will be within the range as shown - but depend on the bearing pressure and the use of fillers if any in the PTFE

all as specified in the revised Section 9.

. PTFE combined with metallic lead or sintered bronze sliding on stainless steel is a general description of glacier bearings using DU(B) material.

Roller bearings if not suitably protected against deterioration of bearing surfaces can have considerably increased friction coefficients above those shown.

J.K. Lloyd

D.O.C. 1974

E


APPENDIX F

OVERTURNING AND UPLIFT

It was considered that the existing clauses 2.22 Uplift and 9.10 Anchorage of the 1970 edition did not adequately warn designers or ensure satisfactory factors of safety against loss of stability of structures as a whole or finite parts of structures.

It is noted by the Sub Committee that some S.R.A.s were using design lateral loads such as wind and stream forces factored by two as design criteria and/or stability checks.

The revised clause is influenced by the stability requirements of BS 153 Part 3A:1972 Clause 18 Anchorage and S.A.A. Concrete Structures Code AS 1480-1974 Clauses 9.3 Stability of the Structure as a Whole, and 13.1.1 Ultimate Conditions and S.A.A. Steel Structures Code AS 1250-1972 Clause 3.3.2 Stability and proposed amendments to Clause 3.3.1 contained in S.A.A. document BD/1/74-2.

The proposed load factor approach recognises the probability of occurrence and also more importantly the dangers in under-assessing dverturning transient loads such as wind or stream forces where an increase in velocity of say 20% can result in an increase of force of 40%.

J.K. Lloyd

D.O.C. 1974

0


IRE FERENCES

INDIAN ROADS CONGRESS (1965). Bridge Loadings Round the World. Comm. Monthly Review.

FRY, A.T., EASTON, G.R., KER, I.R., STEVENSON, J., WEBBER, J.R., (October 1975). A study of the Economics of Road Vehicle Limits. NAASRA Study Team Report R.3

SAWKO, MILLS (1971). Design of Cantilever Slabs for Spine Beam Bridges. mt. Conf. on Developments in Bridge Design and Construction, Cardiff.

MICHAEL, K.C. (1973): Bridge Design Specifications. International Training Course in Road Engineering for African and Asian Engineers, Perth.

. 5. FENVES, S.J., VELETSOS, N.S., SIESS, C.P. (1962). Dynamic Studies of AASHTO Road Test Bridges. Highway Research Board Special Report 73.

CSAGOLY, P.F., CAMPBELL, T.I., AGARWAL, A.C. (1972). Bridge Vibration Study. Ministry of Transport and Communication Report RR 181, Ontario.

PAGE, J. (1976). Dynamic Wheel Load Measurements on Motorway Bridges. T.ft.R.L LabOatory RpOrt 722.

Inquiry into the Basis of Design and Erection of Steel Box Girder Bridges. Interim Design Rules, Part 1, Article 4.21.

GADE, R.H., BOSCH, H.R., PODOLNY, W. (July 1976). Recent Aerodynamic Studies of Long Span Bridges. A.S.C.E., J. Struct. Div. St.7.

SCANLAN, R.H. (1976). Modern Approaches to Solutions of the Wind Problems of Long Span Bridges. A.I.S.C. Engineering Journal, Second Quarter.

. 11. LEONARD, D.R. (1966). Human Tolerances for Bridge Vibrations. Road Research Laboratory Report No. 34, Harmondsworth.

WRIGHT, R.H., WALKER, W.H. (Jan.. 1972). Vibration and Deflection of Steel Bridges. A.I.S.C. Engineering Journal.

JAVOR, T. (1973). The Dynamic Effects on Prestressed Concrete Bridges Built Without Falsework. I.A.B.S.E. Symposium, Lisbon.

CHANG, F.K. (Jan. 1973). Human Response to Motions in Tall Buildings. J. Struct. Div. A.S.C.E.

NISHIWAKI, T., HOSHIYA, M. (1975). Quantitative Analysis of Response Against Vibration in Bridge Design. I.A.B.S.E. Symposium, Dresden.

WHEELER, J. (September 1976). Inter-departmental communication. M.R.D. to D.M.R.

Composite Construction in Structural Steel and Concrete. B.S.I. CP 117 Part 2.


U.S. DEPARTMENT OF TRANSPORTATION (1969). Strength and Serviceability Criteria : Reinforced Concrete Bridge Members : Ultimate Design. pp. 22 and 41.

ROWE, R.E. (1962). Concrete Bridge Design. (Wiley).

WARBURTON (1964). The Dynamic Behaviour of Structures. (Pergammon Press).

FORMAN, D. (1973). Bridge Rails and Crash Barriers. M.R.D. W.A. Technical Report No. 1.

MINISTRY OF TRANSPORT, DEPT. OF ENVIRONMENT (August 1973). Standard Highway Loadings. Technical Memorandum (Bridges) BE 5/73.

NEW ZEALAND MINISTRY OF WORKS AND DEVELOPMENT. (Feb.1974). Differential Temperature Analysis. Chief Designing Engineer (Civil) Report.

NEW ZEALAND MINISTRY OF WORKS AND DEVELOPMENT. (July 1973). Highway Design Brief. Issue C.

BLACK, W. (1971). Notes on Bridge Bearings. Road Research Laboratory Report LR 382.

TAYLOR, M.E. (1972). PTFE in Highway Bridge Bearings. Transport and Road Research Laboratory Report LR 441.

NATIONAL COMMITTEE ON EARTHQUAKE ENGINEERING STANDARD ASSOCIATION OF AUSTRALIA. (1974). Seminars on Earthquake Engineering. Adelaide, Perth.

STANDARDS ASSOCIATION OF AUSTRALIA. (1976). Draft for Australian Standard Rules for the Design of Earthquake-Resistant Buildings. DR 76100.

A.A.S.H.T.O. (1975). Interim Specification, Bridges. Interim No. 2.

Earthquake Engineering. (1972). Proc. Fourth F?ropean Symp. on Earthquake Engineering, London.

BRIDGE AND STRUCTURAL COMMITTEE, JAPAN SOC. OF C.E. (1973). Earthquake Engineering for Bridge, Earthquake Resistant Design for Civ. Eng. Structures, Earth Structures and Foundations in Japan.

CHAPMAN, H.E., HUIZING, J.B.S., KENNAIRD, A.R., STANFORD, P.R. (June 15, 1974). Changes in New Zealand Highway Bridge Design. New Zealand Engineering.

GATES, J.H. (Dec. 1976). California's Seismic Design Criteria for Bridges. J. Struct. Div., A.S.C.E.., S112.

SANDERS, W.W., ELLEBY, H.A. (1970). Distribution of Wheel Loads in Highway Bridges. National Co-operative Highway Research Progress Report 83.

CUSENS, A.R. (Dec. 1974). Load Distribution in Concrete Bridge Decks. CIRIA Report 53.


BUCKLE, I.G., JONES, R.A. (1974). Load Distribution in Multi-Beam Decks. 7th ARRB Conference., Adelaide.

JONES, A.R. (1976). A Simple Algorithm for Computing Load Distribution in Multi-Beam Bridge Decks. 8th ARRB Conference., Perth.

CLARK, L.A. (Oct. 1974). Strength and Serviceability Consideration in the Arrangement of the Reinforcement in Concrete Skew Slab Bridges. CIRL4 Report 51.

.4

paper two: design loads and distribution of loads

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