organised bynituk.ac.in/cishrimg/e-proceedings_cishr-2017-min.pdffactor, soil type are considered as...

Proceedings of

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ORGANISING TEAM

Chief Patron

Prof. Shyam Lal Soni

Director, NIT Uttarakhand

Patron

Col. Sukhpal Singh

Registrar, NIT Uttarakhand

Organizing chairman

Dr. Aditya Kumar Anupam

Secretary

Dr. Kranti Jain

Members

Mrs. Smita Kaloni

MR. Shashi Narayan

Mr. Devesh Punera

Mr. Laiju A.R.

Mr. Amardeep

Mr. Bibhash Kumar

Mr. Shashank Batra

Mr. Muskan Mayank

Mr. Abhinav Kumar

Mr. Neeraj Kumar

ADVISORY COMMITTEE

Prof. A.K. Dey, NIT Silchar

Prof. Bhupinder Singh, IIT Roorkee

Prof.M.K. Srimali, MNIT Jaipur

Prof. M.N. Viladkar, IIT Roorkee

Prof. Mahesh Pal, NIT Kurukshetra

Prof.Manish Shrikhande, IIT Roorkee

Prof. Praveen Kumar, IIT Roorkee

Prof.Subhasish Dey, IIT Kharagpur

Prof. Surinder Deswal, NIT kurukshetra

Prof.Vinod Tare, IIT Kanpur

Prof. Z Ahmad, IIT Roorkee

Dr. Ankit Gupta, IIT BHU

Dr. Dharamveer Singh, IIT Bombay

Dr. Gargi Singh, IIT Roorkee

Dr. Jagdish Prashad Sahoo, IIT Roorkee

Dr. Priti Maheshwari, IIT Roorkee

Dr. Rajib Sarkar, IIT Dhanbad

Dr. S.D. Bharti, MNIT Jaipur

Dr.S.K. Mishra, IIT Kanpur

Dr.S.T.Ramesh,NIT Tiruchirappalli

Dr.Ajay Chourasia, CSIR-CBRI

Dr.Ashutosh Kainthola, KainGeotech

Er. Vinod Kumar Singh, L&T

Shri Brij Mohan Agarwal, MES

Shri R Chalisaganokar, Irrigation Deptt.

CONTENT

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1

Conference on Infrastructure Sustainability in Hilly Regions (CISHR), 21st -23rd Dec. 2017

Proceedings of CISHR-2017 Page 1

Behaviour of Buckling Restrained Braced Frame

Under Seismic Loads Lingeshwaran Nagarathinam 1, T Venkat Das 2 and Bhargava Laxmi Goli3

1 Assistant Professor, Department of Civil Engineering, K L E F (Deemed to be University), [email protected]

2 Assistant Professor, Department of Civil Engineering, K L E F (Deemed to be University), [email protected]

3 M.tech Student, Department of Civil Engineering, K L E F (Deemed to be University), Guntur, India

ABSTRACT

Buckling Restrained Braced (BRB) frame systems are currently used as the primary lateral force

resisting elements both in new construction and seismic retrofit projects. In the conventional bracing

system, braces are buckled due to the earthquake forces on the structure. To alleviate the issues in the

traditional bracing system, BRB technology was introduced. In recent days, BRB becomes the most

promising technology used in the lateral force resisting system of structures located in high seismic

regions. In this study, design, analysis and comparision of the different brace layout has been carried out.

The selection of BRB configuration has been adopted based on the suitable sway moments. Different

configurations such as Forward inclined, Zig-Zag, X-pattern are considered for the study of building in

order to provide lateral stiffness. The RC structural plan was taken in the seismic zone V with response

reduction factor of 5 and soil type is hard. The building model was considered to analyze the behavior of a

structure with and without BRB to compare the parameters of storey drift, storey forces, storey

displacement, storey stiffness and storey acceleration using response spectrum and time history method of

analysis.

Key Words: Buckling Restrained Brace, Storey drift, Storey forces and Storey Stiffness

1. INTRODUCTION

An earthquake is an effect due to the sudden release of stored energy on the earth’s surface in the

form of seismic waves. Earthquake mainly occurs due to rupture of geological faults, volcanic activities,

landslides and mine blast. The most common loads resulting from the effect of gravity are dead load, live

load and snow load. Besides these vertical loads, buildings are also subjected to lateral loads caused by

wind, earthquake. Lateral loads can develop high stresses, produce sway movement or cause vibration.

Therefore it is very important for the structure to have sufficient strength against vertical loads together

with adequate stiffness to resist lateral forces. By using Shear wall, Dampers, Bracing System we can

reduce the lateral deformation and increase the stiffness of buildings caused by earthquakes.

Braces were normally used for structures where the lateral loads are governing the design of the

structure, regardless of whether the wind or seismic loads. The bracing system is one of the most

prominently used techniques to control the displacements in the structure due to the lateral loads. The

buckling restrained braces were applied to a steel framed structure and response of the structure was

studied for different types of braces configuration. The proper designs of brb systems give good control

over both inter storey drift and total displacement [1]. A large amount of kinetic energy is incorporated

into a structure during major earthquakes. If the braces are too slender they cannot withstand the

compressive forces and if the braces are thicker then the forces on columns and beams are high which

makes the structural elements increase in size. To improve upon this situation, “damage-control

structures” are developed to decrease the comprehensive building damage [2]. The performance-based

plastic design methodology developed for the brbf design, where the design base shear was obtained based

on energy work balance using preselected target drift and yield displacement [3]. The numerical model development of buckling restrained braces used has diagonal members designed for dissipative behavior

and lateral load resistance under seismic action. The non-linear static method of analysis was implemented

to study the best brb system and the location. Results show that double diagonal bracing system

experiences less stress and deformation under applied seismic

mailto:[email protected]


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loads [4]. The stiffness of the brace increases, the maximum displacement, storey drift, and storey shear

decreases which results in better performance of the structure [5]. When the steel brace is attached to the rc

structure, it offers good resistance during the earthquake and reduces the lateral displacements of the structure

[6]. Several cross sections of brbs under different fire scenarios are considered, it is observed that higher

stiffness of brbs can resist higher failure temperature and is strong enough to suffer the whole heating and

cooling phases of fire [7]. The design based earthquake and the maximum considered earthquake hazard levels

based on statistical analysis of past seismicity data. The configurations are taken in the combination of

moment resisting and non-moment resisting. Response reduction factor is considered for the design with rigid

beam column connections and pinned beam column connections. The interstory drift and residual drift ratio

responses of brbf are calculated under dbe and mce level of earthquakes [8]. A study for the brb are designed

using fema 450 and asce-7. The equivalent lateral force procedure and nonlinear time history analysis were

used to design the brb and design curves are obtained from the considered structure [9]. As the building

number is growing due to the over population their is a necessity in the seismic areas to use buckling

restrained braces. The test set-up is made and testing programme is runned with bracing system. Out of these

the unbounded braces performs well and have good hysteresis behaviour. The analytical assessments are also

done by using the fema guidelines [10]. The structure undergoes lateral forces caused due to seismic activity.

The columns and beams of the structures are used to transfers the major portion of the gravity loads and some

portion of lateral loads but that is not significant to the stability of structure. So we provide bracing systems,

shear walls, dampers etc to resist or transfer these lateral forces to the structure uniformly without affecting

the stability and strength of the structure. In a hysteresis-damping system, members absorb seismic energy,

such as the unbounded brace is incorporated in a structure. By means of this system, it is feasible to keep

columns and beams within their elastic range, thus justifying the damage. A hysteresis-damping system allows

the sustained use of damage to the building even after an earthquake.

1.1 Concept of BRB

A BRB consists of a steel core surrounded by an outer casing that restrains local buckling but allows the

core to deform inelastically in tension and compression under strong earthquake loading. The most important

characteristic of a BRB is to yield both in compression and tension without buckling. Gravity loads are the

essential loads on the building. In any case, as the building gets taller, it must have sufficient strength and

stiffness to oppose lateral loads imposed by wind and earthquakes. The height of the building increases

additional stiffness was necessary to control the deflection, rather than the strength of the members, as

deflection dictates the design. Buckling restrained braces are chosen has lateral forces resisting system for the

building because of their large ductility, energy dissipation capability, and also for the ease of repair after a

major earthquake. A technology introduced in late 1990, the BRBF represent the state of art in moment braced

frame design. The major components of buckling restrained brace are steel core, bond preventing layer and

casing as shown in Figure.1.

Figure.1 Schematic view of Buckling Restrained Brace

1 STEEL CORE IS DESIGNED TO RESIST THE AXIAL FORCES DEVELOPED IN THE

BRACING.

Bond preventing layer decouples the casing and core. This allows steel core to resist full axial forces

which develop in bracing.

3



Casing provides lateral support against flexural buckling of the core.

1.2 Advantages

1. It is easy to adopt in seismic retrofitting because it can incorporate into the structural system by means

of a bolted or pinned connection to gusset plates.

2. Since BRB’s are light weight it does not usually require foundation strengthening. 3. Stiffness can be controlled which ultimately leads to good performance to the building.

4. Higher ductile and energy dissipative behavior under axial forces. 5. Post-earthquake investigation and replacement are relatively easy since the damage is concentrated

over a relatively small area.

2. OBJECTIVE

The building under lateral loads coming from the earthquake forces with the application of buckling restrained bracings. The following objectives are proposed for the present study.

1. Different configuration of buckling restrained braces to resist the lateral loads. 2. Analyze and interpret the storey drift, storey displacements, storey forces, storey stiffness and

diaphragm drift using BRB.

3. EXPERIMENTAL ANALYSIS The analysis of G+5 and terrace is carried out using ETABS software for special moment resisting frame

situated in zone V. The RC structure is analysed without bracing and with bracing with different

configurations. The below data consists of the plan area, beam size, column size, slab thickness, the height of

the building. Seismic parameters such as Seismic Zone, Zone factor, Importance factor, Response Reduction

factor, Soil type are considered as criteria for earthquake resistant design of structures as per IS 1893-2002.

The properties of the building and its components are mentioned in Table 1.

Table 1 Details of plan

Plan Area 34.2*19m

Beam size 230*450mm

Column size 230*600mm

Slab thickness 130mm

Utility of building Residential building

Shape of building Unsymmetrical

Height of building 18m

Type of construction RCC framed structure

Grades M30, Fe500

Seismic Zone v

Zone Factor 0.36

Importance factor 1

Response reduction factor 5

Soil Type Hard

The Figure.2 shows G+5 and terrace of RC structural plan, which was used to investigate the seismic

response of the building with BRB. The plan which is unsymmetrical in nature was used for observing the

varying storey drift and displacement. The number of bays in X and Y direction is different in the building, it

has eight bays in X-direction and five bays in Y-direction. All column sizes and beam sizes are assumed are

same through all stories of the building. The building is designed for earthquake loads of the structure by

4



using Indian standard code. Different load cases are taken such as dead load, live load and earthquake load are

applied to the building.

Figure.2 Typical plan of building

2 THE DIFFERENT CONFIGURATIONS OF BRB AS LISTED BELOW AND SHOWN IN

FIGURE. 3 WERE USED FOR ANALYSIS.

1. Bare Frame

2. Forward-inclined

3. Zig-Zag

4. X-pattern

Figure.3 Different configurations

4. RESULTS

The BRBs are modelled for the building with different configurations and comparison was made to

propose the suitable configuration. Here in order to look at the benefit of BRB system in the lateral load

conditions the comparison has been made and finalized and which gives the better performances among all

types of BRB.

4.1 Story Drift

From the Figure.4 it can be observed that building without BRB shows more storey drift compared to

the building with different types of BRB. From the above three different types of BRB, X-pattern showed to

have less storey drift. It can also be observed that the storey drift at third floor is maximum because of the

more variation in the displacement.

5



Figure.4 Storey drift

4.2 Storey Displacement

In the Figure.5 it was observed that forward inclined and Zig-Zag are having approximately same

amount of displacement. When BRB is used as lateral support to the building the displacement is reduced

compared to the normal building.

Figure.5 Storey displacement

4.3 Story Stiffness

Figure.6 shows the plot of storey stiffness in x-direction along storey height for different types of BRB.

For the different types of BRB there is a partial increase from the storey two and gradually increases to the

last storey. Bare frame looks like the stiffness is same at all storeys. The storey stiffness in y-direction gives

the best result and out of all types of BRB, Zig-Zag and X-pattern gives more stiffness.

6



Figure.62 Storey stiffness in X&Y-direction

4.4 Storey Acceleration

Figure.7 represents the storey accelerations in X- direction for different configurations of bracing.

From the figure it can be observed that type-4 shows more acceleration. There was less acceleration

difference between different types of bracings in Y-direction.

Figure.7 Storey acceleration in X&Y-direction

2.1 CONCLUSION

The selected frame model was analysed with different types of bracings such as Forward inclined, Zig-

Zag, X-pattern. From this analytical study it was observed that among various configurations, X-pattern offers

better resistance to the applied lateral loads especially seismic governing. The other configurations also

exhibit better performance for these loads when compared to the bare frame. With the help of bracings we can

reduce lateral displacements of structures. Axial forces in the columns increases when we use bracings. The

selection of bracing configuration, however, depends upon the seismic zone, functional utility and the cost

estimated.

2.2 REFERENCES

[1] W. N. Deulkar, C. D. Modhera, and H. S. Patil, “Buckling Restrained Braces for Vibration Control of Building Structure”, International Journal of Recent Research and Applied Studies, Vol.4, No.4, pp.363-372, 2010.



[2] Eric Ko, Arup, and Caroline Field, Arup, “The unbounded brace from research to California practice”, San Francisco.

[3] Dipti R. Sahoo, and Shih-Ho Chao, “Performance-Based Plastic Design Method for Buckling Restrained Braced Frames”, Engineering Structures, Vol.32, No.9, pp.2950–2958, 2010.

[4] K. P. Shadiya, and R. Anjusha, “Bracing Configuration Effect on Buckling Restrained Braced Frames”, International Journal of Innovative Research in Science, Engineering and Technology, Vol.4, No.4, pp.2552-2560, 2015.

[5] Jinkoo Kim, and Hyunhoon Choi, “Behavior and Design of Structures with Buckling-Restrained Braces”, Engineering Structures, Vol.26, No.6, pp.693–706, 2004.

[6] K. G. Viswanath, K. B. Prakash, and Anant Desai, “Seismic Analysis of Steel Braced Reinforced Concrete Frames”, International Journal of Civil and Structural Engineering, Vol.1, pp.114-122, 2010.

[7] Elnaz Talebi, Mahmood Md Tahir, Farshad Zatmatkesh, and Ahmad B. H. Kueh, “Comparative Study on the Behaviour of Buckling Restrained Braced Frames at Fire”, Journal of Constructional Steel Research, Vol.102, pp.1-12, 2014.

[8] Ahmad Fayed Ghowsi and Dipti Ranjan Sahoo, “Seismic Performance of Buckling-Restrained Braced Frames with Varying Beam-column Connections” International journal of steel structures, Vol.13, No.4, pp.609-621, 2013.

[9] Richard J. Balling, Lukas J. Balling and Paul W. Richards, “Design of Buckling-Restrained Braced Frames Using Nonlinear Time History Analysis and Optimization” Journal of Structural Engineering, Vol.135, No.5, pp.461-468, 2009.

[10] Stephen Mahin, Patrix Uriz, Ian Akin, Caroline Field and Eric Ko, “Seismic Performance of Buckling Restrained Braced Frame systems” 13th World Conference on Earthquake Engineering, 2004.

[11] Indian standard code of practice for Earthquake Resistance Design of Structures, Bureau of Indian standards, New Delhi, IS 1893-2000.

[12] Indian standard code of practice for wind loads, Bureau of Indian standards, New Delhi, IS 875(3)- 1897.

[13] Blake M. Andrews, Larry A. Fahnestock and Junho Song, “Ductility Capacity Models for Buckling- Restrained Braces”, Journal of Constructional Steel Research, Vol.65, pp.1712-1720, 2009.

[14] Young K. Ju, Myeong-Han Kim, Jinkoo Kim and Sang-Dae Kim, “Component tests of Buckling- Restrained Braces with Unconstrained Length”, Engineering Structures, Vol.31, pp.507-516, 2009.

[15] R. Sabelli, S. Mahin and C. Chang, “Seismic Demands on Steel Braced Frame Buildings with Buckling- Restrained Braces”, Engineering Structures, Vol.25, pp.655-666, 2003.

[16] Shawn Kiggins and Chia-Ming Uang, “Reducing Residual Drift of Buckling-Restrained Braced Frames as a Dual System” Engineering Structures, Vol.28, pp.1525-1532, 2006.

[17] H. Y. Chang and C. K. Chiu, “Performance Assessment of Buckling Restrained Braces” Science Direct, Vol.14, pp.2187-2195, 2011.

[18] L. Di Sarno and G. Manfredi, “Seismic Retrofitting of Existing RC Frames With Buckling Restrained Braces” ATC & SCI Conference on Improving the Seismic Performance of Exixting Building and Other Structures, 2009.

[19] Seyed Taghi Rasouli Amreie, Leila Kalani Sarokolayi and Alireza Mohseni Saravi, “the effect of different connections of steel structures on the seismic behaviour of buckling-restrained braced(BRB)”, International Journal of Scientific Research Engineering and Technology, Vol.4, No.8, pp.841-845, 2015.

[20] Lingeshwaran Nagarathinam, “Analysis and design of G+5 Residential buiding by using ETABS”, International Journal of Civil Engineering and Technology (IJCIET), Volume 8, Issue 4, April 2017.

8



Effects of Hilly Region Topography on Rayleigh Wave

Neeraj Kumar 1 and J.P. Narayan

2

1,2 Department of Earthquake Engineering, Indian Institute of Technology Roorkee, Roorkee,

Uttrakhand-247667, [email protected] , [email protected]

ABSTRACT

In hilly regions, the surface topography is made-up of the string of ridges and valleys. In last few

decades, Government of India has made a huge investment in the development of infrastructures in

Himalayan region by construction roads, bridges, Dams etc. Spatial variability in ground motion affects the

long-span structures like bridges and Dams in the hilly areas. In this paper, we have simulated the Rayleigh

wave responses of topography models with a string of ridges and valleys using forth order accurate staggered

grid viscoelastic P-SV wave algorithm, developed by Narayan and Kumar (2014). We have observed the

insulation effects of hill topography on the Rayleigh wave characteristics. The analysis of simulated results

revealed that there is an amplification of the horizontal component of the Rayleigh wave whereas de-

amplification of the vertical component of the Rayleigh wave at the top of the triangular ridge. Furthermore,

the energy of Rayleigh wave has reduced to less than 10% in vertical component after passing through a

string of ridges and valleys having the total horizontal distance of 4.5 Km. It is concluded that the hill

topography act as a natural insulator for the Rayleigh wave for those frequencies whose wavelength is less

than the width of ridge/valley.

Key Words: Surface waves, Hill topography, Finite Difference method, Numerical simulation.

1. INTRODUCTION

A lot of research has been done to quantify the effects of surface topography on body waves (Geli et al.,

1988; Pedersen et al., 1994; Spudich et al., 1996; Kamalian et al., 2006; Zhao, 2010; Gao et al., 2012;

Narayan and Kumar, 2015). But still, limited studies have been done on the effects of surface topography on

surface wave characteristics. The surface waves (Rayleigh and love wave) are more devastating for civil

structures than body waves, especially to the long-span structures like bridges, dam, pipe lines etc in the hilly

region. The characteristics of ground motion at a particular site depend mainly upon three factors - source,

the path of propagation and local topographical features. In the past, local topography played a crucial role in

determining the extent of damage during an earthquake. Sanchez-Sesma et al. (1988) found that the

amplification of acceleration is no more than 2 at the crest, peaking when the wavelength is about equal to

the ridge width and also that neighbouring ridges may have a greater effect on site response than layering.

Narayan and Rao (2003) have also simulated the responses at different elevations on both the weathered and

non-weathered ridges. Aki (1988) proposed that for a triangular wedge, the amplitude of vertically

propagating SH-wave is amplified whereas, in triangular wedge type valley, the amplitude is de-amplified at

the base of trough due to defocusing effects.

The effect of the valley was observed in the Mandal valley during Chamoli earthquake of March 29,

1999 (Narayan and Rai, 2001). The damage was much less in the Mandal proper village and Khalla village

compared than other villages since these villages were situated at the base of the valley. The effects of single

triangular ridge and valley were studied by Savage (2004), mentioning a large amplification of the horizontal

component and de-amplification of the vertical component of Rayleigh wave at top of ridge. He also

reported the de-amplification of both the components of Rayleigh wave at the base of a valley. The increase

of infrastructure and population in the hilly region in recent few decades calls for an urgent need to the

quantification of effects of hilly region topography with different numbers of ridges and valleys on the

Rayleigh wave characteristics for seismic hazard and risk evaluations and cost-effective sustainable

earthquake engineering.



9



2. MODEL PARAMETERS AND SOURCE IMPLEMENTATION

The effects of hilly region topography consisting of a number of ridges and valleys lying between the

epicenter and the site of interest are studied considering four topography models. The dimension of each

ridge and valley is taken as width 720 m and height/depth 360 m. The position of the source, reference point

and the receiver array are fixed. Each model is headed by a ridge, whose position is fixed and the successive

valley or ridge are added thereafter. The reference point is taken as the axis of the first ridge (R5) for the

measurement of the horizontal distances. Similarly, free surface is considered as a reference level for

measurement of vertical distances. Height and depth are denoted by (+ve) and (-ve) sign, respectively. The

four topography models, T1R, T1RV, T1RVR and T3RV represents single ridge, ridge and valley, ridge-

valley-ridge and string of three ridge-valley, respectively, as shown in figure 1.

Figure 1: Sketche for a hill topography of T3RV model having three string of ridge -valley. R is representing the

locations of receiver points (Note: horizontal distances are measured wrt to the receiver R5 and vertical distances are measured wrt to the mean elevation across the topography).

Table 1 Rheological parameters for the visco-elastic rock.

Velocity at FR Quality factor at FR Density

(Kg/m3)

Unrelaxed moduli (GPa)

VP (m/s) VS (m/s) QP QS µu Ku λu

1600 920 160 92 2200 22.22 63.45 19.00

The GMB-EK rheological model for the viscoelastic homogeneous rock like P-wave and S-wave

velocities and quality factors measured at the reference frequency (1.0 Hz) in the field, density and the

computed unrelaxed moduli are given in table 1. The body waves were generated at a distance of 900 m

towards the left of the reference point and at a depth of 102 m. The body waves in the P-SV wave FD grid is

generated by applying shear stress σXZ in the form of Gabor wavelet. The amplitude of generated P-wave is

negligible as compared to that of the SV-wave since only shear stress is applied at the focus. The

mathematical formulation for the Gabor wavelet is given below

𝑆(𝑡) = 𝐸𝑥𝑝(−𝛼)𝑐𝑜𝑠[𝜔𝑃(𝑡 − 𝑡𝑆) + 𝜑] (1)

where 𝛼 = [𝜔𝑃(𝑡−𝑡𝑆)

𝛾]

2

, 𝜔𝑃 is predominant frequency, controls the oscillatory character, tS controls

the duration and is phase shift. The particle velocity and its spectra at the focus for 𝑓𝑃= 4 Hz, =1.5,

tS=0.25 s and =0. The frequency bandwidth in the Gabor wavelet is 0-15 Hz. A uniform grid size 3m is

used in both directions. The time step is taken as 0.001 sec to avoid the instability.

3. ANALYSIS OF RESPONSES OF HILLY TOPOGRAPHY

Rayleigh wave is generated by considering a very shallow focal depth so that the entire SV-wave

energy propagating towards the free surface is converted into the Rayleigh wave (Narayan and Kumar,

2010). The Rayleigh waves generated in the epicentral zone have propagated towards the right and interact

with the hill region topography in the model (Fig. 1). The seismic responses have been recorded at

equidistant 32 receiver points (180 m apart horizontally) extending -720 m left to 4860 m right of the

reference point. The horizontal and vertical components of the Rayleigh wave responses of the homogeneous

model (without hill topography) is also computed to quantify the topography effects on the Rayleigh waves,

as shown in figure 2. Figure 2 shows that the amplitude of P-wave is negligible as compared to the Rayleigh

Distance in meter

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wave and also generated Rayleigh wave is vertically polarized. The spectra of horizontal and vertical

components at a distance of 4860 m (R32) are computed. There are considerable spectral amplitudes in the

generated Rayleigh wave in the frequency bandwidth 0.5-10.0 Hz. The dominant frequency in Rayleigh

wave is around 4.0 Hz at the reference point. The spectral amplification or de-amplification of a particular

component along the flanks of the topography has been computed just by taking the ratio of spectra of the

response along the flank with the spectra of the response of homogeneous model at reference level. Further,

in the case of 2D simulation of Rayleigh wave propagation, there is no divergence effect and only damping

is responsible for the decrease of amplitude with the epicentral distance. It is assumed that damping effects is

to some extent same for both the homogeneous and topography models for a particular epicentral distance.

Figure 2: The horizontal and vertical components of Rayleigh wave responses of the homogeneous model.

3.1. Raleigh wave response

Figure 3a shows the horizontal (left panel) and vertical (right panel) components of responses of the

single ridge (T1R) model, respectively. Figure 3a depicts reflected Rayleigh waves from the left base and top

of the ridge as well as the diffracted Rayleigh waves in the form of P- and SV-waves from the left base of the

ridge. It appears that the splitting of Rayleigh wave has occurred just near the ridge-top. The seismic phases

recorded just after the ridge topography are P-wave, diffracted P-wave, diffracted SV-wave and the two

phases of the Rayleigh waves. The horizontal and vertical components of Rayleigh wave recorded at the top

of the ridge (shown by red colour) depicts that there is a very large amplification of the horizontal

component and de-amplification of the vertical component. The sudden increase of the amplitude of the

horizontal component of Rayleigh wave at/near the top of the ridge calls for the special attention in risk

analysis since the horizontally polarized Rayleigh wave may trigger the landslides under favourable

condition. Figure 3b depicts the response of T1RV model. This figure depicts similar effects due to the ridge

topography but later as the wave reaches the valley topography, the reflected Rayleigh waves from the base

of the valley and the diffracted P- and SV-waves from the base of the valley are also observed. The seismic

phases recorded just after the valley are P-wave, diffracted P-wave, diffracted SV-wave and the split

Rayleigh waves. There is de-amplification of both the components of Rayleigh wave at the base of the

valley. Further, a comparison of the amplitude of Rayleigh waves recorded at the last receiver depicts that

insulating effect of combine ridge and valley is more than that of single ridge topography. It is observed

through figure 3c and 3d that the amplitude of Rayleigh wave, both horizontal and vertical components,

diminishes as more number of ridges and valleys are added in hill model. The Rayleigh wave gets split into

more number of Rayleigh waves, diffracted P-waves, diffracted SV-waves and reflected Rayleigh waves as

the more number of ridges and valleys increases in hill topography model.

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Figure 3: Horizontal and Vertical components of responses of the (a) T1R, (b) T1RV, (c) T1RVR and (d) T3RV topography models, respectively.

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3.2. Spectral Ratio

Figure 4 shows the comparison of the spectral ratio for the horizontal and vertical components at the

last station (4.86 km towards the right of reference point) for the topography models T1R, T1RV, T1RVR

and T3RV, respectively. The spectral ratio at the last station in all the models is more or less same in both

the components of Rayleigh wave. This figure very clearly depicts the decrease of the spectral ratio for all

the frequencies with an increase of number of ridge and valley topography in the path of Rayleigh wave. It is

inferred that in case of a single ridge, the spectral ratio shows ups and downs in the frequency band of 0.5-

8.0 Hz. Furthermore, the spectral ratio is larger in the lower frequencies (less than 1.5 Hz) as compared to

the other frequencies in all the models (wavelengths are larger than the width of base or top of ridge &

valley). Furthermore, reflection from the first ridge or valley is also larger for these frequencies. In case of

T3RV model, the spectral ratio has reduced to less than 4% for frequencies larger than 1.5 Hz over a

topography span of only 4.5 km.

Figure 4: The spectral ratio for the horizontal and vertical components at the last station for the topography models T1R, T1RV, T1RVR and T3RV, respectively.

4. INSULATING EFFECT OF CONSIDERED TOPOGRAPHY MODELS

The spectral ratio is defined as the ratio of spectra of the respective component of responses of Rayleigh

wave with and without topography in the model. This spectral ratio is an indicator of amplification and de-

amplification of the particular component of Rayleigh along the hill topography as well as an indicator of the

insulating effect of topography, if the spectral ratio is computed after crossing the topography. Figure 5

illustrates the comparison of spectral ratios for the horizontal (left) and vertical (right) components of the

T3RV topography model at receiver R5 (reference point), R9, R13, R17, R21 and R25. On an average, the

spectral ratio of horizontal component is more than the spectral ratio of the vertical component for all

frequency range at the top of the ridge in T3RV model. In case of first ridge (R5 reference point), this ratio

has gone even more than 3.0. On the other hand in case of the vertical component of the first ridge of the

T3RV models, the spectral ratio is little less than 1.0 for all frequencies range. Similarly, the spectral ratio

for the horizontal (left) and vertical (right) component at the base of the first valley of T3RV model (R9

position) are shown in figure 5b. A considerable de-amplification of both the vertical component and

horizontal at the base of the valley can be inferred for all the frequencies. An overall de-amplification of both

the components of Rayleigh wave is observed at further receivers R13, R17, R21 and R25 of the T3RV

model, except minor amplification in the horizontal component at the top of ridges (R13 and R21) for all

frequency range.

13



In order to infer the relative

insulating effects of the considered hill

topography models, the ASR (Average

Spectral Ratio) was computed at last

recording point (4.86 km), represented in

brackets in figure 4. Also a comparison of

decreasing trend of ASR with addition of

each hill topography feature in T3RV

model is shown in figure 5, at the top of

ridge and base of valley respectively. On an

average, the ASR is larger in the horizontal

components as compared to the vertical

components topography models and this

difference is decreasing with an increase of

number of ridge and valley in the string.

The main cause of insulating effects of the

topography for the Rayleigh waves is the

splitting of Rayleigh waves while crossing

the particular ridge or valley as well as

strong diffraction of Rayleigh waves in the

form of body waves at the base and top

corners of the ridges and valleys. The ASR

value at the top of first ridge of the T3RV

model for the horizontal and vertical

component is 3.12 and 0.63, respectively.

The insulating capacity for the horizontal

components of Rayleigh wave of the hill

T3RV model has decreased the ASR value

to an order of 0.045 at R25. It is clear from

the figures 4 & 5 that the insulating effect

of topography is proportional to the number

of ridges and valleys in the path of the

Rayleigh wave. The value of ASR for the

horizontal components at last receiver point

(R32) in the case of the T1R, T1RV,

T1RVR, and T3RV topography models was

0.43, 0.14, 0.12 and 0.06, respectively.

5. CONCLUSIONS

Analysis of the Rayleigh wave

simulated responses of the single ridge

model revealed the amplification of

horizontal component and de-amplification

of the vertical component of Rayleigh wave

at the top of the ridge (Savage, 2004). The

obtained ASR for the horizontal component

at the top of T1R model was of the order of

3.1. The insulating effect of topography

was proportional to the number of ridges

and valleys falling into the path of the

Rayleigh wave. For example, ASR for the

horizontal components after crossing the

T1R, T1RV, T1RVR, and T3RV

topography models was

0.43, 0.14, 0.12 and 0.06, respectively. The insulating effect of a string of topography was more for the

Rayleigh wave whose wavelength was lesser than or comparable to the width of a particular ridge and

valley (Ma et al., 2007).

Figure 5: Spectral Ratio of horizontal and vertical component of

Rayleigh wave at receivers placed at the top of ridges and base of

valleys in T3RV model. Value of ASR is represented in brackets.

14



REFERENCES

[1] Aki, K. Local site effect on strong ground motion, proceeding earthquake engineering and soil dynamics II-recent advances in Ground Motion Evaluation, ASCE, 1988, pp. 103-155.

[2] Gao, Y., Ning Zhang, Dayong Li, Hanlong Liu, Yuanqiang Cai, and Yongxin Wu. Effects of topographic amplifications induced by U-shaped canyon on seismic waves, BSSA, 2012, pp. 1748-1763.

[3] Geli, L., Bard, P.Y. and Beatrice, J.The effect of topography on earthquake ground motion: a review and new results, BSSA, 1988, 78, pp. 42-63.

[4] Kamalian M., Jafari M. K., Sohrabi-Bidar A., Razmkhah, A. and Gatmiri, B. Time- Domain Two-Dimensional Site Response Analysis of Non-Homogeneous Topographic Structures by A Hybrid FE/BE Method, SDEE, 2006, 26, pp. 753-765.

[5] Ma Shou, Archuleta, R.J. and Page, M.T. Effects of Large Scale Surface Topography on Ground Motions, as Demonstrated by A Study of the San Gabriel Mountains Los Angeles, California, BSSA, 2007, 97, pp. 2066-2079.

[6] Narayan, J.P. and Rai, D.C. An observational study of local site effects in Chamoli earthquake, Proceedings of ‘Workshop on recent earthquakes of Chamoli and Bhu’, Indian Society of Earthquake Technology, Roorkee, 2001, pp. 273-280

[7] Narayan, J.P. and Rao P.V. Prasad. Two and half dimensional simulation of ridge effects on the ground motion characteristics, Pure and Applied Geophysics, 2003, 160, pp. 1557-1571.

[8] Narayan, J.P. and Kumar, V. P-SV wave time-domain finite-difference algorithm with realistic damping and a combined study of effects of sediment rheology and basement focusing, Acta Geophysica, 2014, 62, pp. 1224-1245.

[9] Narayan, J.P. and Kumar, V. A numerical study of effects of ridge-weathering and ridge-shape-ratio on the ground motion characteristics, J. Seismo., 2015, 19, pp. 83-104.

[10] Pedersen, H. A., LeBrun, B., Hatzfeld, D., Campillo, M. and Bard, P.Y. Ground motion amplitude across ridge BSSA, 1994, 84, pp. 1786–1800.

[11] Sanchez-Sesma, F. and Campllo M. Topographic effects for Incident P, SV and Rayleigh waves, Techno-physics, 1993, pp 113-125

[12] Savage, W.Z., An Exact Solution for Effects of Topography on Free Rayleigh Waves, BSSA, 2004, 94, pp. 1706-1727.

[13] Spudich, P., Hellweg, M. and Lee, W. H. K. Directional topographic site response at Tarzana observed in aftershocks of the 1994 Northridge, California, earthquake: implications for main shock motions, BSSA, 1996, 86, 193–208.

[14] Zhao C. Coupled method of finite and dynamic infinite elements for simulating wave propagation in elastic solids involving infinite domain. Sci. China Tech. Sci., 2010, 53, pp. 1678−1687.

15



The Utilization of Local Available Material

in Base and Sub-base Layer

Sudhir Narayan Bharati1*

, Aditya Kumar Anupam1

1 Department of Civil Engineering, National Institute of Technology,Uttarakhand

*Email id: - [email protected], [email protected]

3 ABSTRACT

Nowadays there is a huge scarcity of aggregate materials. For the saving purpose of costly aggregates it

is desirable to use the local available natural materials since it will be cost effective. In the present study, the

local natural materials are Alaknanda river bed material (RBM), local aggregates and debris materials. The

various properties like gradation, moisture content, specific gravity, water absorption, crushing strength,

abrasion value, impact value are analyzed. On the basis of these properties we analyze whether the material

is suitable or not for the replacement of 100% base and sub-base material. If suitability not found, then the

properties which are not found to be satisfactory will be assessed and tend to improve these properties by

replacing the local material to conventional aggregates. For the testing purpose the tests which will be

performed are different physical properties including CBR.

Key Words: Pavement, River bed material, Sub-base course, Base course

1. INTRODUCTION

The literature review mainly focused on the waste material use in base layer, sub base layer of the

pavement. Different author was proposed different type of waste material used in pavement and also test was

performed to check the basic criteria of the pavement. Taha et al. [1] investigate the laboratory evaluation of

RAP and virgin aggregate blends and to make recommendations about its potential use as road base and sub-

base materials for highway construction in the Sultanate of Oman. Shahu et al. [2] study to quantify the

influence of important factors such as fly ash content, dolime content, and curing period on the shear

strength and stiffness characteristics of copper slag –fly ash –dolime (CFD) mix for its effective utilization in

the base course of flexible pavement. Mathur et al.[3] studied the physical and chemical characteristics of

various steel plant solid wastes such as air-cooled slag, steel slag, and granulated slag have been discussed.

Taha et al. [4] proposed A pavement design analysis of using various cement stabilized RAP-virgin

aggregate mixtures as base materials. Portland cement with RAP-virgin aggregate, mixtures in the road base

construction in the Sultanate of Oman. Rakshvir et al. [5] study various physical and mechanical properties

of recycled concrete aggregates were examined. Recycled concrete aggregates are different from natural

aggregates and concrete made from them has specific properties. Kumar et al. [6]evaluate the various

properties like modulus of elasticity, resilient strain, permanent strain, compressive strength, shear strength,

failure load. Pattanaik et al. [7] study about symbolic regression with genetic programming was used to

develop the empirical model for BPN by using experimental observations. The developed model for BPN is

be able to predict the skid resistance of the pavement satisfactorily, irrespective of the type of aggregate

gradation, binder, and aggregate sources. Mohammadinia et al. [8] investigated that cement-treated

construction and demolition (C&D) materials are viable construction materials for pavement base/sub-base

applications. Mohammadinia et al. [9] studied the geotechnical properties of geopolymer-stabilized C&D

materials were evaluated to assess their performance in pavement base sub-base applications. Mohammad

et al. [10] study to evaluate the effect that providing a stronger and more durable base or sub-base layer will

have on the performance of a pavement. Lav et al.[11] Utilizing an accelerated full scale road test data for

the fatigue performance of cement stabilized fly ash and performing a mechanistic-empirical design

procedure, required layer thickness for different lives were obtained for different amount of cement content.

Kumar et al. [12] investigate the stress-strain behaviour of the four most frequently encountered local

materials that can be utilized in the lower layer sub-base of a pavement. Cho et al. [13] reported the research

concentrates on the application of waste aggregates to highway pavement, and in particular to the surface

slab and lean concrete sub-base. Basic material properties of waste aggregates including strength of concrete



16



were measured. Ahmed et al. [14] proposed the influence of mixture variables on the mechanical properties of

cement treated recycled aggregate (CTRA). Toutanji [15] proposed that using rubber tire chips and particles

as a replacement for the mineral aggregates in Portland-cement concrete and study the compressive and

flexural strengths of rubber tire concrete are evaluated, and the effect of the volume contents of the rubber tire

chips on these strengths is also examined

2. METHODOLOGY

Figure 1: Flow Chart of Methodology

3. COLLECTION OF MATERIAL

The debris materials which used in this study collected from the four different locations nearby Srinagar

Uttarakhand and the one more material is Alaknanda riverbed material.

Figure 2: Material 1 Figure 3: Location 1


Physical Property

Collection of

Materials

Material Testing

Result And Discussion

Conclusion

Strength

property

17





4. RESULT AND DISCUSSION

4.1 Impact Testing

With respect to concrete aggregates, toughness is usually considered the resistance of the material to

failure by impact. a sample of standard aggregate kept in a mould is subjected to fifteen blows of a metal

hammer of weight 14kgs. Falling from a height of 38cms. The quantity of finer material (passing through

2.36 mm) resulting from pounding will indicate the toughness of the sample of aggregate. The ratio of the

weight of the fines (finer than 2.36 mm size) formed, to the weight of the total sample taken is expressed as a

percentage.

According to the ministry of road transportation and highways (MORTH)[16], government of India has

specified the aggregate impact value should not normally exceed 30% for aggregate to be used in wearing

course of pavements. The maximum permissible value is 35% for bituminous macadam and 40% for water

bound macadam base course.

Table 1 Result of Impact values

Materials Avg. Impact Value (%)

M1 22.146

M2 14.24

M3 17.339

M4 41.85

M5 17.139

18



The impact values obtained after the test shows that material M1, M2, M3and M5 are satisfy the

upper limits for surface course as well as base course also so these material are suitable in impact value for

base, sub-base and surface course also but the material M4 is not suitable for pavement materials.

4.2 LOS ANGLES ABRASION TEST

Los Angeles test was developed to overcome some of the defects found in Deval test. Los Angeles test is

characterized by the quickness with which a sample of aggregate may be tested. The involves taking specified

quantity of standard size material along with specified number of abrasive applicability of the method to all

types of commonly used aggregate makes this method popular. The test charge in a standard cylinder and

revolving if for certain specified revolutions. The particles smaller than 1.7 mm size is separated out. The

loss in weight expressed as percentage of the original weight taken gives the abrasion value of the aggregate.

As per the given specifications of MORTH[16] the abrasion value should not be more than 30 percent for

wearing surfaces and not more than 40 per cent for sub-base and base layer.

Table 2 Results of Abrasion Test

Material Abrasion value (%)

M1 35.36

M2 21.64

M3 30.90

M4 60.62

M5 23.56

The above results of abrasion test shows that material M2 and M5 are suitable for base sub-base and

surface course also but material M1 and M3 are suitable only for base and sub-base layers the material M4 is

not suitable for any layers.

4.3 CRUSHING VALUE TEST

Aggregate crushing value gives a relative measure of the resistance of an aggregate sample to

crushing under gradually applied compressive load. Generally, this test is made on single sized aggregate

passing 12.5 mm and retained on 10 mm sieve. The aggregate is placed in a cylindrical mould and a load of

40 ton is applied through a plunger. The material crushed to finer than 2.36 mm is separated and expressed

as a percentage of the original weight taken in the mould. This percentage is referred as aggregate crushing

value. According to the IRC and BIS the crushing value to be used in base course shall not exceed 45% and

the value for surface course shall not be more than 30% for cement concrete pavement. There is no any

specification which is provided by the MORTH for the flexible pavement.

Table 3 Results of Crushing Value

Materials Crushing Value (%)

M1 27.79

M2 21.35

M3 25.50

M4 41.92

M5 21.14

19



So on the bases of IRC all the materials are usable in all layers inspite of material M4.

4.4 SPECIFIC GRAVITY AND WATER ABSORPTION TEST

The specific gravity of an aggregate is considered to be a measure of strength or quality of the material.

Stone having low specific gravity are generally weaker than those with higher specific gravity values. The

specific gravity test helps in identification of stone. The specific gravity value of aggregate are made use of

for making weight-volume conversions and for calculating the void content in compacted bituminous mixes.

According to the MORTH and IRC the specific gravity of aggregate used in road construction range from 2.5

to 3.2 with an average value about 2.70.

Water absorption gives an idea of strength of rock. Stone having more water absorption are more porous

in nature and are generally considered unsuitable unless they are found to be acceptable based on strength,

impact and hardness tests. The acceptable range of water absorption is 0.1 to 2 percent. Up to 1 percent for

aggregate used in bituminous surface dressing and up to 2 percent for base course.

Table 4 Impact Abrasion Crushing Test

Materials Specific Gravity Apparent Specific

Gravity

Water

Absorption (%)

M1 2.647 2.683 0.5

M2 2.694 2.825 1.7

M3 2.988 3.121 1.4

M4 2.60 2.843 3.2

M5 2.69 2.76 0.92

As per the MORTH specifications the water absorption value of material M1 and M5 is less than 1%

so these material may be suitable for surface as well as base layers also but material M2 and M3 are suitable

only for base layers and the material M4 is not suitable for both.

4.5 GRADING OF AGGREGATE

Good grading implies that a sample of aggregates contains all standard fractions of aggregate in

required proportion such that the sample contains minimum voids. A sample of the well graded aggregate

containing minimum voids will require minimum paste to fill up the voids in the aggregates.

4 4.5.1 SIEVE ANALYSIS

The operation of dividing a sample of aggregate into various fractions each consisting of particles of the

same size. The sieve analysis is conducted to determine the particle size distribution in a sample of aggregate,

which we call gradation.

20



Table 5 Sieve Analysis of Material 1

Is Sieve

Weight of Aggregate Retained

(gm)

%Age of

Total

Weight

Retained

Cumulative

% of Total

Weight

Retained

%Passing

1 2 3 Avg

40mm 84 0 0 28 0.56 0.56 99.44

20mm 749 583 614 648.67 12.973 13.53

3 86.467

16mm 436 660 416 504 10.08 23.61

3 76.387

12.5mm 586 753 629 656 13.12 36.73

3 63.267

10mm 410 435 313 386 7.72 44.45

3 55.547

4.75mm 1228 1122 1272 1207.33 24.147 68.6 31.4

PAN 1507 1447 1756 1570 31.4 100 0


Is Sieve


(gm)

%Age of

Total

Weight

Retained

Cumulative %

of Total Weight

Retained

%Passing

1 2 3 Avg

40mm 84 0 0 28 0.56 0.56 99.44

20mm 749 583 614 648.67 12.973 13.533 86.467

16mm 436 660 416 504 10.08 23.613 76.387

12.5mm 586 753 629 656 13.12 36.733 63.267

10mm 410 435 313 386 7.72 44.453 55.547

4.75mm 1228 1122 1272 1207.33 24.147 68.6 31.4

PAN 1507 1447 1756 1570 31.4 100 0


Is Sieve


(gm)

% of Total

Weight

Retained

Cumulative % of

Total Weight

Retained

%Passing

1 2 3 Avg

40mm 0 0 0 0 0 0 100

20mm 1395 1344 1160 1299.67 25.99 25.99 74.01

16mm 886 737 775 799.33 15.99 41.98 58.02

12.5mm 725 978 972 891.67 17.83 59.81 40.19

10mm 780 581 648 669.67 13.39 73.20 26.80

4.75mm 981 977 915 957.67 19.15 92.35 7.65

PAN 233 383 530 382 7.64 99.99 0

21




Is Sieve

Weight of Aggregate

Retained (gm) % of Total

Weight

Retained

Cumulative % of

Total Weight

Retained

%Passing

1 2 3 Avg

40mm 0 0 97 32.33 0.65 0.65 99.35

20mm 1161 756 877 931.33 18.63 19.28 80.72

16mm 774 552 511 612.33 12.25 31.53 68.47

12.5mm 912 489 588 663 13.26 44.79 55.21

10mm 651 527 589 589 11.78 56.57 43.43

4.75mm 1123 1975 1665 1587.67 31.75 88.32 11.68

PAN 379 701 673 584.33 11.69 100 0


Is Sieve

Weight of Aggregate

Retained (gm)

% of Total

Weight

Retained

Cumulative % of

Total Weight

Retained

%Passing

1 2 3 Avg

40mm 0 0 0 0 0 0 100

20mm 609 875 705 729.67 14.59 14.59 85.41

16mm 1265 948 940 1051 21.02 35.61 64.39

12.5mm 1224 1002 902 1042.67 20.85 56.46 43.54

10mm 666 676 644 662 13.24 69.7 30.3

4.75mm 1171 1440 1710 1440.33 28.81 98.51 1.49

PAN 65 59 99 74.33 1.49 100 0

Figure 11: Comparison of Normal Gradation Graph

22



Figure 12: Comparison of Modified Gradation Graph

4.6 FLAKINESS INDEX TEST

The flakiness index of aggregate is the percentage by weight of particles in it whose least dimension

(thickness) is less than three-fifths of their mean dimension. The test is not applicable to sizes smaller than 6.3

mm.

Table 10 Result of Flakiness Index

Material Flakiness Index (%)

M1 14.59

M2 34.01

M3 39.95

M4 56.37

M5 17.41

5 4.7 ELONGATION INDEX

The elongation index on an aggregate is the percentage by weight of particles whose greatest dimension

(length) is greater than 1.8 times their mean dimension. The elongation index is not applicable to sizes

smaller than 6.3 mm. This test is conducted by using metal length gauge. A sufficient quantity of aggregate

is taken to provide a minimum number of 200 pieces of any fraction to be tested. Each fraction shall be

gauged individually for length on the metal gauge. The total amount retained by the gauge length shall be

weighed.

Table 11 Results of Elongation Index

Material Elongation index (%)

M1 34.73

M2 49.81

M3 54.39

M4 64.17

M5 43.36

23



6 4.8. FINAL RESULTS

Table 12 Physical Properties of All The Procured Materials.

Materials Impact

Value %

Crushing

Value %

Abrasion

Value %

Specific

Gravity

Apparent

Specific

Gravity

Water

Absorp -tion %

Flakiness

Index

Elongatio

n

Index%

M1

22.146

27.794

35.36

2.647

2.683

0.5

14.59

34.73

M2 14.24 20.82 21.64 2.694 2.825 1.7 34.01 49.81

M3 17.339 25.5 30.9 2.988 3.121 1.4 39.95 54.39

M4 41.85 41.92 60.62 2.6 2.843 3.2 56.37 64.17

M5 17.136 21.141 23.56 2.692 2.761 0.92 17.41 43.36

5 CONCLUSION

The material M5 is satisfying all the physical properties for the base and surface layers.

The material M4 not fulfil the criteria give by MORTH for surface layer as well as base layers. Material

M1 is suitable only for base layers Cause the abrasion value is higher than 30%.

Material M2 has higher water absorption value than 1% so this material is not suitable for surface course

but it is suitable for base layers.

Material M3 is suitable for base layers cause the abrasion and water abrasion value is higher than surface

layer criteria.

In future the strength property will be find out with the help of California bearing ratio (CBR) test and we

also find out that the material can be used directly in the surface layer if not then at which extent we can use

these materials.

7 REFERENCES

[1] Taha, Ramzi, et al. "Evaluation of reclaimed asphalt pavement aggregate in road bases and subbases." Transportation Research Record: Journal of the Transportation Research Board 1652 (1999): 264-269.

[2] Shahu, J. T., S. Patel, and A. Senapati. "Engineering properties of copper slag–fly ash–dolime mix and its utilization in the base course of flexible pavements." Journal of Materials in Civil Engineering 25.12 (2012): 1871-1879.

[3] Mathur, Sudhir, S. Soni, and A. V. S. R. Murty. "Utilization of industrial wastes in low-volume roads." Transportation Research Record: Journal of the Transportation Research Board 1652 (1999): 246-256.

[4] Taha, Ramzi, et al. "Cement stabilization of reclaimed asphalt pavement aggregate for road bases and subbases." Journal of materials in civil engineering 14.3 (2002): 239-245.

[5] Rakshvir, Major, and Sudhirkumar V. Barai. "Studies on recycled aggregates-based concrete." Waste Management & Research24.3 (2006): 225-233.

[6] Kumar, Praveen, and Shashi Kant Sharma. "Prediction of Equivalency Factors for Various Subbase and Base Courses." Journal of Materials in Civil Engineering 25.10 (2012): 1357-1365.

[7] Pattanaik, MadhuLisha, RajanChoudhary, and Bimlesh Kumar. "Evaluation of Frictional Pavement Resistance as a Function of Aggregate Physical Properties." Journal of Transportation Engineering, Part B: Pavements 143.2 (2017): 04017003.

[8] Mohammadinia, Alireza, et al. "Laboratory evaluation of the use of cement-treated construction and demolition materials in pavement base and subbase applications." Journal of Materials in Civil Engineering 27.6 (2014): 04014186.

[9] Mohammadinia, Alireza, et al. "Stabilization of demolition materials for pavement base/subbase applications using fly ash and slag geopolymers." Journal of Materials in Civil Engineering28.7 (2016): 04016033.



[10] Mohammad, Louay, and ShadiSaadeh. "Performance evaluation of stabilized base and subbase material." GeoCongress 2008: Geosustainability and Geohazard Mitigation. 2008. 1073-1080.

[11] Lav, A. Hilmi, M. AysenLav, and A. BurakGoktepe. "Analysis and design of a stabilized fly ash as pavement base material." Fuel85.16 (2006): 2359-2370.

[12] Kumar, Praveen, Satish Chandra, and R. Vishal. "Comparative study of different subbase materials." Journal of Materials in Civil Engineering 18.4 (2006): 576-580.

[13] Cho, Yoon-Ho, and Sung-Hun Yeo. "Application of recycled waste aggregate to lean concrete subbase in highway pavement." Canadian Journal of Civil Engineering 31.6 (2004): 1101-1108.

[14] Behiry, Ahmed Ebrahim Abu El-Maaty. "Utilization of cement treated recycled concrete aggregates as base or subbase layer in Egypt." Ain Shams Engineering Journal 4.4 (2013): 661-673.

[15] Toutanji, Houssam A. "The use of rubber tire particles in concrete to replace mineral aggregates." Cement and Concrete Composites 18.2 (1996): 135-139.

[16] Ministry of road transport and highways specifications for road and bridge works 5th revision.

25



Study of physical, chemical and engineering

behaviour of municipal solid waste of Bhimavaram, India Vamsi Nagaraju. T

1, Venkata Rao. M

2 and Krishnam Raju.

G.L.V3

1Assistant Professor of Department of Civil Engineering, S.R.K.R Engineering College,

[email protected]

2,3Assistant Professor of Department of Civil Engineering, S.R.K.R Engineering

College, Bhimavaram-534204

ABSTRACT

Municipal solid waste (MSW) is heterogeneous nature, and has emerged as a big challenge not only

because of the health and environment concerns but also due to huge quantities of waste generated. The

stability of landfill is governed by engineering properties of MSW. These properties play a vital role in

interactions within the landfill system involving the waste body and landfill structure: cover liner, MSW,

leachate, leachate collection system and gas collection system. The severity of landfill is governed by

leachate, which causes significant threat to surface water and ground water. In this paper presents the

investigation of the quantity, chemical characteristics and geotechnical properties of MSW. Quantity analysis

and chemical composition tests were conducted. And also other tests conducted are moisture content, grain-

size distribution, compaction, shear strength and consolidation tests. The waste samples for the tests were

collected from the sites located on the outskirts of Bhimavaram. The influence of those engineering

properties on the stability of the landfill, and chemical properties on severity of leachate in the designing

of landfill were discussed.

Keywords: Municipal solid waste, leachate, geotechnical properties of MSW, environment

8 INTRODUCTION

In India, municipal solid waste (MSW) production dramatically increases rapidly, keeping pace with the

massive urbanization and rapid industrialization, and also emerged as a big challenge not only because of the

health and environmental concerns but also due to huge quantities of waste generated (Syamala and Satpal,

2015). Until recently, landfills have been the primary method of municipal solid waste (MSW) management.

However, although land filling is one of the cheapest ways of disposing of MSW, there is a risk that serious

environmental problems may result from contaminated sites in the future (Ruokojarvi et al. 1995; Hansen

and O’Keefe, 1996). In order to give a push to MSWM in cities, the Central government of India, has designed schemes

under Ministry of New and Renewable Energy (MNRE) to promote waste to energy projects. Some of State

governments of Andhra Pradesh, Haryana, Gujarat, Karnataka, Maharashtra, Madhya Pradesh, Rajasthan,

Tamil Nadu and Uttar Pradesh have announced policy measures pertaining to allotment of land, supply of

garbage, and facilities for evacuation, sale and purchase of power to encourage the setting up of waste to

energy projects.

As a general rule, leachate is characterized by high values of COD, pH, ammonia nitrogen and heavy

metals, as well as strong color and bad odor. At the same time, the characteristics of the leachate also vary

with regard to its composition and volume, and biodegradable matter present in the leachate against time

(Malina and Pohland, 1996; Im et al. 2001). All these factors make leachate treatment difficult and

complicated.

Engineering properties of waste such as density, moisture content, unit weight, hydraulic conductivity,

compressibility and shear strength are the basis in designing of the engineered landfills as the knowledge

of these properties helps in assessing the settlement and potential modes of failure. The composition of

Municipal Solid Waste (MSW) is very important in this aspect as it influences some of engineering

properties of the waste. The percentage of organic content in MSW may affect the settlement due to the


26



degradation of organics along with time. The permeability of the MSW may be hindered by the large

plastics and other plastic like materials present in MSW. Incorrect estimation of the MSW permeability

may lead to leachate accumulation in some parts of the landfill, resulting in a non-uniform degradation of

the waste which can cause differential settlement and structural failure of the landfill components. In

India bulk of the waste is being landfilled. The strength, stability and the settlement depends on the

composition of the waste. Perhaps this is a first comprehensive approach towards the geotechnical

characterisation of MSW in India (Penmethsa, K.K, 2007).

This paper presents the laboratory data of solid waste of bhimavaram. The influence of those

engineering properties on the stability of the landfill, and chemical properties on severity of leachate in the

designing of landfill were discussed.

EXPERIMENTAL INVESTIGATION

Material collection

The samples were collected from different locations of open dumps at the outskirts of bhimavaram.

Samples collected were sent to laboratory for analysis. After weighing each sample accurately, composite

samples of each category were prepared for composition and physico-chemical analysis.The maximum

percent passing was 78% through the 4.75 mm sieve. The Cu and Cc values for the samples were

calculated as and 12 and 1.86. The values indicate that the samples are well graded and the absence of

coarse sand and clay like particles. The MSW constituted of fine sand and clay like particles.

Characterisation of solid waste

The physical characterisation of the fresh and the aged MSW (after composting) is done in order to

measure the quantity of the recoverable and to study the effect of the physical composition on the strength

and stability characteristics of the MSW. The waste is segregated by hand sorting into paper, plastics, inerts

(rubber, leather), Glass, stones and the organic fraction of the waste. The physical characterisation of the

waste passing through 63mm was done by hand sorting and on the weight basis. The age of the sample was

4-5 weeks. The quantity of waste taken for composition analysis was 10 kg. The MSW samples used for

all the experiments were those passing through the 16mm trammel and retained by the 4mm trammel. Therefore the composition analysis of the 4mm trommel retained waste was done and

mentioned below figure.

3.75

4

.15

1

0.4

16.8

5.05

1.05

52.4

6.4

Food waste paper

Plastic

Rags/cloth

green waste/ coconuts metals/glass/ceramics

soil/earth

others

Characterisation of MSW



Chemical composition of solid waste

Solid waste landfills may cause severe environmental impacts if leachate and gas emissions are not

controlled. Leachate generated in municipal landfill contains large amounts of organic and inorganic

contaminants (Kettunen and Rintala, 1998). Leachate is collected from the solid waste and the leachate

composition will be as given in Table 1.

Table: 1 Chemical composition of solid waste

Property Sample 1 Sample 2

BOD, PPM 2200 2450

COD, PPM 4200 4840

PH 7.98 7.54

Calcium (Ca), % 0.95 0.88

Sodium (Na), % 2.05 2.56

Potassium (K), % 1.65 1.12

Phosphate (PO4), % 1.35 1.42

Leachate may also have a high concentration of metals and contain some hazardous organic

chemicals. The removal of organic material based on COD, BOD and ammonium from leachate is

the usual prerequisite before discharging the leachates into natural waters (Kettunen and Rintala,

2009). The leachate composition from the transfer station can vary depending on several factors, including

the degree of compaction, waste composition, climate and moisture content in waste

Geotechnical properties of solid waste

In the present study the direct shear tests were performed with bulk density 1050 kg/m3 and for

confining pressures of 50, 100 and 150 kPa. The size of sample was 60mm in length, 60mm in width and

30mm in height. The stress-strain response of the waste are plotted and the cohesion and the friction angle

values were obtained. Test results were shown in Table-2.

Table -2 Geotechnical properties of solid waste

Property Sample-1 Sample-2

Natural moisture content 14% 13%

Bulk density (kg/m3) 190 195

Cohesion, kPa 16 18

Friction angle, degrees 38 42

There is great variability in the reported shear strengths in the literature. Cohesion values from 0

to 80 kPa and friction angles from 0–60° have been reported. The deviator stress increased constantly in the

initial stages until 30% strain and there was a sudden increase in the rate of stress from 40% to 50% strain

levels. The cohesion and the friction angle values were obtained as 18kPa and 42° for 20% deformation.

28



CONCLUSIONS

More than 92% of the MSW generated in country is directly disposed on open dumping in an unsecure

and unplanned manner. Basically in India, cities are congested and crowded and required special attention to

MSW management. Currently there is no any specific site for segregation of solid waste in

Bhimavaram.

This investigation on municipal solid waste of Bhimavaram Municipality gives field data about the

quantity and characteristics of solid waste, most of the waste quantity is food waste from the hotels.

Biodegradable content of MSW is a good source of compost for agriculture purpose. Non biodegradable

content can be used for recycle, reuse or landfill. The direct shear tests yielded values of 16-18kPa for cohesion and 38°-42° as the friction angle

respectively. These values are required for the assessment of slope stability of landfills. For example, a

slope of 1:1 and height of 30m of MSW landfill, using bulk density of 190kg/m3, cohesion of 15 kPa and

friction angle of 40° gives a factor of safety of slope in the range of 1.25 using tri-axial test data. Use of direct shear data leads to a factor of safety of 1.65. Hence proper understanding of slope stability issues in landfill is very essential and improper data or lack of data may lead to failures.

REFERENCES

[1] J.F. Malina, F.G. Pohland, Design of anaerobic processes for the treatment of industrial and municipal

wastes, Water Qual. Manage. 7 (1996) 169–175.

[2] J.H. Im, H.J. Woo, M.W. Choi, K.B. Han, C.W. Kim, Simultaneous organic and nitrogen removal

from municipal landfill leachate using an anaerobic–aerobic system, Water Res. 35 (2001) 2403–2410.

[3] Penmethsa, K.K., (2007). Permeability of Municipal Solid Waste in Bioreactor Landfill with

Degradation. Ms.C. thesis, University of Texas at Arlington, USA.

[4] R.H. Kettunen, J.A. Rintala, (1998) Performance of an on-site UASB reactor treating

leachate at low temperature, Water Res. 32, 537–546. [5] R.H. Kettunen, T.H. Hoilijoki, J.A. Rintala, Anaerobic and sequential anaerobic–aerobic treatments of

municipal landfill leachate at low temperatures, Bioresour. Technol. 58 (2009) 40– 41.

[6] Ruokoja¨rvi P, Ettala M, Rahkonen P, Tarhanen J, Ruuskanen J (1995) Polychlorinated dibenzo-p-dioxins and -furans (PCDDs andPCDFs) in municipal waste landfill fires. Chemosphere 30:1697–1708

[7] Shyamala Mani and Satpal Singh (2015) Sustainable Municipal Solid Waste Management in India: A Policy Agenda, International Conference on Solid Waste Management, 5 Icon SWM 2015

29



Factorial Design based Stability Analysis of Infinite Slopes

Vikas Pratap Singh1

1Assistant Professor, Department of Civil Engineering, Institute of Infrastructure, Technology, Research and

Management, Ahmedabad – 380026, Gujarat, India. Email: [email protected]

ABSTRACT

Stability of infinite slopes is of paramount importance in hilly areas, as an unstable slope might result in

catastrophic landslides. In practice, limit equilibrium approach forms the underlying principle for majority of

the conventional stability analysis methods for infinite slopes. One of the major limitations of the limit

equilibrium based methods is that these methods fail to account for inherent variability of in-situ soil and its

influence on the assessment of slope stability. This study presents a simple approach to account for

heterogeneity of soil in slope stability analysis. In order to incorporate the influence of in-situ soil variability

in conventional limit equilibrium approach, factorial design of experiments methodology is used in

conjunction. Factors of safety for slope stability are computed analytically using in-situ soil properties at

various levels as decided in accordance with factorial design methodology. The in-situ soil properties

considered as variables included cohesion, angle of internal friction and unit weight. Using the computed

factor of safety values, a regression model is fitted-in. The factor of safety regression model is then utilised

to study the influence of variability in in-situ soil properties on the assessment of the infinite slope stability.

It is evident from the various observations that the methodology adopted provided a significant insight

into the role of in-situ soil variability in slope stability analysis. From the study, it can be concluded that the

variability in in-situ soil internal friction angle is most critical to slope stability followed by cohesion and the

least influenced by the unit weight. Also, it can be noted that the both cohesion and friction angle have a

positive influence on slope stability, on the other hand, unit weight has a negative influence.

Key Words: Factorial Design, Regression model, Infinite Slope, Stability Analysis

1. INTRODUCTION

The hilly terrains in India are often subjected to the landslides due to the failures of naturally occurring

infinite soil slopes [1]. In practice, the stability analysis of natural slopes is conducted using limit equilibrium

approach [2]. The safety of the slope is demarcated based on a factor of safety defined as the ratio of

resisting to driving forces on a potential sliding surface. A factor of safety greater than one indicates a safe

slope. Further, it is well established that even within uniform soil layers the properties of natural soil deposits

vary considerably [3]. Consequently, it is essential that the influence of variability of in-situ soil properties

be accounted in the slope stability analysis. In this context, many studies such as [4-6] are available on the

use probabilistic and reliability methods for slope stability analysis.

In this study, a simple methodology is demonstrated to incorporate the influence of in-situ variability

in the assessment of the stability of finite slopes. Factorial design of experiments [7] is used in conjunction

with conventional limited equilibrium method. Using the factor of safety values computed in accordance

with factorial design, a regression model is fitted-in and used to study the influence of soil variability on

slope stability.

2. CONVENTIONAL INFINITE SLOPE STABILITY ANALYSIS

Figure 1 shows a schematic diagram of an infinite slope. The analysis [2] is conducted by considering

a vertical slice ‘1234’of weight W, width b, unit dimension normal to the plane of paper and depth of failure

plane z. The shear strength of the slope along failure plane is given by Mohr-Coulomb failure criterion. In

Figure 1, N and T are directions normal and tangential (or parallel) to the failure plane, respectively. The

various forces are resolved along the failure plane (i.e. tangential or T-direction), and resisting and driving

components are determined. Stability of slope is then defined by a factor of safety (FS) obtained by taking

ratio of resisting forces and driving forces. In this study, slope is assumed to be in dry state, for which the

factor of safety (FS) is given by the Eq. (1).

30



(1)

where c, ϕ and γ represents the in-situ soil cohesion, angle of internal friction, and unit weight, respectively;

z = depth of potential failure plane, i = slope angle with respect to the horizontal.

Figure 1: Typical Schematic of an Infinite Slope

3. FACTORIAL DESIGN BASED INFINITE SLOPE STABIITY REGRESSION MODEL

Montgomery [7] discusses in detail about the factorial design of experiments widely used in the

experiments involving several factors. A special case of factorial design is that of k factors, each at only two

levels. These levels may be quantitative or qualitative. A complete replicate of such a design requires 2 x 2 x

…x 2 = 2k observations and is called a 2

k factorial design. The 2

k factorial design provides the smallest

number of runs with which k factors can be studied in a complete factorial design. Because there are only

two levels for each factor, it is assumed that the response is approximately linear over the range of factor

levels chosen. In the present study, soil parameters c, ϕ and γ each at two levels are considered as the three

design factors (i.e. k = 3) for the experimental design and therefore, further discussion is restricted to the

method of 23 factorial design of experiments. Sub-sections 3.1 to 3.4 provide a brief on the procedure

involved in the development of factor of safety regression model (i.e. FS-model) using 23 factorial design of

experiments.

3.1. Fixing Levels for Design Factors

The 23 factorial design of experiments need to specify values of each factor at two levels i.e. high and

low. These levels are fixed based on the 95% confidence intervals [8] such that the low level xL and high

level xh values are related to mean value μ and standard deviation σ with the relationships xL = μ-1.65σ and

xh = μ+1.65σ, respectively. Coefficients of Variation (COV) are adopted from literature [3]. Table 1 presents

statistical details of the three design factors considered in study.

Table 1 Statistical Information on Study Variables

Study Variable

(or Design Factor)

Design

Notation

Coefficient of

Variation,

COV

Mean,

μ

Standard

Deviation, σ

Low Level

Value, xL

High Level

Value, xh

Cohesion, c (kPa) A 0.12 5.0 0.60 4.01 5.99

Friction Angle, ϕ (o) B 0.06 35.0 2.10 31.54 38.47

Unit Weight, γ (kN/m3) C 0.06 18.9 1.13 17.04 20.76

Note: xL= μ – 1.65σ ; xh = μ +1.65σ ; σ = μ.COV

3.2. Combinations for 23 Factorial Design of Experiments

Following the standard notations, three design factors namely in-situ soil cohesion c, angle of internal

friction of in-situ soil ϕ and in-situ soil unit weight γ are represented as the A, B and C, respectively. Table 2

shows the eight design runs for the 23 design using the “+ and -” notation to represent the low and high levels

2c zcos i tanFS

zcosisin i

31



of the factors. Factor combinations in the design are usually represented by lower case letters. High level of

any factor in the factor combination is denoted by the corresponding lower case letter and that the low level

of any factor in the factor combination is denoted by the absence of the corresponding letter. For example, a

represents the factor combination of A at high level and B, C at low level, b represents B at high level and A,

C at low level, ab represents A, B at high level and C at low level and so on. By convention, (1) is used to

denote all factors A, B, and C at the low level. In Table 2 column (v) i.e. ‘Run label’ indicates the standard

order of eight experimental run labels for different factor combinations as (1), a, b, ab, c, ac, bc, and abc.

Table 2 Design Run Label, Algebraic Signs for Contrast Constants and Computed Factors of Safety

Run

Number

Factor Run

Label

Algebraic Signs for Determination of Contrast

Constants

Computed

Factor of

Safety, FS A B C A B AB C AC BC ABC Column (i) (ii) (iii) (iv) (v) (vi) (vii) (viii) (ix) (x) (xi) (xii) (xiii)

1 - - - (1) - - + - + + - 1.61

2 + - - a + - - - - + + 1.87

3 - + - b - + - - + - + 1.92

4 + + - ab + + + - - - - 2.19

5 - - + c - - + + - - + 1.51

6 + - + ac + - - + + - - 1.73

7 - + + bc - + - + - + - 1.82

8 + + + abc + + + + + + + 2.04

3.3. Experimental Runs and Statistical Analysis

In the present study, an ‘experiment run’ refers to the analytical determination of the factor of safety

values for the infinite slope stability using Eq. (1). For illustration, slope is assumed to be in dry state with

slope angle equal to 300 and depth of failure plane z = 1 m. Table 2 column (xiii) summarises the factor of

safety (FS) determined in accordance with the eight design combinations. To develop a FS regression model,

parameters such as contrast constants, effect estimates and percent contribution of main factors (A, B and C)

and interaction factors (AB, AC, BC and ABC) are to be determined for identifying the most significant

main factors / interaction factors. For example, contrast constant for main factor A is equal to algebraic sum

of observations from eight experimental runs. Column (vi) of Table 2 shows the algebraic sign convention

for the algebraic sum to determine contrast of A. Therefore,

Contrast A 1 a b ab c ac bc abc (2)

Further, for the main factor A, the effect estimate and percent contribution to the response quantity are

calculated using Eqs. (3)-(6).

1Effectestimate, A [Contrast A]

4n (3)

A

T

SSPercentcontribution A

SS (4)

2

A

Contrast ASS

8n (5)

where n = number of experiment replicates (= 1 in the present case), SSA = sum of squares for A and

TSS total sum of squares given by

T

squareof sumof allobservationsSS sumof squareof eachobservation

8n

(6)

Table 3 summarises the above parameters computed for all main/interaction factors. The verification of

the main/interaction factors identified above is also done by plotting a normal probability of the effect

estimates [7]. Figure 2 shows the normal probability plot of effect estimates in the present case. The

significant factors are those that lie far away from the straight line in the probability plot. Thus, both from

Table 3 and Figure 2, the important factors that emerge out for factor of safety are A (i.e. in-situ soil

cohesion c), B (i.e. angle of internal friction ϕ) and C (i.e. unit weight of in-situ γ).

32



Table 3 Summary of Factor Effect Estimates and Percent Contributions

Model Term Design Factor Factor of Safety Model (FS-Model)

Effect Estimate Sum of Squares Percent Contribution

A C 0.24 0.12 34.19

B ϕ 0.31 0.20 56.78

AB cϕ 0.00 0.00 0.00

C γ -0.12 0.03 8.72

AC cγ -0.02 0.00 0.29

BC ϕγ 0.00 0.00 0.00

ABC cϕγ 0.00 0.00 0.00

Note: Bold numbers indicate most influencing main factor/interaction factor

Figure 2: Normal Probability Plot of effect Estimates of the FS-Model

3.4. Fitting a Regression Model

Based on the observations and their subsequent analysis presented in previous sub-section, the regression

model for factor of safety (i.e. FS-model) of the infinite slope stability is

o 1 1 2 2 3 3FS x x x (7a)

1 2 3FS 1.84 0.12x 0.155x 0.06x (7b)

where βo, β1, β2, β3 and β12 are the regression coefficients (βo is the average of all eight observations of

the corresponding response quantity given in Table 2 column (xiii) and all other are one-half the effect

estimate of the corresponding main factor/interaction factor given in Table 3), and x1, x2, and x3 are the

coded factors representing main factors A, B, and C respectively. The coded factors x1, x2, and x3 can be

expressed in terms of design factors as

1x c 5 /0.99 where 4.01 kPa c 5.99 kPa (8a)

2x 35 /3.47 where 0 031.54 38.47 (8b)

3x 18.9 /1.86 where 3 317.04 kN / m 20.76kN / m (8c)

When design factors have only two levels, coded factors given by Eq. (8a-c) produce the familiar 1

notation for levels of the coded factors. For example, Eq. (8a) yields x1 = +1 when c is at high level chigh

(equal to 5.99 kN/m2), x1 = -1 when c is at low level clow (equal to 4.01 kN/m

2) and x1 = 0 when c is at mean

value (equal to 5.0 kN/m2). Coded factors also enable graphical representation of variation of different

design factors between two levels (i.e. high and low level) on the same axis.

Model adequacy can be checked by computing residuals as the difference between the factors of safety

values computed using Eq. (1) and the respective predicted values from regression model i.e. Eq. (7b). The

residuals were found to be within an acceptable range of ±0.01. The regression model (i.e. FS-model) for

predicting factor of safety of an infinite slope developed in Eq. (7) is then utilised to study the influence of

variability in-situ soil parameters on the stability of infinite slope. The results of the analyses using FS-model

are discussed in the following section.

33



4. RESULTS AND DISCUSSIONS

Figure 3 shows the influence of variability in in-situ soil cohesion on infinite slope stability. Keeping

other soil properties at their respective mean values, only the cohesion is increased from 4.01 kPa (i.e. low

level value) to 5.99 kPa (i.e. high level value). From Figure 3, it can be observed that as cohesion increases,

slope stability increases by about 14%. With respect to the factor of safety corresponding to the mean value

of all in-situ parameters, i.e. FS = 1.84, a variation of ±7% can be observed.

Figure 3: Influence of Variability in In-situ Soil Cohesion on Infinite Slope Stability

Figure 4 shows the influence of variability in in-situ soil angle of internal friction on infinite slope

stability. Keeping other soil properties at their respective mean values, only angle of internal friction is

increased from 31.540 (i.e. low level value) to 38.47

0 (i.e. high level value). From Figure 4, it can be

observed that as angle of internal friction increases, slope stability increases by about 18%. With respect to

the factor of safety corresponding to the mean value of all in-situ parameters, i.e. FS = 1.84, a variation of

±8% can be observed.

Figure 4: Influence of Variability in In-situ Soil Angle of Internal Friction on Infinite Slope Stability

Figure 5 shows the influence of variability in in-situ soil unit weight on infinite slope stability. Keeping

other soil properties at their respective mean values, only soil unit weight is increased from 17.04 kN/m3 (i.e.

low level value) to 20.76 kN/m3 (i.e. high level value). From Figure 4, it can be observed that as unit weight

of soil increases, slope stability decreases by about 7%.

Thus, based on the above following two general observations can be made: (a) the angle internal friction

is the most critical in-situ soil parameter to the slope stability, and (b) as cohesion and friction angle of soil

increases, stability of the slope increases, whereas, with the increase in unit weight of soil beyond a certain

limit, the stability of the slope decreases.

34



Figure 5: Influence of Variability in In-situ Soil Unit Weight on Infinite Slope Stability

From basics it is well known that both c and ϕ bears a positive correlation with γ, and are negatively

correlated with each other. Figure 6 plotted in terms of coded variables shows the influence of correlation (ρ)

on the stability analysis. While considering positive correlation, both the coded variables are simultaneously

increased from -1 to +1, and in case of negative correlation, one coded variable is increased from -1 to +1

and other decreased from +1 to -1 simultaneously. In Figure 6, curves ρϕ-γ and ρc-γ indicates that stability of

the slope is increased due to the positive correlation of c and ϕ with γ. On the contrary, curve ρc-ϕ indicates a

decrease in stability due to the negative correlation between c and ϕ.

Figure 6: Influence of Correlation among In-situ Soil Parameters on Slope Stability Assessment

5. CONCLUDING REMARK

The prime objective of the study was to illustrate the use of factorial design of experiments

methodology in context of slope stability analysis. It is evident from the observations of the study that

factorial design methodology provided a useful insight into the influence of in-situ soil variability in infinite

slope stability analysis, and as illustrated, could be easily used in conjunction with any conventional method

for a more exhaustive analysis.

REFERENCES

[1] N. Vasudevan, K. Ramanathan, Geological factors contributing to landslides: case studies of a few landslides in

different regions of India, IOP Conf. Series: Earth Env. Sci. 30 (2016) 012011. [2] J.M. Duncan, S.G. Wright, T.L. Brandon, Soil Strength and Slope Stability, John Wiley & Sons, New Jersey, 2014. [3] K.K. Phoon, F.H. Kulhawy, Characterization of geotechnical variability, Can. Geotech. J. 36 (1999) 612-624. [4] J.T. Christian, C.C. Ladd, G.B. Baecher, Reliability applied to slope stability analysis, J. Geotech. Eng. 120 (1994)

2180-2207. [5] B.K. Low, S. Lacasse, F. Nadim, Slope reliability analysis accounting for spatial variation, Georisk 1 (2007) 177-

189. [6] S. Lari, P. Frattini, G.B. Crosta, A probabilistic approach for landslide hazard analysis, Eng. Geol. 182 (2014) 3-14. [7] D.C. Montgomery, Design and Analysis of Experiments, Wiley, Singapore, 2001. [8] T.L.L. Orr, Selection of characteristic values and partial factors in geotechnical designs to Eurocode 7, Comput.

Geotech. 26 (2000) 263-279.

35



Seismic Stability of Reinforced Retaining Wall Using the

Pseudo-Dynamic Approach under Horizontal and Vertical

Seismic Accelerations for c-ϕ Soil Backfill

Ashish Gupta1 and Vishwas A. Sawant

2

1Research Scholar, Department of Civil Engineering, IIT Roorkee. Roorkee 247 667, India

E-mail: [email protected] 2Associate Professor, Department of Civil Engineering, IIT Roorkee, Roorkee 247 667, India

E-mail: [email protected]

ABSTRACT

Pseudo-static and pseudo-dynamic methods are, two widely used methods for analyzing the reinforced

retaining walls under seismic conditions for cohessionless soil backfill. For c-ϕ soil backfill, most of the

researchers had used pseudo-static method for the seismic stability analysis of reinforced retaining walls.

Seismic stability analysis of reinforced retaining walls using the pseudo-dynamic approach are still very

limited for c-ϕ soil backfill. In the present study, pseudo-dynamic method is used for analyzing the

reinforced retaining walls under seismic condition. In the present study propagation of primary and shear

waves under both, horizontal and vertical seismic accelerations has been considered also. A simplified

formulation has been also presented here to obtain the maximum strength of reinforcements, inclination of

failure angle and safety factor for analyzing the reinforced retaining walls under seismic condition. A

parametric study has been done to show the effect of shear strength parameters, each moment of lateral

shaking, horizontal and vertical seismic coefficients. For cohesionless backfill, numerical predictions are in

good agreement when compared with available studies in literature for validation purpose.

Key Words: Pseudo-dynamic approach, Maximum strength of reinforcement, Inclination of failure

angle, Safety factor

1. INTRODUCTION

It is seen earlier that many historic earthquakes have caused permanent deformation of various concrete

retaining walls. Sometimes these deformations were very small and sometimes the concrete retaining walls

have collapsed during earthquakes. To support the soil backfills in various civil infrastructure projects,

reinforced retaining walls have been used as the alternatives to conventional concrete retaining walls. Soil

reinforcement has become extensively used earthwork construction technique due to its technical and

economical advantages. Mononobe and Okabe did the pioneer work for determining the seismic earth

pressure using the pseudo-static method without considering the time dependent effect. Time dependent

effect was then incorporated by Steedman and Zeng (1990) for analysing the retaining walls under seismic

conditions. The method incorporated by Steedman and Zeng (1990) is known as pseudo-dynamic method.

The only drawback of this study was the consideration of finite shear waves in soil backfill. Nimbalkar et al.

(2006) and Choudhury et al. (2007) then improve the pseudo-dynamic method incorporated by Steedman

and Zeng (1990) by considering the propagation of shear and primary waves with in the soil medium.

Nimbalkar et al. (2006), Choudhury et al. (2007) and Reddy et al (2009) analysed the reinforced soil wall

considering the pseudo-dynamic method for cohessionless soil backfill. Then, Ghanbari and Ahmadabadi

(2010) proposed pseudo-dynamic approach for analysing the reinforced retaining walls under seismic

conditions considering c-ϕ soil backfill. In the present work, a simplified limit equilibrium method is used

for the analysis of reinforced retaining wall using pseudo-dynamic method considering the effect of shear

and primary waves.

2. METHODOLOGY

ABC is a reinforced retaining wall system of height H, shown in figure 1. The retaining wall is

retained the soil backfill of cohesion c and soil friction angle ϕ, having unit weight γ. Reinforced

36



retaining wall system is having n number of planar reinforcement of length Lr. Reinforcement

layers are having the spacing Sv=H/n except the reinforcements provided in the top and bottom. In

the top and bottom, 0.5 Sv spacing is provided. On the base of wall, seismic acceleration are

subjected in horizontal and vertical directionh ha k g and

v va k g . kh and kv are the seismic

coefficients in horizontal and vertical direction. Assumed failure soil wedge AB makes an angle, α.

Various forces in the soil wedge is shown in figure 2. The resultant of shear and normal force acting

on the failure wedge is F. Shear and primary wave velocity, Vs and Vp are assumed to act within the

soil backfill. / 1.87p sV V is taken as given in most of the literature. The analysis consists the period of

lateral shaking 2 /T , where ω is the angular frequency.

Figure 1: Reinforced retaining wall system Figure 2: Forces acting on the reinforced

retaining wall system in seismic condition

Tensile force generated in the reinforcement

For stabilizing the RR wall, 1

n

i

i

T

is required sum of the tensile forces, where Ti is the tension

force generated in the ith reinforcement layer. Considering the dynamic equilibrium of forces on RR

wall system in horizontal and vertical directions and 1

n

i

i

T

can be expressed as;

1

tan tan cot tann

i v h a

i

T W Q t Q t cH c H

(1)

Total weight of the assumed failure wedge is 20.5 cotW H . The required strength of

reinforcement (K) can be expressed as;

2

2 21

0.5 tan cot * 1 *cot0.5 0.5

nv h

i f

i

Q t Q tK T H c a c

H H

(2)

2

1

2where, ; 0.5 and *

n

f a i

i

ca c c K T H c

H

Pseudo-dynamic inertia forces

The mass of small element of thickness dz at depth z is;

( )tan

H zm z dz

g

. At depth z and time

t, the seismic accelerations in horizontal and vertical direction can be written as;

, sin and , sinh h v v

s p

H z H za z t k g t a z t k g t

V V

(3)

The total inertia force in the horizontal direction, Qh(t) can be derived as;

37



0

( ) ,tan

H

hh h

kQ t m z a z t I

(4)

Assuming,

; ; ; ; 2 ; 2 and 2s p t

s p

H H t H t H tTV t TV t

V V T T T

2

2 cos sin sin4

I H t

(5)

On simplifying, 2 2 2

32 cos sin sin

4 tan

hh t

kQ t H H

H

(6)

Similarly, total inertia force in the vertical direction, Qv(t) can be stated as;

2 2 2

32 cos sin sin

4 tan

vv t

kQ t H H

H

(7)

On substituting Qh(t) and Qv(t) in equation 2;

2 2

3 3

2 2

3 3

tan 1 * 1 tan * 2 cos sin sin1 2

tantan2 cos sin sin

2

hf t

vt

kc a c H H H

HK

kH H H

H

(8)

Safety Factor, FS

On applying the load in the reinforced retaining wall system, the axial pullout of reinforcement

causes the shear resistance. The tension fully mobilized in the reinforcement layers over the

effective length. Hence, 1

n

i

i

t

can obtain by the following expression:

1 1

2 1 tann n

i v i i r

i i

t k H L

(9)

Using, 0.5 ; ; cot and 2 3i v v i r i cri rH i S S H n L L H H

In which, ϕr is the internal friction angle between soil and reinforcement. For ith

layer, Hi is

the depth of embedment, Li is the effective length and ti is the tensile force mobilized due to bond

resistance.

3

2

1

4tan 1 cot cot

6

n

i v r v r cri cri v

i

n nt S k n L H S

(10)

The safety factor is the ratio of the total mobilized bond resistance, to the maximum tensile

force generated in the reinforcement layers, which can be expressed as; 1 1

n n

i i

i i

FS t T

.


Figures (3-4) show the variation between required maximum strength of reinforcement, Kmax with kh

for different values of kv/kh for c=0 and 10kPa. It is clearly shown in figures (3-4) that with an increase in the

kh value, values of Kmax increases with considerable value. It can be also observed in figures (3-4) that, for

any value of kh, Kmax is more for the higher values of kv/kh. For example, for kh=0.2 and kv/kh=0.5 for c=0,

Kmax is 0.85. Kmax value is 0.59 for the c=10kPa for kh=0.2 and kv/kh=0.5. From this example, it can be

noticed the considerable effect of cohesion value of soil backfill.

Figures (5-6) show the variation between safety factor, FS with kh for different values of kv/kh for c=0

and 10kPa. It is clearly shown in figures (5-6) that with an increase in the kh value, values of SF decreases

continuously. From these figures it can be also noticed that the decrease in FS is very fast for the smaller

38



values of kh for both c=0 and 10kPa. For example, for kh=0.15 and kv/kh=0.5, FS is approximately 3.1 for c=0

and approximately 5.0 for c=10kPa.

Figure 3: Kmax with kh for different kv for c=0

Figure 4: Kmax with kh for different kv for c=10kPa

Figure 5: SF with kh for different kv for c=0

Figure 6: FS with kh for different kv for c=10kPa

Figure 7: with t for different ϕ for c=10kPa

Figure 8: (ƩTi)max with t for different ϕ for c=10kPa

Figure 9: tcritical with T for different H

39



Variation of inclination of failure wedge and maximum required strength of reinforcements for each

moment of lateral shaking has been shown in figures (7-8) for c=10kPa and T=0.3s and ϕ varies from 200 to

300. From the figures (7-8), it can be noticed that inclination of failure wedge increases and maximum

required strength of reinforcements decreases continuously with significant value, when value of ϕ increases

for each moment of lateral shaking.

The critical time (tcritical) for which required strength in the reinforcements would be maximum when the

period of lateral shaking is increasing. From the given formulation in this study, the calculated tcritical does not

vary with changes in the cohesion values, soil friction angle values and the seismic coefficients. But from

figure 9, it can be reported that tcritical value linearly increases with increase in the height of wall. The tcritical

value increases approximately 412%, 332% and 266% for height of wall 6m, 8m and 10m respectively when

T increases from 0.1s to 1.2s.

Validation of the present work

Kmax and critical inclination of the failure angle (cri) has been compared with Ghanbari and Ahmadabadi

(2010) as shown in Table 1. For a set of parameters (kv=0, kh=0.2, δ=0, c=0, ϕ=300, =20kN/m

3, T=0.2s,

Vs=150m/s and Vp=1.87Vs), K and cri show a good agreement.

Table 1 Comparison between results obtained from the present study by Ghanbari and Ahmadabadi (2010) for the case

of (kv=0, kh=0.2, δ=0, c=0, ϕ=300, =20kN/m

3, T=0.2s, Vs=150m/s and Vp=1.87Vs)

t = 0.0 t = 0.02 t = 0.04 t = 0.08 t = 0.10 t = 0.12

K cri K cri K cri K cri K cri K cri

H= 3 m Present Method 0.29 63.2 0.36 58.2 0.43 52.8 0.45 51.4 0.38 56.3 0.31 61.7

Ghanbari and

Ahmadabadi (2010)

0.29 62.5 0.36 58.2 0.43 54 0.45 53 0.38 57 0.31 61.8

4. CONCLUSIONS

In the available literature, very limited solutions to analyse the reinforced retaining walls under

seismic conditions are available considering c-ϕ soil backfills. The present formulation provide a solution in

simplest form to analyse the reinforced retaining walls for calculating the required strength of reinforcements

and inclination of failure wedge under seismic loading condition. The conclusions drawn from the present

study are as follows:

The required maximum strength of reinforcement Kmax increases with significant value on

increasing in kh values for higher values of kv/kh (for more than 0.5 value).

Safety factor decreases very fast for the smaller values of kh (less than or equal to 0.15) for both,

cohessionless and cohesive soil backfill.

On increasing the soil friction angle, angle of Inclination of failure wedge increases and the

maximum required strength of reinforcements decreases continuously in considerable amount

for each moment of lateral shaking.

The effect of cohesion value is very significant for the design of reinforced retaining wall, which

is clearly noticed in the present work.

On increasing the period of lateral shaking, tensile forces does not effected very much when the

height of wall increases.

On validating Kmax and critical obtained from the present work by Ghanbari and Ahmadabadi

(2010) is in a very good agreement.

REFERENCES

[1] A. Ghanbari and M. Ahmadabadi, New analytical procedure for seismic analysis of reinforced retaining wall with

cohesive-frictional backfill, Geosynthetics International. 17(6), (2010), 364-379.

[2] D. Choudhury, S.S. Nimbalkar and J.N. Mandal, External stability of reinforced soil walls under seismic conditions,

Geosynthetics International. 14(4), (2007), 211-218.

[3] G.V.N. Reddy, D. Choudhury, M.R. Madhav and S.E. Reddy, Pseudo-dynamic analysis of reinforced soil wall

subjected to oblique displacement, Geosynthetics International. 16(2), (2009), 61-70.

40



[4] R.S. Steedman and X. Zeng, The influence of phase on the calculation of pseudo-static earth pressure on a retaining

wall, Geotechnique, 40(1), (1990), 103-112.

[5] S.N.M. Tafreshi and M. Rahimi, A simplified pseudo-dynamic method of reinforced retaining-wall subjected to

seismic loads, In proceedings of 15th

World Conference on Earthquake Engineeing (15WCEE), Lisbon, Portugal,

Volume 27, September 24-28, 2012.

[6] S.S. Nimbalkar, D. Choudhury andJ.N. Mandal, Seismic stability of reinforced soil wall by pseudo-dynamic

method, Geosynthetics International. 13(3), (2006), 111-119.

41



Effect of Steep Slope on Single Model Pile subjected to

Lateral Load near Sloping Ground

Bhishm Singh Khati

1 and Vishwas A. Sawant

2

1Research Scholar, Department of Civil Engineering, IIT Roorkee. Roorkee 247 667, India

E-mail: [email protected] 2Associate Professor, Department of Civil Engineering, IIT Roorkee, Roorkee 247 667, India

E-mail: [email protected]

ABSTRACT

Most of structures in civil engineering are supported on pile foundation due to the insufficient bearing

capacity at the required depth. A detailed laboratory experimental model tests have been conducted to

investigate the single pile response for varying gradients of soil slopes. The study was carried out on

horizontal ground and to determine the effect of slope, the same is carried on slope of 1 Vertical to 1.28

Horizontal (1V:1.28H) with relative density of 25% and length to diameter ratios (L/D) of 15. The study

includes the effect of ground slope, crest distance, relative density and embedment length on lateral load

capacity. Based on the study, lateral soil resistance, bending moment and lateral deflection for different cases

are determined. From the study it is concluded that the lateral resistance against the lateral load increases

with increase in pile-soil relative stiffness. It is found that the soil resistance increases with increase in the

depth of the soil, relative density of the soil and the embedment length of the pile.

Key Words: Aluminum pile, Experimental model, Slope, Bending moment, Soil resistance

1. INTRODUCTION

Deep pile foundations are used in locations where the use of shallow foundations would lead to

unacceptably low factors of safety against shear failure or excessive settlement. In addition to axial loads, the

piles are often subjected to lateral loads and moments as well. Further, there are many circumstances in

which piles have to be provided on slopes. This in turn adversely reduces the lateral load capacity offered by

the soil and therefore, the capacity of foundation in the direction of the slope gets reduced drastically in

comparison to pile foundations that are located in horizontal ground surface. Depending upon the type of

structure and pile supports, there can be different causes of lateral forces. Wind are the most common cause

of lateral force that a pile has to support. The other major cause of lateral force is earthquake. The horizontal

shaking of the ground during earthquakes generates lateral forces that the piles have to withstand. In the case

of bridge abutments, horizontal forces are caused due to traffic and wind movement. Dam structure are

designed to withstand water pressures which transfer as horizontal forces on the supporting piles. In the case

of earth retaining structures, the primary role of piles is to resist lateral forces caused due to the lateral

pressures exerted by the soil mass behind the retaining wall. Piles are used to support open excavations, there

is no axial force and the only role of the piles is to resist lateral forces. So in many circumstances, the

external horizontal loads act at the pile head. The piles near sloping ground are subjected to lateral loads

which are more predominant than vertical loads. Limited experimental studies have been carried out to

analyze the behavior of single pile subjected to the lateral load near sloping ground. Very few studies have

considered the effect of distance between slope crest and pile. Initially tests were carried out on horizontal

ground condition. Then to determine the effect of slope, 1V:1.28H are used. In the parametric study, the

effect of distance from the crest of the slope on pile response to lateral load was examined. A comparison

was made with response in horizontal ground condition to highlight the effect of ground slope.

2. METHODOLOGY

Proposed work is aimed to be carried in a concrete tank of dimensions 2.50m×1.23m×1.12m. The experimental setup, i.e. the cable and pulley arrangement to provide lateral load on the embedded piles is shown in the figure 1. Apparatus of the test includes a loading frame connected by a pulley, LVDT, strain

42



gauges, data acquisition system and computer. The test is conducted on Solani river sand in dry condition at Roorkee. Parameters considered in the analysis for the experiments is shown in table 1.

Figure 1: Actual laboratory set up for sloping ground

Table 1: Parameters Considered in the Analysis

Parameter Values

Unit weight of soil γ (kN/m3) 14.12

Diameter of Pile (mm) 25

Length to diameter ratio (L/D) of pile 15

Relative density (%) 25

Slope angle θ (degrees) 38

Edge distance (s/D) 0 &10


Pile is instrumented with strain gauges for measurement of bending moment along the length of pile.

With the increase of lateral load, the average micro strain increases. For a particular load, the maximum

value of micro strain is obtained at the strain gauge located at a depth of 3L/8 or 4L/8 as shown in figure 2.

Figure 3 shows the typical lateral load displacement curve, with the increase of lateral load, the top lateral

displacement increases.

0

5

10

15

20

25

30

0 50 100 150 200

Late

ral

Load

(N)

Average Micro Strain

L/8

2L/8

3L/8

4L/8

5L/8

6L/8

Figure 2: Micro strain at different depth for slope 1V:1.28H & s/D=0

43



0

10

20

30

40

50

60

70

0 10 20 30 40 50

Lat

eral

Loa

d (N

)

Lateral Displacement (mm)

Figure 3: Lateral Load Vs Displacement for slope 1V:1.28H & s/D=10

The applied lateral load and corresponding lateral displacement at the ground level was measured and

recorded for each load increment. Figure 4 shows the typical lateral load displacement curves for various s/D

ratio for slope. From the figure, it is very clear that the change in ground surface from horizontal to slope the

lateral load capacity significantly reduced. Beyond 10D there will be very negligible change in lateral

displacement. It can be seen that for slope 1V:1.28H, load-displacement curve for the case of edge distance

s/D = 10, is very close to horizontal response.

This reduction is due to the reduction in passive resistance of the soil in front of the pile and also

reduced initial confining pressure. When the test was conducted for 1V:1.28H slope with relative density of

25%, the slope crest has started to yield when the lateral displacement of the pile exceeds 5 mm. This shows

the instability of steeper slope (1V:1.28H) in loose sand as a result pile experiences more lateral

displacement and bending moments. Generally, determining the ultimate load from lateral pile tests depends

on the tolerance of the structure supported by the piles. If no such criterion is available, the criterion usually

accepted for estimating the ultimate lateral load is the load corresponding to 20% of the pile diameter

(Narasimha Rao et al. 1998) lateral movements normal to the pile area. In the present study, the 20% of the

pile diameter is 5.0mm.

The lateral load capacity of pile in horizontal ground is always greater than piles in sloping ground.

Comparisons of pile response in horizontal ground condition with sloping ground for different positions of

pile are presented in figure 4. As expected piles in horizontal ground are much stiffer than those in sloping

ground. This may be attributed to lesser passive resistance available in sloping ground. As piles are moving

away from crest, pile top displacement is decreasing.

0

5

10

15

20

25

30

35

0 5 10 15

Late

ral

Load

(N)

Lateral Displacement (mm)

Horizontal

s/D=10

s/D=0

Figure 4: Effect of crest distance in slope with horizontal ground

Pile capacity is taken as load corresponding to displacement of 5mm (20% Diameter). For horizontal

ground condition, pile capacity is 30N. As expected, pile capacity is reducing with increase in ground slope.

44



Table 2 summarizes reduction in pile capacity with respect to horizontal ground condition. Maximum

reduction of the order 65% is observed and location of pile is at crest. For same slope, reduction decreases

with edge distance. For s/D = 10, reduction is only 20%.

Table 2: Percentage reduction in pile capacity with edge distance (s/D)

Slope Percentage (%) reduction in pile capacity

1V:1.28H s/D=0 s/D=10

65 20

The bending moment values at the location of strain gauges were calculated using strain measurement

(data logger). The strain response was found to be linear with the applied load. The distribution of bending

moment along the pile shaft were measured and plotted against the depth of pile. The variation of bending

moment (pattern) with depth is almost same in all the cases. It was found that the maximum bending moment

increases with increase in the applied lateral load (Figures 5-7). The maximum bending moment of the pile is

significantly decreases with increase in distance of pile from crest of slope as observed in fig.8. It is clear

that the depth of point of maximum bending moment moves downward as the pile moves towards crest of

slope. Effect of ground slope on bending moments is compared in fig. 8 with moments in level ground.

0

50

100

150

200

250

300

350

400

0 500 1000 1500 2000 2500 3000 3500 4000

Depth

(m

m)

BM (N-mm)

Load (N)

4.905

7.70085

10.4967

13.29255

16.0884

24.47595

30.06765

32.8635

35.65935

38.4552

41.25105

44.0469

46.84275

49.6386

Figure 5: Bending Moment Vs Depth for horizontal ground

0

50

100

150

200

250

300

350

400

0 500 1000 1500 2000 2500 3000

Depth

(m

m)

BM (N-mm)

Load (N)

4.905

7.70085

10.4967

16.0884

18.88425

24.47595

Figure 6: Bending Moment Vs Depth for slope & s/D=0

45



0

50

100

150

200

250

300

350

400

0 500 1000 1500 2000 2500 3000 3500 4000 4500 5000

Dep

th (

mm

)

BM (N-mm)

Load (N)

2.79585

5.5917

8.38755

13.97925

16.7751

22.3668

27.9585

30.75435

33.5502

38.4552

43.3602

53.1702

Figure 7: Bending Moment Vs Depth for slope & s/D=10

0

50

100

150

200

250

300

350

400

0 500 1000 1500 2000 2500 3000

Depth

(m

m)

BM (N-mm)

LL-25 NHorizontal

s/D=10

s/D=0

Figure 8: Bending Moment Vs Depth for slope and horizontal ground

Governing differential equation of pile can be written as; 0)(4

4

zpdz

ydEI . Assuming Winkler theory,

soil resistance p(z), is related to pile deflection y at given point. In the liner analysis, soil resistance p(z) = k

y; where k is modulus of subgrade reaction. But considering nonlinear behavior of soil, k is function of pile

deflection. Therefore, it is required to establish relation between soil resistance p(z), and pile deflection y

through model experiments.

The soil resistance along the pile shaft can be determined from the bending moment values using an

approach similar to that of Muthukumaran et al. (2008). In this approach, the distribution of the bending

moment curve along the pile shaft was fitted by a polynomial function as given in Eq. 1. Since the

experimental result shows more scatter, a minimum third order polynomial function has been chosen for the

curve fitting in order to reduce the error.

dczbzazzM 23 (1)

The soil resistance (force per unit length) was obtained by differentiating the bending moment profile M

(z) twice as given in Eq. 2.

bazdz

zMdzP 26

)()(

2

2

(2)

The displacement of the pile along its shaft was obtained by double integrating the bending moment

function as given in Eqs. 3 and 4.

dzzMEI

zY1

(3)

46



21

2345

261220

1)( CzC

dzczbzaz

EIzY (4)

In which, a, b, c and d are curve fitting constants. C1, C2 are constants of integration obtained from the

boundary conditions. Available bending moment values are fitted to third order polynomial equation with

respect to the depth for each load. This third order polynomial equation is differentiated to get shear force

equation of second order. Again this shear force polynomial equation is differentiated to get soil resistance

equation of first order. The equation of soil resistance is solved to get the values at the corresponding depth.

The polynomial equation for bending moment is integrated to get slope equation of fourth order with

an unknown constant C1. Similarly to get displacement equation, the slope equation was integrated with

respect to depth with constants C1 and C2. An integral constant C1 and C2 was eliminated using two boundary

conditions. First boundary condition is, top lateral displacement at ground level, (when z = 0, y = top lateral

displacement = C2). Second boundary condition is, displacement is zero at tip of pile. (z = L the depth of

embedment, y = 0).

4. CONCLUSIONS

The behavior of single model pile installed near a sloping ground subject to lateral load has been

investigated in the paper through an experimental study. Based on the results obtained from the experimental

study the following conclusions can be drawn; 1. A detailed parametric study is carried out to investigate the effect of edge distance from the slope of

the ground on the lateral pile capacity, displacement and the bending moment along the pile length. From the study, it has been observed that the effect of slope is insignificant when the pile is placed after 10D distance from the crest of slope.

2. The lateral load capacity is significantly increases with increase in distance of pile from crest of slope.

3. The lateral displacement is significantly reduced with increase in distance of pile from crest of slope. 4. The lateral load capacity is significantly reduced when the ground surface changes from horizontal

to slope. The change in ground surface from horizontal to 1V:1.28H slope, at the location of slope

crest, reduces the lateral load capacity to 65 %.

5. The maximum bending moment of the pile significantly increases with increase in ground slope.

However, bending moments significantly increases with decrease in distance of pile from crest of

slope.

REFERENCES

[1] D. Rathod, K. Muthukumaran and T.G. Sitharam, Response of laterally loaded pile in soft clay on sloping ground,

International Journal of Geotechnical Engineering, 10(1), (2015), 10-22.

[2] K. Muthukumaran, R. Sundaravadivelu and S.R. Gandhi, Effect of sloping ground on single pile load deflction

behaviour under lateral soil movement, In proceedings of 13th

World Conference on Earthquake Engineering,

Vancouver. B.C., Canada, August 1-6, 2004.

[3] K. Muthukumaran and N.A. Begum, Experimental investigation of single model pile subjected to lateral load in

sloping ground, International Journal of Geotechnical Engineering, 33, (2015), 935-946.

[4] S.R. Narsimha, V.G.S.T. Ramakrishna and M.R. Babu, Influence of rigidity on laterally loaded pile groups in

marine clay, Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 124(6), (1998), 542-549.

[5] N.A. Begum and K. Muthukumaran, Numerical modelling for laterally loaded piles on a sloping ground, In

proceedings of 12th

International Conference of International Association for Computer Methods and Advances in

Geomechanics, (IACMAG), Goa, India, October 1-6, 2008.

[6] V.A. Sawant and S.K. Shukla, Effect of Edge Distance from the Slope Crest on the Response of a Laterally Loaded

Pile in Sloping Ground, Geotechnical and Geological Engineering, 32(1), “(2014), 197-204.

47



Flexural behavior of Steel Fiber Reinforced

Concrete beam Bichitra Singh Negi

1 and Kranti Jain

2

1-M.Tech Student, Department of Civil Engineering, National Institute of Technology, Uttarakhand-246174,

Email: [email protected], Contact number: 09557817464.

2-Assistant Professor, Department of Civil Engineering, National Institute of Technology, Uttarakhand-246174,

Email:[email protected], Contact number: 09528685012

ABSTRACT` In recent time, Steel fiber reinforced concrete (SFRC) is a very advance material in respect of their

structural properties. SFRC has high ductility and this leads to increase the post cracking strength. Due to

these structural properties any one can use SFRC in hilly areas at earthquake prone region so that the more

ductile structure may design.

The main objective of this paper is to analyze the flexural behavior of SFRC based on investigation done

by many researchers in the past. Almost all fiber types used by the researchers enhance the flexural behavior

of concrete beam but some fiber enhanced behavior significantly as compared to other fiber depending on

the fiber geometry and fiber volume fraction. For this we compare flexural strength & toughness at limit of

proportionality (LOP) and Modulus of rupture (MOR) respectively and determine the maximum percentage

increment at those points.

By comparing all the results, we get max percentage increase in flexural strength and toughness for

twisted fiber with 1% & 1.5% macro and micro fiber volume fraction, respectively in concrete matrix. While

maximum value of flexural strength and toughness attain by hooked fiber with 1% and 1.5% macro & micro

fiber respectively. For concrete without micro fiber, matrix with 1.2% volume fraction of 30 mm high

strength twisted fiber show best result.

Key Words: Steel fiber reinforced concrete (SFRC), Fiber geometry, Fiber volume fraction,

Flexuralbehavior.

1. INTRODUCTION

Steel fiber has been widely used as a reinforcing material in concrete. Steel fiber reinforced concrete

now a day used in all type of structural work efficiently. As unreinforced concrete have low tensile strength

and toughness and reinforced concrete with reinforcement bar have tensile strength but are not economic.

With the incorporation of steel fiber in the concrete matrix, the cost of reinforcement as well as labour cost

can be reduce and flexural strength and toughness will increased.

Initially, the beam behaves elastically up to a point where first crack occur. After the first crack actual

function of steel fiber started as the fiber arrest the crack by obstructing the path of crack and hence acts as

crack arrestor. It means the introduction of steel fiber into concrete will enhance the post cracking property

of concrete depending on various factor like workability of matrix, fiber type, fiber geometry, fiber volume

fraction, fiber strength, fiber aspect ratio, fiber content, fiber orientation etc.

In the concrete matrix, fiber act as a multi-directional dispersed reinforcement. The main function of

fiber is to obstruct the cracks thereby prevent them from increasing by transferring the tension across the

crack. Crack in concrete can be generated due to various reasons like thermal crack and shrinkage crack.

Micro crack can also be generated due to initial stage loading which can be resist by fibers.

Reinforcing fiber can be of different type, shape and cross-section. Fiber may have straight end, hook

end and various other shapes with their varied length. Workability of steel fiber concrete can be improved by

using supplementary cementetious materials such as fly ash, silica fume and superplasticizer admixtures etc.

The production of ball effect reduces the mixing ability of different concrete constituent which cause a

reduction in workability and if matrix is not properly mixed resulting the reduction in strength. So

volume fraction should be in limit to achieve greater flexural strength and toughness.

48



2.ABBREVIATIONS USED

fLOP First cracking strength δLOP Deflection at limit of proportionality, mm

LOP Limit of proportionality δMOR Deflection at modulus of rupture, mm

MOR Modulus of Rupture PVA Polyvinyl-Alcohol Fiber

SFRC Steel fiber reinforced concrete LVDT Linear variable differential transformer

L Span length h Height of specimen

b Width of specimen

3. AIM AND OBJECTIVE

The main objective of this paper is to find out the efficacy of types of fiber by measuring flexural

strength & toughness of SFRC beams under four point bend test per ASTM C 1609. As role of fiber comes

into picture after post cracking stage of load deflection curve,so the measure variation due to fiber can be

analyzed at LOP and MOR.

4. EXPERIMENTAL PROGRAM

Flexural Behavior of SFRC beam

The flexural behavior of SFRC beam is classify into deflection hardening and deflection softening

behaviors, according to the change of load carrying capacity after first crack occur. Some parameters are

used for describing the flexural behavior of SFRC to compare the flexural performance of different fibers.

The first cracking point of SFRC is defined as Limit of Proportionality (LOP) (as shown in fig 1) according

to the ASTM C1018-97[5]; and, the maximum equivalent bending strength point of SFRC is defined as

Modulus of Rupture (MOR). The equivalent bending strength at MOR, fMOR, can be calculatedby using Eq.

(1) which was provided by ASTM C 1609/C1609M-05 [6]. And, the energy equivalent to the area under the

load–deflection curve up to MOR is notated as ToughMOR.

(a) (b)

Fig 1.(a) Four Point Bend Test, (b)Typical Load- Deflection Curve[1]

𝑓𝑀𝑂𝑅 = 𝑃𝑀𝑂𝑅

𝐿

𝑏ℎ2

(1)

49



where L is span length, b is the width of specimen, and, h is theheight of specimen. Testing of beam is

carried out in Four-point bend test as shown in Fig 1.

Types of Fiber Used

(a) (b) (c) (d)

Fig 2. (a) 60 mm hooked fiber,(b) 35 mm hooked fiber, (c) 60 mm crimped fiber, (d) 30 mm crimped fiber [4]

(a) (b) (c) (d)

Fig 3 . (a) Torex fiber,(b) High Strength hooked fiber, (c) Spectra fiber, (d) PVA fiber [1]

(a) (b) (c) (d) (e)

Fig 4 . (a) Smooth fiber, (b) Hooked fiber A, (c) Hooked fiber B, (d) Twisted fiber, (e) Micro-Smooth fiber [2]

50



Table 1.Fiber geometry and length Fiber

Code Geometry Cross-section

Length

(mm)

Diameter

(mm)

Aspect

Ratio

Beam Specimen Size

(mm×mm×mm)

F1 [1] Hooked-end Circular 60 0.8 75 100×100×350

F2 [1] Crimped Circular 60 1 60 100×100×350

F3 [1] Crimped Cresent 52 2.3×0.55 40 100×100×350

F4 [1] Twin-cone Circular 62 1 62 100×100×350

LS [2] Long Smooth Circular 30 0.3 100 100×100×350

HA [2] Hooked A Circular 30 0.375 80 100×100×350

HB [2] Hooked B Circular 62 0.775 80 100×100×350

T [2] Twisted Circular 30 0.3 100 100×100×350

T [3] High strength steel Torex (twisted) Circular 30 0.3 100 100×100×350

H [3] High strength steel hooked Circular 30 0.38 80 100×100×350

SP [3] polyethylene spectra Circular 38 0.38 100 100×100×350

PVA [3] PVA-fiber Circular 12 0.2 60 100×100×350

S [4] Hooked Deformed circular 35 0.55 65 180×180×600

W [4] Hooked Deformed circular 35 0.55 65 180×180×600

T [4] Hooked Deformed circular 35 0.55 65 180×180×600

X [4] Hooked Deformed circular 60 0.75 80 180×180×600

C [4] Hooked Deformed circular 60 0.75 80 180×180×600

Y [4] Hooked Deformed circular 60 0.75 80 180×180×600

D [4] Hooked Deformed circular 60 0.75 80 180×180×600

I [4] Crimped Deformed circular 30 0.6 50 180×180×600

CC [4] Crimped Deformed circular 30 0.6 50 180×180×600

J [4] Crimped Deformed circular 30 0.6 50 180×180×600

K [4] Crimped Deformed circular 60 0.7 85 180×180×600

DD [4] Crimped Deformed circular 60 0.7 85 180×180×600

L [4] Crimped Deformed circular 60 0.7 85 180×180×600

E [4] Hooked Deformed circular 60 0.75 80 180×180×600

F [4] Hooked Deformed circular 60 0.75 80 180×180×600

G [4] Crimped Deformed circular 30 0.6 50 180×180×600

H [4] Crimped Deformed circular 30 0.6 50 180×180×600

5. RESULT AND DISCUSSIONS

Table 2 shows torex fiber and hooked fiber with 1.2% volume fraction have higher increase in flexural

strength but considerable toughness of SPA fiber with 1.2% volume fraction show better result than other. At

0.40% volume fraction SPA04, T04, H04 fiber shows increased in flexural strength after first crack while

PVA04 fiber show decreased flexural strength as compared to its strength at first crack. While considering

toughness the SPA12 fiber have far better increment in each volume fraction. Table 3 shows that specimen

contain two type of fiber. As we try to hybrid our concrete matrix with macro steel fiber in combination with

micro fibers, the post cracking behavior of concrete is enhanced. Fig 6 suggest that the flexural strength and

toughness of specimen with long smooth fiber and twisted fiber enhanced more for same proportion of

volume fraction as compared to hooked fiber.

51



Table 2. % Variation in Flexural Strength and Toughness for beam [1]

Volume

Fraction Type

f (LOP)

MPa

f (MOR)

MPa

% Increase in

Flexural

Toughness(LOP)

N-m

Toughness(MOR)

N-m

% Increase in

Toughness

1.20%

T12 2.62 13.08 399.24 0.116 44.117 37931.90

H12 2.6 11.59 345.77 0.112 28.453 25304.46

SP12 2.76 10.03 263.41 0.078 90.679 116155.13

PVA12 3.13 4.72 50.80 147 7.384 -94.98

0.40%

T04 2.28 7.61 233.77 0.129 21.399 16488.37

H04 2.56 6.97 172.27 0.128 22.922 17807.81

SP04 2.24 7.89 252.23 0.077 35.239 45664.94

PVA04 2.74 1.73 -36.86 0.123 2.086 1595.93

Fig 5. Percentage Variation at LOP with respect to MOR (a) Flexural, (b) Toughness

Table 3. % Variation in Flexural Strength and Toughness for beam tested [2] Volume

Fraction Type

f (LOP)

MPa

f (MOR)

MPa

% Increase in

Flexural

Toughness(LOP)

N-m

Toughness(MOR)

N-m

% Increase in

Toughness

0 LS10SS00 16.12 22.45 39.27 2.09 85.856 4007.94

0.5 LS10SS05 14.31 33.8 136.20 1.651 122.443 7316.29

1 LS10SS10 14.68 37.51 155.52 1.673 165.808 9810.82

1.5 LS10SS15 13.53 39.95 195.27 1.46 158.026 10723.70

0 HA10SS00 16.67 26.9 61.37 2.326 90.5 3790.80

0.5 HA10SS05 17.35 31.21 79.88 2.562 117.79 4497.58

1 HA10SS10 16.66 38.26 129.65 2.271 149.586 6486.79

1.5 HA10SS15 17.33 34.35 98.21 2.402 111.674 4549.21

0 HB10SS00 14.63 24.36 66.51 1.519 87.98 5691.97

0.5 HB10SS05 16.09 29.36 82.47 2.119 111.391 5156.77

1 HB10SS10 14.78 35.53 140.39 1.797 168.97 9302.89

1.5 HB10SS15 16.79 47.25 181.42 2.279 226.255 9827.82

0 T10SS00 13.18 28.34 115.02 2.18 90.931 4071.15

0.5 T10SS05 13.78 32.75 137.66 1.55 119.281 7595.55

1 T10SS10 14.25 41.2 189.12 1.625 156.358 9522.03

1.5 T10SS15 14.59 47.22 223.65 1.644 189.588 11432.12


-100

0

100

200

300

400

500

T12

H12

SP

12

PV

A1

2

T04

H04

SP

04

PV

A0

4

% Increase in Flexural

% Increase inFlexural

-50000

0

50000

100000

150000

T12

H12

SP

12

PV

A1

2

T04

H04

SP

04

PV

A0

4

% Increase in Toughness

% Increase inToughness

52



Fig 7 show a little variation in flexural strength and toughness but some of the bar show negative trend it

mean the specimen show deflection softening behavior and contribution of fiber does not enhanced the

flexural behavior of specimen. High strength specimen containing F1 fiber show contradictory effect in

flexural strength characteristics and toughness characteristics. Its post cracking flexural strength decreases

after LOP while toughness increases.

Table 5 and Table 6 shows medium strength and high strength beam specimen details respectively.S(M-

HO-35-0.75) shows medium strength specimen having 35 mm hooked fiber with volume fraction 0.75%. It

can be easily seen from table 5 and Fig 8 that the 30mm crimped fiber with volume fraction 0.75% and 1.0

% show deflection softening behavior (flexural strength decreases from LOP to MOR), while toughness will

increases for all specimen. Table 6 and Fig 9 again show same behavior for crimped fiber of 30 mm size at

volume fraction 0.75. Table 4. % Variation in Flexural Strength and Toughness for beam [3]

Volume

Fraction Beam ID

f(LOP)

MPa

f(MOR)

MPa

% Increase in

Flexural

Toughness(LOP)

N-m

Toughness(MOR)

N-m

% Increase in

Toughness

Normal

Strength

(40kg/m3)

F1 I 6.49 6.76 4.16 5.67 5.7 0.53

F2 I 5.57 5.61 0.72 3.61 3.62 0.28

F3 I 5.9 5.95 0.85 2.8 2.71 -3.21

F4 I 5.9 5.07 -14.07 4.91 4.69 -4.48

Mid Strength

(40kg/m3)

F1 II 6.88 7 1.74 5.11 5.12 0.20

F2 II 6.67 6.71 0.60 3.22 3.21 -0.31

F3 II 6.84 6.92 1.17 2.64 2.6 -1.52

F4 II 6.93 7.01 1.15 4.00 4.01 0.25

High Strength

(40kg/m3)

F1 III 10.49 10.28 -2.00 3.54 4.78 35.03

F2 III 9.98 9.98 0.00 3.06 3.01 -1.63

F3 III 9.55 9.17 -3.98 2.54 2.49 -1.97

F4 III 9.39 9.42 0.32 4.98 4.94 -0.80


050

100150200250

LS1

0SS

00

LS1

0SS

10

HA

10

SS0

0

HA

10

SS1

0

HB

10

SS0

0

HB

10

SS1

0

T10

SS0

0

T10

SS1

0


% Increase

in Flexural0

2000400060008000

100001200014000


% Increase

inToughness

-20

-10

0

10

F1I

F2I

F3I

F4I

F1II

F2II

F3II

F4II

F1III

F2III

F3III

F4III


-20

0

20

40

F1I

F2I

F3I

F4I

F1II

F2II

F3II

F4II

F1III

F2III

F3III

F4III


53



Table 5. % Variation in Flexural Strength and Toughness [4]

Volume

fraction Beam ID

f(LOP)

MPa

f(MOR)

MPa

% Increase in

Flexural

Toughness(LOP)

N-m

Toughness(MOR)

N-m

% Increase in

Toughness

0.75 S (M-HO-35-0.75) 5.44 6.12 12.50 2.5 17.86 614.40

1 W(M-HO-35-1.0) 6.06 7.36 21.45 3.105 29.11 837.52

1.5 T (M-HO-35-1.50) 6.47 8.6 32.92 3.31 41.07 1140.79

0.5 X(M-HO-60-0.50) 5.37 6.06 12.85 2.95 74.6 2428.81

0.75 C(M-HO-60-0.75) 6.06 6.51 7.43 3.23 78.63 2334.37

1 Y(M-HO-60-1.00) 6.81 9.49 39.35 4.31 122.25 2736.43

1.5 D(M-HO-60-1.50) 6.41 12.63 97.04 1.74 164.42 9349.43

0.75 I (M-CR-30-0.75)

5.48 3.36 -38.69 2.5 39.67 1486.80

1 CC(M-CR-30-1.0) 5.77 4.86 -15.77 2.63 28.56 985.93

1.5 J (M-CR-30-1.50) 6.05 6.36 5.12 2.76 17.45 532.25

0.75 K (M-CR-60-0.75) 4.75 5.04 6.11 1.94 25.34 1206.19

1 DD(M-CR-60-1.0) 5.14 5.86 14.01 2.12 49.76 2247.17

1.5 L (M-CR-60-1.50) 5.53 6.68 20.80 2.31 74.18 3111.26


-60

-40

-20

0

20

40

60

80

100

120

S (

M-H

O-3

5-0

.75

)

W (

M-H

O-3

5-1

.00

)

T (

M-H

O-3

5-1

.50

)

X (

M-H

O-6

0-0

.50)

C (

M-H

O-6

0-0

.75

)

Y (

M-H

O-6

0-1

.00)

D (

M-H

O-6

0-1

.50)

I (M

-CR

-30-0

.75

)

CC

(M

-CR

-30-1

.0)

J (M

-CR

-30

-1.5

0)

K (

M-C

R-6

0-0

.75

)

DD

(M

-CR

-60

-1.0

)

L (

M-C

R-6

0-1

.50)


0

2000

4000

6000

8000

10000

S (M

-HO

-35

-0.7

5)

W (

M-H

O-3

5-…

T (M

-HO

-35

-1.5

0)

X (

M-H

O-6

0-0

.50

)

C (

M-H

O-6

0-0

.75

)

Y (M

-HO

-60

-1.0

0)

D (

M-H

O-6

0-…

I (M

-CR

-30

-0.7

5)

CC

(M

-CR

-30

-1.0

)

J (M

-CR

-30

-1.5

0)

K (

M-C

R-6

0-0

.75

)

DD

(M

-CR

-60

-1.0

)

L (M

-CR

-60

-1.5

0)


54



Table 6 % Variation in Flexural Strength and Toughness [4]

Volume

fraction Beam ID

f(LOP)

MPa

f(MOR)

MPa

% Increase in

Flexural

Toughness

(LOP) N-m

Toughness(MOR)

N-m

% Increase in

Toughness

0.75 E(H-HO-60-0.75) 6.84 6.36 -7.0175 3.52 32.04 810.23

1.5 F (H-HO-60-1.50) 11.01 15.07 36.8756 29.7 202.52 581.89

0.75 G (H-CR-30-0.75) 6.6 2.92 -55.7576 3.64 29.43 708.52

1.5 H (H-CR-30-1.50) 7.67 7.3 -4.8240 2.17 6.15 183.41


6. CONCLUSION

As volume fraction of fiber increases in the concrete matrix, flexural strength and toughness also

increases.It is applicable up to 1 % volume fraction after that the effectiveness of fiber decreases

because of fiber balling effect which affects workability of concrete matrix.

Hooked steel fiber shows more increases in flexural strength and toughness as compared to crimped

fiber.

If concrete matrix contain micro fiber along with macro fiber then its flexural behavior enhanced

more as compared to matrix with only macro fiber.

Fiber length also affect flexural strength and toughness, as length of fiber increases its flexural

strength and toughness also increases but up to a certain limit as above a limit length of fiber is

increased than it will adversely affect its flexural strength.

Among all the above fiber used, matrix with twisted fiber has more enhanced flexural behavior.

REFERENCES

[1] D. joo Kim, A. E. Naaman, and S. El-Tawil, “Comparative flexural behavior of four fiber reinforced

cementitious composites,” Cem. Concr. Compos., vol. 30, no. 10, pp. 917–928, 2008.

[2] D. J. Kim, S. H. Park, G. S. Ryu, and K. T. Koh, “Comparative flexural behavior of Hybrid Ultra High

Performance Fiber Reinforced Concrete with different macro fibers,” Constr. Build. Mater., vol. 25, no.

11, pp. 4144–4155, 2011.

[3] N. Banthia and Trottier J.F., “Concrete reinforced with deformed steel fibers 2. Toughness

characterization,” ACI Mater. J., vol. 92, no. 2, pp. 146–154, 1995.

[4] Jain Kranti Gyanchand, “Shear Behaviour of Steel Fibrous concrete beams without stirrup

reinforcement,Ph.D Thesis, Department of Civil Engineering” IIT Roorkee, India.

[5] ASTM C 1018-97. Structural test method for flexural toughness and first crack strength of fiber

reinforced concrete (using beam with third point loading). American Society of Testing and Materials;

October 1998. p. 514–51.

[6] ASTM C 1609/C 1609M-05. Structural Test method for flexural performance of fiber reinforced concrete

(using beam with third point loading). American Society of Testing and Materials; January 2006. p.

1–8.

0200400600800

1000


-80

-60

-40

-20

0

20

40

60


55



A new Energy Based Seismic Design method for steel moment

resisting multi-story multi-bays Ankush Kumar

1 and Shashi Narayan

1

1Department of Civil Engineering, National Institute of Technology, Uttarakhand

ABSTRACT

Structural engineers mainly aim at précising the complex dynamic effect of seismically induced forces

in the form of lateral loads. The new energy based seismic design (EBSD) of multi-story multi-bay Moment

resisting frame, utilizes the plastic behavior of the steel and the energy dissipation by the moment-curvature

hysteresis. Bilinear Plastic (BP) Model for Hysteresis is assumed for the steel. The energy demand during an

earthquake can be predicted and that the energy supply of the structural system can be established. In a

satisfactory design, the energy supply is more than the energy demand. In the present study, a six-story three

bays steel frame is designed using the EBSD method proposed

Key Words: Energy-Based Seismic Design, Bilinear plastic Model, Input Energy Distribution and

Plastic Analysis.

1. INTRODUCTION

The main aim of earthquake resistant design is to convert the complex dynamic effect of seismically

induced ground displacement in the form of lateral loads or displacement. For the past four decades, many

researchers have proposed different procedures for earthquake resistant design of structures. This continuous

effort has resulted in several revisions of Indian Standard code of practice “Criteria for Earthquake Resistant

Design of Structure” by the Bureau of Indian Standard (BIS), New Delhi[1].

An earthquake resistant design of a structure is not only based on the peak response demand

(displacement or force) but also transient response demand. In order to use the full capacity of the structure

during an earthquake, the structure may go in the inelastic zone or plastic zone. The nonlinear response, like

plastic rotation, is a measure of nonlinear performance of the structure during the earthquake. The plastic

energy, product of plastic moment and rotation, is an indication of the total damage to the structure. The

ability of the structure to absorb and dissipate the energy due to an earthquake governs the energy based

earthquake-resistant design objective. There are various approaches to design the structure based on energy

demand and capacity. Decanini and Mollaioli[2] evaluated the hysteretic energy demand of the structure

based on the intensity and spectral distribution of the hysteretic energy to input energy ratio. The ratio is

influenced by damping, ductility ratio, soil class and hysteretic model. Terapathana [3] computed seismic

energy demand of three-story reinforced concrete frame subjects to 20 LA10/50 earthquake records of SAC

project. The demand is given by their mean and mean plus standard deviation of energy demand for each

story level. Using modal characteristics of the MDOF system, Mezgebo and Lui[4], developed energy

demand along with seismicity of the site.

Gaetano Manfredi[5] developed a procedure to obtain input energy spectra in which damage potential

index capable of taking into account the effect of the duration of the ground motions. The input energy and

hysteretic energy for MDOF system are approximately determined from the equivalent SDOF system. An

empirical formula to determine the absorbed energy in MDOF systems using the energy spectra for SDOF

system is given by Chou and Uang [6]. Wang and Yi [7] derived the relation between the hysteretic seismic

energy of multi-story buildings and equivalent SDOF system.

The story-wise distribution of hysteretic energy over the height of a MDOF system has been

experimentally calculated by Senviratna and Krawinkler[8]. The distribution of hysteretic energy along the

height of the frame is linear, for regular frames which have a damping ratio of ξ = 0.02 [9]. The inelastic

hysteretic energy can be evaluated by total work done by force distribution height to cause the displacement

of each story using pushover analysis [4,6,7].

Energy-based seismic design (EBSD) method was first proposed by G.W.Housner[10], which is based

on the assumption that the energy demand during an earthquake can be predicted and the energy capacity of

the structural element can be established. In the satisfactory design, the energy capacity is more than energy

56



demand. Several authors [3,4,11] have developed earthquake resistant design procedure for a multistory-

single bay frame for Reinforced Concrete and steel structures. whereas, the design procedure for the multi-

story-multi-bay frame is in a primitive state.

In this study, a methodology is presented for earthquake resistant design of multi-story-multi-bay

moment resisting frame using energy principle. A six story-three bays moment resisting steel frame is

designed as an example problem using the purposed method. The proposed method is based on energy

demand of a ground motion and energy capacity of the structure. The energy capacity of the structure is

determined by plastic analysis assuming bilinear plastic model. The elastic energy capacity of the structure is

minimal compared to the plastic energy. Therefore, not considered in the energy capacity of the structure.

2. ENERGY DEMAND OF AN EARTHQUAKE

A viscously damped SDOF system having mass, m, stiffness, k ,and damping, c, as shown in figure

1a subjected to the earthquake is considered to evaluate energy demand of an earthquake. The dynamic

equation can be written as

𝑚𝑢�̈� + 𝑐𝑢 ̇ + 𝑘𝑢 = 0 (1)

where, 𝑢𝑡 = absolute displacement of the mass; 𝑢 = relative displacement of the mass w.r.t. the

ground; 𝑢𝑔= ground absolute displacement, 𝑢𝑡 , can be expressed in terms of 𝑢 and 𝑢𝑔and is given as

𝑢𝑡 = 𝑢 + 𝑢𝑔 (2)

Double differentiation of equation 2 with respect to .time, t, & using in equation 1, we get

𝑚�̈� + 𝑐𝑢 ̇ + 𝑘𝑢 = −m𝑢�̈� (3)

Hence, the SDOF system in figure 1(a) can be conveniently treated as the equivalent SDOF system in figure

1(b) with a fixed base and subject to an effective horizontal dynamic force.

For MDOF system which is subjected to earthquake ground motion, equation 3 can be written as

𝑀�̈�(𝑡) + 𝐶�̇�(𝑡) + 𝐾𝑢(𝑡) = 𝑀𝑢�̈�(𝑡) (4)

where, 𝑀, 𝐶 𝑎𝑛𝑑 𝐾 are the mass, damping and stiffness matrices of size (𝑛 × 𝑛), respectively

�̈�(𝑡), �̇�(𝑡) and 𝑢(𝑡) are relative acceleration, velocity and displacement vectors of 𝑛 order respectively.

Let, 𝑢(𝑡) = 𝜱𝑥(𝑡) = ∑ ∅𝑖𝑛𝑖=1 𝑥𝑖(𝑡) where, 𝜱 is the 𝑛 × 𝑛 mode shape matrix composed of 𝑛 mode

shape vector ∅𝑖 each of dimension 𝑛 × 1 and 𝑥𝑖(𝑡), 𝑖 = 1,2, … . 𝑛.Then equation 4 can be written as

𝑀 ∑ ∅𝑖𝑛𝑖=1 �̈�𝑖(𝑡) + 𝐶 ∑ ∅𝑖

𝑛𝑖=1 �̇�𝑖(𝑡) + 𝐾 ∑ ∅𝑖

𝑛𝑖=1 𝑥𝑖(𝑡) = −𝑀𝑢�̈�(𝑡) (5)

For 𝑟𝑡ℎ mode, multiply ∅𝑟𝑇both side to above equation, we get

∅𝑟𝑇𝑀 ∑ ∅𝑖

𝑛𝑖=1 �̈�𝑖(𝑡) + ∅𝑟

𝑇𝐶 ∑ ∅𝑖𝑛𝑖=1 �̇�𝑖(𝑡) + ∅𝑟

𝑇𝐾 ∑ ∅𝑖𝑛𝑖=1 𝑥𝑖(𝑡) = −∅𝑟

𝑇𝑀𝑢�̈�(𝑡) (6)

(a) Moving base system 𝑢�̈�

𝑐

𝑘

𝑢𝑔(𝑡) 𝑢𝑡(𝑡)

𝑚 𝑚

(b) Equivalent fixed base system

−𝑚�̈�𝑔(𝑡)

𝑢�̈�

𝑐

𝑘 𝑢𝑡(𝑡)

𝑚 𝑚

Figure 1 Mathematical model of a SDOF system subjected to an Earthquake Ground Motion

57



In equation 6, For linear system, ∅𝑟𝑇𝑀∅𝑖 = 0 and ∅𝑟

𝑇𝐾∅𝑖 = 0, if 𝑟 ≠ 𝑖 and 𝐶 can be chosen such that

∅𝑟𝑇𝐶∅𝑖 = 0 for 𝑟 ≠ 𝑖. when 𝑟 = 𝑖 the equation 6 reduces to SDOF as given in equation 7. The response of

a MDOF structure can be evaluated by weighted sum of response of n-SDOF structure during different

modes. It can also be concluded that different modes are orthogonal and uncoupled with each other whereas

for nonlinear they are coupled.

∅𝑟𝑇𝑀∅𝑟�̈�𝑟(𝑡) + ∅𝑟

𝑇𝐶∅𝑟�̇�𝑟(𝑡) + ∅𝑟𝑇𝐾∅𝑟𝑥𝑟(𝑡) = −∅𝑟

𝑇𝑀𝑢�̈�(𝑡) (7)

Assuming, ∅𝑟𝑇𝑀∅𝑟 = 𝑀𝑟 and ∅𝑟

𝑇𝐶∅𝑟 = 𝐶𝑟, then equation 7 can be written as

𝑀𝑟�̈�𝑟(𝑡) + 𝐶𝑟�̇�𝑟(𝑡) + ∅𝑟𝑇𝐾∅𝑟𝑥𝑟(𝑡) = −∅𝑟

𝑇𝑀𝑢�̈�(𝑡) (8)

Multiplying equation 4 by �̇�(𝑡) and integrating w.r.t. time varies from 𝑡 = 0 to 𝑡 which represent energy

balanced equation which is written as

∫ 𝑀𝑡

0�̈�(𝑡)�̇�(𝑡)𝑑𝜏 + ∫ 𝐶

𝑡

0�̇�(𝑡)�̇�(𝑡)𝑑𝜏 + ∫ 𝐾

𝑡

0𝑢(𝑡)�̇�(𝑡)𝑑𝜏 = ∫ −𝑀

𝑡

0𝑢�̈�(𝑡)�̇�(𝑡)𝑑𝜏 (9)

Right-hand side of term equation 9 represent input energy, 𝐼𝐸 = ∫ −𝑀𝑡

0𝑢�̈�(𝑡)�̇�(𝑡)𝑑𝜏, and in left hand side

of the above equation, first term is relative kinetic energy, 𝐸𝑘 = ∫ 𝑀𝑡

0�̈�(𝑡)�̇�(𝑡)𝑑𝜏, second term is damping

energy, 𝐸𝜉 = ∫ 𝐶𝑡

0�̇�(𝑡)�̇�(𝑡)𝑑𝜏 and third term is absorbed energy, 𝐸𝑎 = ∫ 𝐾

𝑡

0𝑢(𝑡)�̇�(𝑡)𝑑𝜏 and equation

can be written as

𝐸𝑘 + 𝐸𝜉 + 𝐸𝑎 = 𝐼𝐸 (10)

In similar way, equation 8 can be written as equation 11, which represent energy content in 𝑟𝑡ℎ mode

∫ 𝑀𝑟�̈�𝑟(𝑡)

𝑡

0

�̇�𝑟(𝑡)𝑑𝜏 + ∫ 𝐶

𝑡

0

�̇�𝑟(𝑡)�̇�𝑟(𝑡)𝑑𝜏 + ∫ ∅𝑟𝑇𝐾∅𝑟

𝑡

0

𝑥𝑟(𝑡)�̇�𝑟(𝑡)𝑑𝜏 = ∫ −∅𝑟𝑇𝑀

𝑡

0

𝑢�̈�(𝑡)�̇�𝑟(𝑡)𝑑𝜏

Dividing both sides with 𝑀𝑟, we get

∫ �̈�𝑟(𝑡)𝑡

0�̇�𝑟(𝑡)𝑑𝜏 + ∫ 2𝜉𝑟𝜔𝑟�̇�𝑟(𝑡)

𝑡

0�̇�𝑟(𝑡)𝑑𝜏 +

∫ ∅𝑟𝑇𝐾∅𝑟

𝑡0 𝑥𝑟(𝑡)�̇�𝑟(𝑡)𝑑𝜏

𝑀𝑟=

∫ −∅𝑟𝑇𝑀

𝑡0 𝑢�̈�(𝑡)�̇�𝑟(𝑡)𝑑𝜏

𝑀𝑟 (11)

Left-hand side of equation 11 that is ∫ −∅𝑟

𝑇𝑀𝑡

0 𝑢�̈�(𝑡)�̇�𝑟(𝑡)𝑑𝜏

𝑀𝑟 is input energy per unit mass for 𝑟𝑡ℎ mode and is

denoted by (𝐼𝐸

𝑚)

𝑟. Total energy can be evaluated by weighted sum of energy content by different modes and

𝐼𝐸𝑇𝑂𝑇𝐴𝐿 can be written as equation 12.

𝐼𝐸𝑇𝑂𝑇𝐴𝐿 = 𝑀1 × (𝐼𝐸

𝑚)

1+ 𝑀2 × (

𝐼𝐸

𝑚)

2+. . . 𝑀𝑝 × (

𝐼𝐸

𝑚)

𝑝 (12)

where p is a number of modes required for total energy. Ideally, it should be equal to degrees-of-

freedom. As the response of the structure is governed by first few modes, in a similar way, energy content

can be estimated by first few modes. The modes considered in this is the modes for which total modal mass

participating factor is 90%, i.e., the energy content of these modes are cumulating to 90% of total energy. In

a similar way, other energy terms of equation 10 can be evaluated for MDOF system by a weighted sum of

energy per unit mass. The input energy can be normalized on the basis of different hysteretic models used for

dissipation of energy and different soil site condition, Mezgebo and Lui[12] formulate empirical formula as a

function for Normalized Input energy(𝑁𝐸) which depend on the ductility of the structures given as

𝑁𝐸 = √𝐼𝐸/𝑚

𝑉𝐼= √

𝐼𝐸/𝑚

𝐶𝐴𝑉 𝑋 𝑃𝐺𝑉 (13)

where 𝑉𝐼 is velocity index; 𝐶𝐴𝑉is absolute cumulative velocity; 𝑃𝐺𝑉 is peak ground velocity of design

earthquake;

Input Energy is distributed over height by using pushover analysis considering a total deformation is 4%

of total height [1].Total deformation is chosen as 4% based on the performance-based design of structure and

comfort level of human. Storey shear and deformation at each story of the structure is obtained from

pushover analysis, and the product of both gives the energy demand for the story. Total external work done

during pushover analysis for the MRF (figure 2a), is given by 𝑊𝐸𝑡𝑜𝑡𝑎𝑙 = ∑ 𝐹𝑖𝑑𝑖𝑛𝑖=1 , where 𝑑𝑖is

displacement at 𝑖𝑡ℎstorey and 𝐹𝑖is storey shear at 𝑖𝑡ℎstorey and ∆𝑖= 𝑑𝑖 − 𝑑𝑖−1. External work done for the

1st storey displacement can be estimated assuming the structure deformation as shown in figure 2b and is

given by 𝑊𝐸𝑖 = (∑ 𝐹𝑖𝑁𝑖=1 )∆1 In similar way external work done for k

th storey is given by 𝑊𝐸𝑖 =

(∑ 𝐹𝑖𝑁𝑖=𝑘 )∆𝑘 (figure 2c). These total work done is equal to smmation of work done for all storey. Therefore,

58



the energy capacity of structure is distributed over height as the work done for each story and distribution

ratio for ith story (𝑅𝑖) is given as

𝑅𝑖 =𝑊𝐸𝑖

𝑊𝐸𝑡𝑜𝑡𝑎𝑙 (14)

Figure 2. Energy Distribution Mechanisms

3. ENERGY CAPACITY OF THE STRUCTURE

The Input energy of the earthquake is dissipated by the structure in the form of damping by hysteretic

energy. The instantaneous increase in energy is converted into kinetic energy which in turn is converted to

strain energy and hysteretic energy over cycles of deformation. Strain energy as compared to hysteretic

energy is very less. Therefore, only hysteretic energy is used for evaluating the Energy capacity of the

structure. The hysteretic energy is evaluated using plastic analysis and lumped plasticity model.

The plastic energy is evaluated using plastic analysis modes of failure of the multi-story-multi-bays

frame using mechanisms method. Then there are four possible types of collapse mechanisms for multi-story-

multi-bays frame these are (a) beam mechanisms (b). sway mechanisms (c). combine sway mechanisms (d).

combine sway and beam mechanisms. Plastic analysis of two story two-bay frame of different mechanisms is

shown in figure 3. For, a number of bays be N and number of the story be k, and assuming that the end

moment and mid-span moment carrying capacity of the beam is same. The internal work done for different

possible mechanisms are given as: For Beam mechanisms

𝑁 × (2𝑀𝑃,𝐵𝑃 × 𝜃𝑃 + 𝑀𝑃,𝐵

𝑃 × 2𝜃𝑃) = 𝐼𝑛𝑡𝑒𝑟𝑛𝑎𝑙 𝑊𝑜𝑟𝑘 𝐷𝑜𝑛𝑒 (15)

For sway mechanisms

(𝑁 − 1)(2 × 𝑀𝑘,𝐶𝐼𝑃 × 𝜃𝑃) + (2)(2 × 𝑀𝑘,𝐶𝐸

𝑃 × 𝜃𝑃) = 𝐼𝑛𝑡𝑒𝑟𝑛𝑎𝑙 𝑊𝑜𝑟𝑘 𝐷𝑜𝑛𝑒 (16)

For combine sway mechanisms

(𝑁 − 1)(𝑀𝑘,𝐶𝐼𝑃 × 𝜃𝑃) + (2)(𝑀𝑘,𝐶𝐸

𝑃 × 𝜃𝑃) + (𝑁)(2 × 𝑀𝑘,𝐵𝑃 × 𝜃𝑃) = 𝐼𝑛𝑡𝑒𝑟𝑛𝑎𝑙 𝑊𝑜𝑟𝑘 𝐷𝑜𝑛𝑒 (17)

For combine sway and beam mechanisms

(𝑁 − 1)(𝑀𝑘,𝐶𝐼𝑃 × 𝜃𝑃) + (2)(𝑀𝑘,𝐶𝐸

𝑃 × 𝜃𝑃) + 𝑁(𝑀𝑘,𝐵𝑃 × (2𝜃𝑃 + 𝜃𝑃) = 𝐼𝑛𝑡𝑒𝑟𝑛𝑎𝑙 𝑊𝑜𝑟𝑘 𝐷𝑜𝑛𝑒 (18)

where, 𝑀𝑘,𝐶𝐼𝑃 = plastic moment capacity for 𝑘 story interior column; 𝑀𝑘,𝐶𝐸

𝑃 = plastic moment capacity for 𝑘

story exterior column; 𝑀𝑘,𝐵𝑃 = plastic moment capacity for 𝑘 story beam; 𝜃𝑃 =plastic rotation

59



Internal work done is the work done for static analysis. In order to take into count dynamic effects, the above-mentioned equation is multiplied by a factor four. This factor can be attributed to the ratio

of a factor for a dynamic to monotonic hysteretic energy ATC 40[14]. It can be explained by the energy of

one cycle of an idealized hysteresis (Bilinear Plastic Model as used in this study) is four times than that of

monotonic loading. The dynamic energy for the system is the area under the hysteresis loop, and the energy

due to monotonic loading is the shaded area as shown in figure 4. It can be seen that with stable hysteresis

behavior free from any stiffness or strength degradation, four times the monotonic area is equal to the full

cycle area. FEMA 355F[15] recommended plastic rotation of 0.025 to 0.030 for steel moment connection.

Therefore, in this study a design plastic rotation (𝜃𝑃) of 0.030 is used in equation 15-18 and plastic moment

carrying capacity can be calculated once the external energy demand is known. The energy capacity of

structure is 4times internal work done (equations 15-18) as explained.

Figure 4 Bilinear Hysteretic model

(b) sway mechanisms

(c) combine sway mechanisms (d) combine sway and beam mechanisms

(a) beam mechanisms

Figure 3 Plastic mechanisms of two story two-bay frame

60



4. OPTIMIZATION OF MEMBER SIZES

The energy constrained for each story and total energy pose an optimization problem for efficient design.

The story-wise optimization is involved in which formulating a series of story-wise collapse mechanisms and

solving an optimization problem with weight as the objective function and constrained as a mention. The

optimization relates the story level hysteretic energy demand to the collapse mechanism based on internal

work done. The optimization of members is done by top-down approach i:e; optimization roof level

than intermediate levels and lastly ground level. The weight function can be expressed in terms of

the length of the member and their plastic moment capacities, which can be written as 𝑀𝑖𝑛𝑖𝑚𝑖𝑧𝑒, 𝑊 = (𝑁 − 1) × 𝐿𝐶 × 𝑀𝑘,𝐶𝐼

𝑃 + 2 × 𝐿𝐶 × 𝑀𝑘,𝐶𝐸𝑃 + 𝑁 × 𝐿𝐵 × 𝑀𝑘,𝐵

𝑃 (19)

The use of the above equation as a design parameter for the EBSD to an optimized level and hence, it is

considered as an objective function of the design problem and the constraints are defined on the various

designing parameters as described in equations 20, 21 and 22. Also, ASI code limits plastic hinges

formation, to prevent the total collapse of the structure due to side-sway mechanism has to occur only at the

end of beam instead of at the end of the column.

𝑀𝑘,𝐶𝐸𝑃 − 1.2𝑀𝑘,𝐵

𝑃 ≥ 0 (20)

𝑀𝑘,𝐶𝐼𝑃 − 1.2(2 × 𝑀𝑘,𝐵

𝑃 ) ≥ 0 (21)

In this study, optimized member sizes are obtained simplex method by minimizing a linear function of

several plastic moment variables under given constraints of design parameter.

The plastic moment carrying capacity of compressive members is affected by both bending moments as

well as the axial forces. In order to include induced axial force effect in plastic energy, recommendations

given by AISC 360-10[16] for axial force-bending moment interaction equation is used. The Modified

moment capacity𝑀𝑐, will depend upon modification factor𝛽𝑚, times required moment capacity𝑀𝑟, where

𝛽𝑚 =8

9{

1

(1−𝑃𝑟𝑃𝑐

)} 𝑓𝑜𝑟

𝑃𝑟

𝑃𝑐≥ 0.2 and 𝛽𝑚 =

8

9{

1

(1−𝑃𝑟

2𝑃𝑐)} 𝑓𝑜𝑟

𝑃𝑟

𝑃𝑐< 0.2 where,𝑃𝑟 is required axial force; 𝑃𝑐 is axial force

carrying capacity of the structure.

5. DESIGN EXAMPLE

For the example problem, a six-story, three-bay rectangular office building frame is designed using the

proposed Method. The dimensions of building frame are shown in figure 6. The chosen building is designed

example 2 for six-story by Popov[13]. The roof level beams are subjected to uniform dead and live loads of

87.74 kN/m and 17.55 kN/m, respectively while the dead and live loads on the remaining floor beams are

94.2 kN/m and 70.12 kN/m, respectively. The frame is assumed to be built in a location with site soil Class

C. A set of five earthquakes from the PEER Beta Data Base are used and scaled to match an IBC(2012)

response spectrum[4]. Using the preliminary member sizes, as designed by Popov, modal properties are

evaluated and is illustrated in table 1, in which the sum of total modal masses of all the three modes is more

than 90% of the total seismic mass. In this study, the building structure is considered at soil site C and

Bilinear Plastic hysteretic model and design ductility level 𝜇 = 4. Using the building parameter and the

normalized energy, modal normalized energy and energy distribution per mode is obtained and is given in

table 2. The total hysteretic energy demand is distributed to the different levels of the frame according to the

hysteretic energy distribution using equation 19. The forces and displacements to be used in eq. 3 are

obtained from modal pushover analysis results for Mode 1. The result of the pushover analysis is given in

table 3. The energy demand distribution over height (for Storey) is given in table 4. Ductility of structure is

taken as 4.

61



Table 2 Modal Property

Modes Period(S) Mr*(kg) m

1 2.02 48000 0.8431

2 0.71 42740 0.1065

3 0.39 47650 0.0316

0.9812

Table 3 Input energy per unit mass

Mode NE VI (IE/m)

(m2/s

2)

1 0.349 52.325 6.383

2 0.411 52.325 8.388

3 0.411 52.325 8.388

Table 3 Pushover Analysis

Storey Force(kN) Displacement(cm) Drift(cm)

1 202.75 30.42 30.42

2 462.98 50.16 19.74

3 622.02 65.74 15.59

4 793.03 75.17 9.43

5 946.29 83.07 7.90

6 966.58 87.84 4.77

Table 4 Hysteretic energy demand at different level

Stories(i) WEi WETotal Ri IEi(kNm)

1 1198.09 1105.29 0.408 451.40

2 758.18 1105.29 0.258 285.66

3 532.47 1105.29 0.182 200.61

4 243.53 1105.29 0.083 91.75

5 153.03 1105.29 0.052 57.66

6 48.33 1105.29 0.016 18.21

Figure 5 Six Story Three bay office building

62



Table 5 Plastic capacity demands

storey column

size

Axial Force(kN) βm

Plastic Moment(kNm)

capacity applied required modified

1 W14 X 159 6176.66 2353.21 1.436 433.75 622.85

W24 X 146 5525.05 4944.42 8.458 587.96 4973.17

2 W14 X 159 6598.06 1954.73 1.263 294.71 372.24

W24 X 146 5939.10 4079.12 2.838 355.61 1009.33

3 W14 X 132 5435.97 1550.01 1.243 167.83 208.69

W21 X 122 4932.82 3228.58 2.573 273.63 704.01

4 W14 X 132 5435.97 1143.82 1.126 112.35 126.48

W21 X 122 4932.82 2382.70 1.719 75.78 130.30

5 W14 X 99 4069.72 736.17 1.221 45.83 55.95

W18 X 86 3418.28 1537.95 1.616 80 129.27

6 W14 X 99 4069.72 342.91 1.092 15.18 16.58

W18 X 86 3418.28 734.76 1.132 30.35 34.36

The required plastic moment carrying capacities obtained are given in table 5 and is compared with the

plastic moment carrying capacities of the preliminary member sizes. If the plastic moment required is more

than plastic moment capacity of the preliminary member size, then the preliminary size needs to be changed

and if the plastic moment required is less than plastic moment capacity of the preliminary member size, then

the current size need not be changed.

Table 6 Member Sizes

Item Story Current

size

Plastic

moment(kNm) Member Size

Required Capacity initial iteration last iteration

Decision member Decision member

Co

lum

ns

1 W14 X 159 622.85 1167.34 Keep W24 X 162 Keep W24 X 162

W24 X 146 4973.17 1700.25 Change W24 X 146 Keep W27 X 178

2 W14 X 159 372.24 1167.34 Keep W14 X 159 Keep W14 X 159

W24 X 146 1009.33 1700.25 keep W24 X 162 Keep W24 X 162

3 W14 X 132 208.69 951.89 Keep W14 X 132 Keep W14 X 132

W21 X 122 704.01 1248.75 Keep W21 X 122 Keep W24 X 146

4 W14 X 132 126.48 951.89 Keep W14 X 132 Keep W14 X 132

W21 X 122 130.30 1248.75 Keep W21 X 122 Keep W21 X 122

5 W14 X 99 55.95 703.68 Keep W14 X 99 Keep W14 X 99

W18 X 86 129.27 756.55 Keep W18 X 86 Keep W18 X 86

6 W14 X 99 16.58 703.68 Keep W14 X 99 Keep W14 X 99

W18 X 86 34.36 756.55 Keep W18 X 86 Keep W18 X 86

Bea

ms

1 W27 X 94 1138.9 1138.9 Keep W27 X 94 Keep W27 X 94

2 W27 X 94 1138.9 1138.9 Keep W27 X 94 Keep W27 X 94

3 W24 X 94 1040.58 1040.58 Keep W24 X 94 Keep W24 X 94

4 W24 X 94 1040.58 1040.58 Keep W24 X 94 Keep W24 X 94

5 W21 X 68 655.48 655.48 Keep W21 X 68 Keep W21 X 68

6 W21 X 68 655.48 655.48 Keep W21 X 68 Keep W21 X 68

63



For multi-story multi-bays frames, the required column member sizes can be optimized in successive

iterations. New member sizes selected as required and the plastic moment capacity versus demand are given

in table 6. Hence after, the current member sizes have more capacity than demand needs. Therefore, current

size is taken as the final design sections.

Conclusion

As Energy-Based Seismic Design directly deals with energy, so it is considered as a more sensible

design method, whereas others methods deal with forces and displacement. The proposed seismic design

method is proposed by (i) relating the input and hysteretic energies for MDOF and equivalent SDOF

systems; (ii) proposing a hysteretic energy distribution for the multi-story-multi-bays frame and (iii)

developing design procedure for applying the EBSD procedure to steel moment resisting frame for multi-

bays.

REFERENCES

[1] Bureau of Indian Standards, Criteria for Earthquake Resistant Design of Structures, vol.

1893-2016.

[2] Decanini LD, Mollaioli F. An energy-based methodology for the assessment of seismic

demand. Soil Dynamics Earthquake Engineering (2001).

[3] Terapathana S. An energy method for earthquake resistant design of RC structures (2012).

[4] Mezgebo MG, Lui EM. A new methodology for energy-based seismic design of steel

moment frames. Earthquake Engineering (2017).

[5] Manfredi G. Evaluation of seismic energy demand, Earthquake Engineering Structural

Dynamics (2001).

[6] Chou C-C, Uang C-M. A procedure for evaluating seismic energy demand of framed

structures, Earthquake Engineering Structural Dynamics (2003).

[7] Wang F, Yi T. A Methodology for Estimating Seismic Hysteretic Energy of Buildings. Civ.

Eng. Urban Plan. 2012 (CEUP 2012), vol. 2012, 2012, p. 17–21.

[8] Seneviratna, Krawinkler H. Evaluation of inelastic MDOF effects for seismic design (1997).

[9] Akbas B, Shen J, Hao H. Energy approach in performance-based seismic design of steel

moment resisting frames for basic safety objective, Structure Design Tall Build (2001).

[10] G.W.Housner. Limit design of structures to resist earthquake. Earthquake Engineering

Structural Dynamics (1956).

[11] Merter O, Ucar T, Duzgun M. Determination of earthquake safety of RC frame structures

using an energy-based approach (2017).

[12] Mezgebo MG, Lui EM. Hysteresis and Soil Site Dependent Input and Hysteretic Energy

Spectra for Far-Source Ground Motions. Advance Civil Engineering (2016).

[13] Tsai K-C, Popov EP. Steel beam-column joints in seismic moment resisting frames(1988).

[14] ATC 40 1996. Seismic evaluation and retrofit of concrete building (1996).

[15] Federal Emergency Management agency (FEMA)- 355F, State of the Art Report on

Performance Prediction and Evaluation of Steel Moment-Frame Buildings (2000).

[16] American Institute for Steel Constructions. Specification for Structural Steel Buildings

2010.

64



Swelling behaviour of a remoulded expansive clay

blended with GGBS VAMSI NAGARAJU. T

1, PHANIKUMAR. B. R

2 AND MOUNIKA. K. N

3

1Assistant Professor of Department of Civil Engineering, S.R.K.R Engineering College, Bhimavaram

2Professor of Department of Civil Engineering, S.R.K.R Engineering College, [email protected]

3UG Student of Department of Civil Engineering, S.R.K.R Engineering College, Bhimavaram-534204

ABSTRACT

This paper presents the effect of ground granulated blast furnace slag (GGBS) on swelling

behaviour of a remoulded expansive clay. One-dimensional swell-consolidation tests were performed on

GGBS-clay blends varying GGBS content as 0%, 4%, 8% and 12% by dry weight of soil. Rate and

amount of heave, swell potential (S%) and swelling pressure (ps) were studied by performing the above

tests. It was found that the amount of heave and swell potential of blends decreased with increasing GGBS

content. Swelling pressure (ps) is decreased significantly with increasing GGBS content. The experimental

data suggests that ground granulated blast furnace slag (GGBS) can be considered as a useful additive in

expansive soil engineering.

Keywords: expansive clay, rate and amount of heave, swell potential, swelling pressure, GGBS

INTRODUCTION

The problems posed by expansive soils such as volume increase or swelling in monsoons and

volume reduction or shrinkage in summers, have been recorded all over the world. Volume increase or

swelling upon absorption of water is attributed due to the presence of mineral montmorillonite, high dry

densities and affinity for water (Chen, 1988; Lu and Lykos, 2004). Thus field expansive clay beds are

subjected to alternate swelling and shrinkage, lightly loaded civil engineering structures such as residential

buildings, pavements and canal linings founded in them experience severe distress. This is resulted in

heavy financial loss all over the world (Gourley et al. 1993).

Many innovative foundation techniques have been suggested for counteracting the detrimental

problems posed by expansive soils. These include special foundation techniques such as drilled piers,

belled piers (Chen, 1988), under-reamed piles (Sharma et al. 1978) and granular pile anchors

(Phanikumar, 1997), physical alteration techniques (Ranganatham and Satyanarayana, 1966), cushion

techniques (Satyanarayana, 1966; Katti, 1978). Chemical alteration technique as also becomes quite

successful in the amelioration of expansive clays (Chen, 1988).

In chemical alteration method different chemicals such as lime, cement, calcium chloride and fly

ash are added to expansive clays to reduce their swelling properties. Lime is quite effective in reducing

plasticity and volume change behaviour of expansive clays. Further, lime treatment of expansive clays

also increases their shear strength (Chen 1988; Evans 1998; Vamsi Nagaraju and Satyanarayana, 2016). A

technique called lime slurry pressure injection (LSPI) is also quite useful in fissured expansive clays.

Cement as an additive to expansive clay also reduces their plasticity and swelling and increases their shear

strength (Chen, 1988). Calcium chloride (CaCl2) which is a hygroscopic material that absorbs moisture

from atmosphere reduces shrinkage cracks in expansive clays and increases shear strength (Phanikumar

and Sastry, 2001).

GGBS is an industrial waste, is also a siliceous pozzolanic material being used as an additive to clays

and cements. It reduces plasticity and compressibility of clays (Higgins, 2005; Yadu and Tripathi, 2013;

Vamsi Nagaraju et al. 2017).


65



This paper presents experimental data on free swell index (FSI) of an expansive clay powder

blended with varying amounts of lime and RHA. Further, two series of swell-consolidation tests were

performed. In one series, clay-RHA blends were the samples in which the RHA content was varied. And

in another series, 4% lime was added to the above clay-RHA blends, and the tests were conducted to

study rate and amount of heave, swelling pressure, rebound and linear shrinkage.

EXPERIMENTAL INVESTIGATION

Test materials

A highly swelling expansive clay collected at a depth of 1 meter from the ground level from the town

of Bhimavaram, AP, India, was used in the experimental investigation. It had a free swell index (FSI) of

145%. Based on its LL of 78% and PI of 49%, the soil may be classified CH. GGBS was collected from

NTPS, AP, India. It was a non-plastic, silt-sized material. Tables 1 show the index properties of the soil.

Table 1. Index properties of expansive clay

Property Value/Remarks

Specific gravity 2.69

Liquid limit (%) 78

Plastic limit (%) 29

Plasticity index (%) 49

Gravel (%) (>4.75 mm) 0

Sand (%) (4.75 - 0.75 mm) 01

Silt (%) (0.075 – 0.002 mm) 39

Clay (%) (<0.002 mm) 60

Free swell index, FSI (%) 145 USCS classification CH

Quantities determined and variables studied Rate and amount of heave, swell potential (S%) and swelling pressure (ps) of the GGBS-

clay blends were determined. GGBS content was varied as 0%, 4%, 8%, and 12% by dry weight

of the soil. In swell- consolidation tests, the initial water content (wi) of the specimens was kept

constant at 0% and the dry unit weight (γd) was kept constant at 12kN/m3.

Tests and procedures

Swell-consolidation tests The oven-dry expansive clay passing 4.75mm sieve was weighed corresponding to the γd chosen and

the volume of the consolidation ring (diameter = 60mm, height = 20mm) for conducting one-dimensional

swell-consolidation tests. However, in the tests on GGBS-clay blends, the clay was replaced by the

required amount of GGBS based on its dosage. The GGBS-clay blends were thoroughly mixed and

statically compacted in the consolidometer ring in four layers each of 5mm thickness so as to ensure a

uniform γd. A filter paper and porous stone were placed at each end of the sample, and this unit was placed

in the consolidometer after positioning the loading pad. This assembly was mounted on the loading frame,

and the samples were allowed to undergo free swell by inundation for three days (exactly 72 hours or 4320

minutes) under a token surcharge of 5kPa. After the equilibrium heave, the samples were subjected to

consolidation under increased vertical stresses.

Swell potential (S%) was determined as the ratio of increase in thickness (ΔH) to the original thickness

(H), expressed as S%. And swelling pressure (ps) was determined from the e-log p curves as the pressure

corresponding to the initial void ratio.

66



e RESULTS AND DISSCUSION

Performances of GGBS-clay blends

To study the effect of GGBS on one-dimensional swell-consolidation of expansive soil, various

amounts of GGBS such as 0%, 4%, 8% and 12% by dry weight of soil were added and effectively mixed

and tested. Table-2 and Figuers 1-4 shows the entire test results.

Table 2. Effect of GGBS on swell-consolidation properties

Property determined GGBS content (%)

0 4 8 12

Heave (mm) 1.41 1.12 1.0 0.79

Swell potential (%) 7.05 5.6 5 3.95

Swelling pressure, ps (kPa) 88 80 65 60

1.65

1.55

1.45

1.35

1.25

1.15

1.05

1 10 log p (kPa)100 1000

Fig 3. e-log p curves (GGBS)

90

80

70

60

50

0 5 10 15

GGBS content (%)

Figure 4. Influence of GGBS content on swelling pressure

Figure 1 shows the rate of heave profiles of clay-GGBS blends in the form of heave (mm) and log time

(minutes) plots. The data pertain to different GGBS contents (0%, 4%, 8% and 12%). The equilibrium

heave was attained by the specimens in 3 days (4320 minutes). The unblended specimen (0% GGBS)

Sw

elli

ng

pre

ssure

(kP

a)

0% additive

4% GGBS 8% GGBS

12% GGBS



resulted in a heave of 1.41 mm. However, when GGBS was added to the clay in increasing amounts,

heave decreased significantly. The measured heave (mm) was 1.41mm, 1.12mm, 1.0mm and 0.88mm

respectively for GGBS contents of 0%, 4%, 8% and 12%. When expansive clay particles are replaced by

non-expansive GGBS particles, interaction between the clay particles decreases, resulting in reduced

amount of heave. Further, flocculation taking place upon adding GGBS to the expansive clay would also

effectively reduce heave. Therefore, swell potential (S%) of the blend samples was 7.0%, 5.6%, 5% and

3.9% respectively for the GGBS contents of 0%, 4%, 8% and 12%.

Figure 2 shows the variation of swell potential (S%) with additive content of the GGBS-clay blends.

S% decreased continuously with increasing GGBS content.

Swelling pressure (ps) of GGBS-clay blends was determined from e-log p curves as the pressure

corresponding to the initial void ratio eo, which was 1.24. Figure 3 shows the e-log p curves of GGBS- clay

blends for different GGBS contents. The swelling pressure (ps) was 88kPa, 80kPa, 65kPa and 60kPa

respectively for the GGBS contents of 0%, 4%, 8% and 12%. As GGBS content in the blends increased,

heave decreased and so, swelling pressure also decreased.

Figure 4 shows the variation of swelling pressure (ps) with GGBS content. When GGBS content

increased from 0% to 12%, ps decreased from 88 kPa to 60 kPa, indicating a reduction of 32%.

CONCLUSIONS

The following conclusions can be drawn from the foregoing experimental study:

1) Upon addition of GGBS to the expansive clay, both the amount of heave and rate of heave decreased.

When GGBS content increased from 0% to 12%, heave decreased from 1.41mm to 0.88mm, showing a

reduction of 38

2) Swelling pressure (ps) also decreased with increasing GGBS content in the blends. Swelling pressure

decreased from 88kPa to 60kPa when GGBS content increased from 0% to 12%, indicating a reduction of 32%.

REFRENCES

[1] Chen, F. H. (1988). “Foundations on expansive soils”, Elsevier Scientific Publishing Co., Amsterdam. [2] Evans, P. (1998). Lime stabilization of black clay soils in Queens land, Australia, Presentation to the

National Lime Association Convention, San Diego, California. [3] Gourley, C. S., Newill, D. and Schreiner, H. D. (1993). Expansive soils: TRL’s research strategy,

Proceedings, 1st International Symposium on Engineering characteristics of arid soils, London. [4] Katti, R.K. (1978). “Search for solutions to problems in black cotton soils”, First I.G.S Annual Lecture,

Indian Geotech. Society at I.I.T., Delhi. [5] Lu, N. and Lykos, W. (2004). Unsaturated Soil Mechanics, Wiley New York. [6] Phanikumar, B. R. (1997). A study of swelling characteristics of and granular pile-anchor (GPAF)

foundation system in expansive soils. PhD thesis. JN Technological University, Hyderabad, India. [7] Phanikumar, B. R., and Sastry, M. V. B. R. (2001). Stabilizing swelling subgrades with calcium chloride,

Highway Research Bulletin, Vol. 65, Journal of Indian Roads Congress, pp. 77-82. [8] Ranganatham, B.V. and Sathyanarayana, B. (1965). “A rational method of predicting swelling potential

for compacted expansive clays”, Proc. 6th Int. Conf. S.M. & F.E., Canada, Vol.1, pp. 92-96. [9] Sharma, D., Jain, M. P. and Prakash, C. (1978). Handbook on underreamed and bored compaction pile

foundations. Roorkee, India: Central Building Research Institute. [10] Vamsi Nagaraju.T, Surya Narayanaraju.J, Sairam. M.V.K (2017). Effect of lime and ground

granulated blast furnace slag on engineering behaviour of expansive clays, National conference on recent advancement in geotechnical engineering, ISBN: 978-81-931273-7-7, pp 18-21, March 2017, GCT, Coimbatore

[11] Vamsi Nagaraju.T, Satyanarayana P.V.V (2016). Improving the Characteristics of Expansive Sub grade Soils Using Fly Ash, National conference on sustainable materials and management systems in civil engineering, Dec 2016, CBIT, Hyderabad, pp 18-21, ISBN: 978-81-932824-8-9

68



Structural Health Monitoring Using Non-Destructive

Testing

Ajeet Kumar1, Vijay Pal Singh

2

1Postgraduate Student, National Institute of Technology Kurukshetra, India, E-mail: [email protected]

2Professor, National Institute of Technology Kurukshetra, India, E-mail: [email protected]

ABSTRACT

Any Civil Engineering structure is made off by assembling various members which transfer loads from

one to another and ultimately to the foundation. For structure to be safe all members must be sufficiently

able to carry these loads. During the design stage of a structure all the possible loads and their combinations

which may occur during the life of structure are considered and members are designed accordingly using

different materials like concrete, steels woods etc. As most common material used in civil engineering

structure is concrete which deteriorates with passage of time because of excessive loading and environmental

factors like freezing and thawing, temperature stresses etc. assessment of strength of various members at this

time becomes essential to avoid the economic loss and loss of life due to failure of structures. In the past,

various methods and techniques known as Non-Destructive Testing has been developed to monitor the

structural health of the structures. In the present study an attempt has been made to evaluate the strength and

quality of concrete using Rebound Hammer and Ultrasonic Pulse Velocity. The effects of various factors as

moisture content, deterioration of concrete, stress levels on relative strength and quality prediction has been

studied. For the present study 24 Nos. of cubes have been casted using M25 grade of concrete and effect of

various factors has been studied.

Key Words: Ultrasonic pulse velocity, Rebound Hammer, Cracks.

INTRODUCTION

Infrastructure of a country, whether developed or developing, consists many old as well as new

structures like bridges, roads, tunnels, high rise buildings, water tanks, power plants etc. and huge cost is

invested to keep them in working condition without any failure. In all above mentioned structures concrete is

generally most commonly used construction material. Concrete is very susceptible to a variety of

environmental degrading factors like freezing and thawing, sulfate attack, temperature etc. which tends to

limit the service life of the structures. This degrading nature of concrete has brought about the need to

develop the test methods to measure the in-situ materials properties for quality assurance and evaluation of

the structure at different interval of times. These test methods are expected not to damage the structure and

allow the re-testing at the same position to evaluate the change in properties if desired at later. These

methods and techniques are called non-destructive testing.

There are many occasions when the various performance characteristics of concrete in a structure are

required to be assessed. In most of the cases, an estimate of strength of concrete in the structure is needed,

although parameters like overall quality, uniformity etc. also becomes important in others. The various

methods that can be adopted for in-situ assessment of strength properties of concrete depend on the various

aspect of the strength. At present the direct test used mainly as a basis of quality control is compression

testing of cubes, cylinders etc. and it represents the potential strength of the concrete used in the structure.

The main parameters determining the qualities of concrete are its composition of mix, degree of compaction

and curing. At the most it can be ensured that the composition of concrete going into the cubes and that

going into the structure is the same. However, the methods of compaction and curing usually are different for

the cubes and the structural members as quality control in laboratory is better that quality control in field. Due to this variation in quality control, results obtained on cubes may not truly represent the quality of

concrete in the structure. Hence the use of Non-Destructive Testing on the newly constructed structure

becomes necessary.

Over last few decades, many researchers carried out structural health monitoring using various types of

non-destructive method. The very common non-destructive technique like Rebound Hammer, Ultrasonic

Pulse Velocity, and Permeability has been used to predict the damage as well as strength of existing

structures. Objective of the present study is to find the effects of various factors such as moisture content,



69



deterioration of concrete, damage under uniaxial stress etc. on the non-destructive testing of the concrete

using ultrasonic and rebound hammer.

TEST SPECIMENS AND METHODOLOGY

An experimental study has been carried out to achieve the objectives of the present study. Concrete mix

design of grade M25 has been used to cast the specimens. 24 Nos. of cubic specimens of size 150mm x

150mm x 150mm have been casted in cast iron moulds. These specimens were cured for 28 days in water

curing tank at 28̊C. After curing, testing have been performed to find the effects of various factors such as

moisture content, deterioration of concrete and damage under uniaxial stress etc. Testing have been done

using Proceq Ultrasonic tester and Rebound Hammer to find the effects of various above mentioned factors.

For the study of effects of moisture on UPV and RH, testing has been done at the interval of one week and

until there were consistency in results and average value became constant. Concrete specimens have been

damaged by shear stress and UPV have been measured before and after damage to find the effect of

deterioration of concrete on UPV. To find the effect of different level of compressive stress on ultrasonic

pulse velocity and compressive strength by rebound hammer, both ultrasonic and rebound hammer testing

has been done at different loads. For this load has been kept at hold on values of 0 kN, 151.73 kN, 315.15

kN, 642.65 kN and 806.17 kN and ultrasonic pulse velocity and compressive strength by rebound hammer is

measured at each loads. Loading is further increased and ultimate strength of cubes has been determined.

RESULTS

Effect of Moisture on Ultrasonic Pulse Velocity and Rebound Hammer Testing

To find the effect of moisture on ultrasonic pulse velocity and rebound hammer, cubic specimens has

been casted and dried in sunlight during day time and tested at 1day, 7days, 14days, 21days and 28days. The

results of effect of moisture on UPV and RH are presented in the Table 1 and Table 2 respectively. These

results have also been plotted in Figure1 and Figure 2 respectively.

Table 1: Variation of UPV with Age of Drying of Cube

Cube No. UPV at 1day

Dry (m/sec)

UPV at 7day

Dry (m/sec)

UPV at 14day

Dry (m/sec)

UPV at 21day dry

(m/sec)

UPV at 28day

Dry (m/sec) 301/1 4601 4464 4491 4499 4560

301/2 4532 4491 4630 4545 4532

301/3 4601 4539 4630 4491 4435

301/4 4605 4532 4532 4398 4582

301/5 4673 4464 4491 4495 4545

301/6 4747 4601 4425 4530 4491

301/7 4618 4615 4559 4491 4478

301/8 4601 4532 4602 4559 4543

301/9 4532 4747 4491 4559 4491

301/10 4535 4505 4430 4561 4473

301/11 4673 4532 4425 4587 4573

301/12 4532 4601 4464 4559 4460

311/1 4601 4532 4491 4491 4601

311/2 4635 4688 4762 4673 4505

311/3 4673 4573 4360 4532 4532

311/4 4673 4601 4630 4486 4499

311/5 4714 4525 4777 4518 4491

311/6 4664 4532 4559 4532 4601

311/7 4545 4601 4559 4558 4573

311/8 4549 4532 4532 4560 4601

311/9 4747 4535 4491 4559 4560

311/10 4532 4615 4491 4630 4673

311/11 4605 4559 4702 4593 4559

311/12 4601 4601 4491 4492 4591

Sum 110789.0 109517.0 109015.0 108898.0 108949.0

Average

Value 4616.2 4563.2 4542.3 4537.4 4539.5

70



Table 2: Variation of Compressive Strength by RH with Age of Drying of Cube

Cube No.

Compressive

Strength by

R.H. at 1day dry

(MPa)

Compressive

Strength by

R.H. at 7day

dry (MPa)

Compressive

Strength by

R.H. at 14day

dry (MPa)

Compressive

Strength by

R.H. at 21day

dry (MPa)

Compressive

Strength by

R.H. at 28day

dry (MPa)

301/1 25.5 30.0 33.0 33.5 33.0

301/2 27.0 29.5 30.5 29.5 30.5

301/3 24.5 31.0 33.5 34.5 31.5

301/4 26.0 32.0 32.5 30.5 32.5

301/5 27.0 31.5 31.0 30.5 31.0

301/6 25.5 28.0 30.5 32.5 31.0

301/7 24.0 30.5 30.5 32.0 32.0

301/8 24.5 29.5 32.5 31.5 30.5

301/9 28.5 32.0 33.0 34.0 33.5

301/10 29.0 30.0 30.5 30.5 32.5

301/11 25.0 31.0 32.5 31.5 33.5

301/12 24.5 30.5 32.5 33.5 33.0

311/1 26.0 31.0 32.0 33.0 32.0

311/2 24.5 30.0 31.5 29.5 31.0

311/3 26.5 30.0 30.5 32.0 32.0

311/4 24.5 30.5 33.0 31.0 31.0

311/5 24.5 29.0 32.5 34.0 34.5

311/6 26.0 31.0 32.5 31.0 31.0

311/7 27.5 31.5 31.0 30.5 31.5

311/8 27.0 30.0 34.0 33.0 33.0

311/9 27.0 30.5 30.0 32.5 33.5

311/10 26.0 29.5 29.0 31.5 31.0

311/11 28.5 32.0 33.5 32.5 32.5

311/12 26.5 31.0 31.5 33.5 32.5

Sum 625.5 731.5 763.5 768.0 770.0

Average

Value 26.1 30.5 31.8 32.0 32.1

Figure 1: Effect of moisture on UPV Figure 2: Effect of moisture on compressive strength

by rebound hammer

4500.0

4525.0

4550.0

4575.0

4600.0

4625.0

4650.0

4675.0

4700.0

0 10 20 30

Ult

raso

nic

Pu

lse

Ve

loci

ty (

m/s

)

Age of Drying (Days)

Average Value

20.0

22.0

24.0

26.0

28.0

30.0

32.0

34.0

0 10 20 30

Co

mp

ress

ive

str

en

gth

by

Re

bo

un

d

Ham

me

r (M

pa)

Age of Drying (Days)

Average Value

71



The results of the ultrasonic pulse velocity and rebound hammer testing shows that ultrasonic pulse

velocity decreases and compressive strength by RH increases with age of drying and after some days these

becomes constant. An increase in ultrasonic pulse velocity has been observed about 1.69% and a decrease in

compressive strength by RH has been observed about 18.77% due to moisture in the specimens.

3.2. Effect of Deterioration of Concrete on Ultrasonic Pulse Velocity

Deterioration of the concrete has been simulated by intentionally giving micro cracks in concrete cubes

due to shear stress. Ultrasonic pulse velocity has been measured before and after deterioration of concrete

across the crack developed due to shear stress and results are shown in Table 3.

Table 3: Effect of Deterioration Concrete on UPV

Cube No. Shear Strength

(MPa)

UPV Before

Damage (m/sec)

UPV After Damage

(m/sec) Percentage Decrease

301/1 3.49 4658 2389 48.71

301/2 3.54 4357 2318 46.80

301/3 3.88 4502 3833 14.86

301/4 2.82 4539 1574 65.32

301/5 3.73 4601 1836 60.10

301/6 3.49 4445 2869 35.46

From the above table it can be seen that UPV after cracking due to shear stress UPV decreased very

large. It can be said that deterioration of concrete due to presence of even micro cracks can cause to decrease

UPV significantly.

3.3. Effect of Uniaxial compressive Stress on Ultrasonic Pulse Velocity and Rebound Hammer

To find the effect of compressive stress on UPV and RH, cubes have been subjected to uniaxial

compressive stress in universal testing machine and testing has been done by ultrasonic apparatus and

rebound hammer. Both UPV and RH Testing have been done on average loads of 0kN, 151.78kN, 315.12kN,

478.74kN, 642.72kN and 806.60kN. Results of test have been shown in Table 4.

Table 4: UPV and RH test Results on Cube under Uniaxial Stress

Sl.

No.

Cube No: 301/8

Cube No: 301/9

Cube No: 301/10

Load

(kN)

UPV

(m/sec)

Comp.

Strength

by RH

(MPa)

Load

(kN)

UPV

(m/sec)

Comp.

Strength

by RH

(MPa)

Load

(kN)

UPV

(m/sec)

Comp.

Strength

by RH

(MPa)

1 0.00 4445 33.0 0.00 4515 32.0 0.00 4454 33.0

2 151.61 4438 35.0 152.14 4573 32.5 152.01 4438 32.5

3 315.23 4438 34.0 315.73 4486 32.0 316.06 4490 33.0

4 478.84 4310 34.5 480.01 4316 32.0 479.63 4263 31.5

5 643.36 4190 26.5 643.95 3258 28.0 642.28 2339 27.0

6 807.22 2869 19.0 807.21 2290 24.5 758.63

7 817.56 814.62

Continue-

72



Cube No: 311/2 Cube No: 311/3 Cube No: 311/4

Load

(kN)

UPV

(m/sec)

Comp.

Strength

by RH

(MPa)

Load

(kN)

UPV

(m/sec)

Comp.

Strength

by RH

(MPa)

Load

(kN)

UPV

(m/sec)

Comp.

Strength by

RH (MPa)

0.00 4495 33.50 0.00 4491 33.0 0.00 4505 33.0

152.77 4425 33.50 151.37 4473 32.0 150.21 4559 33.5

317.43 4425 32.50 314.83 4398 30.5 312.31 4559 33.0

480.98 4306 30.50 478.03 3729 30.0 475.25 4491 32.0

642.38 3067 27.00 642.57 1999 26.5 640.67 3563 32.5

806.80 1838 24.50 676.21 805.16 1456 29.5

812.82 816.95

Continue-


Load

(kN)

UPV

(m/sec)

Comp.

Strength

by RH

(MPa)

Load

(kN)

UPV

(m/sec)

Comp.

Strength

by RH

(MPa)

Load

(kN)

UPV

(m/sec)

Comp.

Strength by

RH (MPa)

0.00 4535 31.0 0.00 4410 31.50 0.00 4659 32.50

151.31 4493 31.5 151.62 4476 32.00 151.78 4601 31.50

314.76 4336 30.0 313.87 4365 33.00 314.95 4265 31.50

481.32 3976 29.5 477.79 3882 27.50 479.40 3989 30.50

644.00 1834 26.5 642.10 2433 25.00 642.01 1836 29.50

699.73 801.36 682.01

Continue-


Load

(kN)

UPV

(m/sec)

Comp.

Strength

by RH

(MPa)

Load

(kN)

UPV

(m/sec)

Comp.

Strength

by RH

(MPa)

Load

(kN)

UPV

(m/sec)

Comp.

Strength by

RH (MPa)

0.00 4513 32.0 0.00 4530 31.0 0.00 4425 32.0

151.61 4458 33.0 154.44 4491 33.0 151.13 4487 32.5

315.23 4362 33.0 315.49 4165 32.0 315.80 4376 32.0

478.84 3936 31.5 478.85 3816 32.5 479.84 3724 30.5

643.36 2408 28.5 642.11 2102 29.0 643.94 2679 30.0

698.24 806.59 1736 28.5 702.10

811.84

Table 5: Average UPV and RH Test Results on Cubes under Uniaxial Stress

Sl. No. Load (kN)

Average Compressive

Strength under Crushing

(MPa)

Average UPV

(m/sec)

Average Compressive

Strength

by RH (MPa)

1 0

33.64

4503.87 32.50

2 151.78 4498.80 32.73

3 315.12 4403.53 32.37

4 478.89 4041.33 31.30

5 642.74 2652.53 28.13

6 806.60 2037.80 25.20

Figure 3 and Figure 4 shows the plot of load versus average UPV and load versus average compressive

strength by RH respectively.

73



Figure 3: Plot between UPV and Uniaxial Figure 4: Plot between Compressive Strength

Loads on Cubes by RH and Uniaxial Loads on Cube.

Test results of UPV and RH under uniaxial loads on cube shows that there is no significant variation in

UPV and compressive strength by RH until the load reaches a value of upto nearly 60 percent of ultimate

strength of cubes then both shows decrease in values. As soon as load reaches a value of nearly 60 % of

ultimate load cracking appears on the surface and both UPV and RH values starts to decrease.

CONCLUSIONS

The compressive strength in the presence of moisture has been found to be decreased about 18.77%. An

increase in ultrasonic pulse velocity of about 1.69% has been observed for cubic specimens.

Presence of crack in concrete causes decrease in UPV and test of UPV on deteriorated concrete by shear

stress indicated a large decrease in Ultrasonic pulse velocity due to presence of single micro crack.

From test results of UPV and RH on cubes under uniaxial stress, it is concluded that growth of micro-

crack in concrete can be evaluated by UPV. It can also be concluded that UPV and RH value is not

affected significantly upto as large stress as nearly 60% of ultimate strength but after that sharp decrease

both in UPV and RH noted.

REFERENCES

1. IS: 13311 (Part 1)-1992, “Non-Destructive Testing of Concrete- Methods of Test”, Bureau of Indian

Standards.

2. IS 13311 (Part 2) 1992, “Non-Destructive Testing of Concrete- Methods of Test”, Bureau of Indian

Standards.

3. Kolaiti, E., and Papadopoulos, Z., “Evaluation of Schmist Rebound Hammer Testing: A Critical

Approach,” Bulletin of the International Association of Engineering Geology, Paris-N°48-October1993.

4. Yaman, I.O.; Inci, G.; Yesiller, N.; and Atkan, H.M., “Ultrasonic Pulse Velocity in Concrete Using

Direct and Indirect Transmission,” ACI Materials Journal, Title no. 98-M48, 2001.

5. Subramaniam, K.V.; Mohsen, J.P.; Shaw, C.K.; and Shah, S.P., “Ultrasonic Technique for Monitoring

Concrete Strength Gain at Early Age,” ACI Materials Journal, Title no. 99-M46, 2002.

6. Lin, Y.; Lai, C.P; and Yen, T., “Prediction of Ultrasonic Pulse Velocity (UPV) in Concrete,” ACI

Materials Journal, Title no. 100-M3, 2003.

7. Akkaya, Y.; Voigt, T.; Subramaniam, K.V.; and Shah, S.P., “Nondestructive Measurement of Concrete

Strength Gain by an Ultrasonic Wave Reflection Method,” Materials and Structures / Matdriaux et

Constructions, Vol. 36, October 2003, pp 507-514.

8. Lee, H.K.; Lee, K.M.; Kim, Y.H.; Yim, H. and Bae, D.B., “Ultrasonic In-situ Monitoring of Setting

Process of High-performance Concrete,” Cement and Concrete Research 34 (2004) 631-640.

0

200

400

600

800

1000

0.00 10.00 20.00 30.00 40.00

Load

(kN

)

Comperessive Strength by RH (Mpa)

Average of all cues

0100200300400500600700800900

0 700 1400 2100 2800 3500 4200 4900

Load

(kN

)

Ultrasonic Pulse Velocity (m/sec)

Average of all Cubes

74



9. Voigt, T.; Grosse, C.U.; Sun, Z.; Shah, S.P.; and Reinhardt, H.W., “Comparison of Ultrasonic Wave

Transmission and Reflection Measurements with P- and S-waves on early Age Mortar and Concrete,”

Materials and Structures 38 (October 2005) 729-738.

10. Stauffer, J.D.; Woodward, C.B.; and White, K.R., “Nonlinear Ultrasonic Testing with Resonant and

Pulse Velocity Parameters for Early Damage in Concrete,” ACI Materials Journal, Title no. 102-M14,

2006.

11. Turgut, P., and Kucuk O.F., “Comparative Relationship of Direct, Indirect and Semi-direct Ultrasonic

Pulse Velocity Measurements in Concrete,” ISSN 1061-8309, Russian Journal of Nondestructive testing,

2006, Vol. 42, No. 11, pp. 745-751. © Pleiades Publishing, Inc., 2006.

12. Ulucan, Z.C.; Turk, K.; and Karatas, M., “Effect of Mineral Admixtures on the Correlation between

Ultrasonic Velocity and Compressive Strength for Self-Compacting Concrete,” ISSN 1061-8309,

Russian Journal of Nondestructive Testing, 2008, Vol. 44, No. 5, pp. 367–374. © Pleiades Publishing,

Ltd., 2008.

75



TO STUDY THE EFFECT OF IMPACT LOADING

ON BRIDGE PIER

Sunil Kumar Yadav1*, Vijay Pal Singh2 1* PG Student, Department Civil Engineering, NIT Kurukshetra, India

2 Professor, Department Civil Engineering, NIT Kurukshetra, India

*Corresponding author (E-mail:[email protected]; +91-9457486710)

ABSTRACT

The role of bridge in today’s infrastructure has become just like a vital organ of infrastructure. The bridges have been classified depending upon type of material used, the type of construction, geometry and design aspects.

There are various component in the bridges. Among various component of bridge like Deck Slab, Longitudinal Girder, Transverse Girder and Bearing, Piers are important component of bridge system. Because Piers takes the load from bearing and transfer it to stratum. So, failure of the pier will fail the whole system of bridges.

Irrespective of bridge type and component, the nature of loads acting on the bridges remains the same.

There are different type of loading acting on bridge like dead load, live load, wind load, etc. and among

them the impact loading generated through live load plays an important role in the life of bridges. The

impact loading on the deck slab, girder and piers/abutment causes deterioration with time.

In the present study, the behaviour of bridge pier under impact loading has been conducted and also the

prediction of life cycle based upon the failure of material under impact loading has been conducted. The

study has been carried out with the help of the Finite Element Analysis. For the analysis ANSYS 18

software has been used. The study has been carried out on a 3D model of hammerhead pier generated in

ANSYS.

The major factor that affects the behaviour of pier is Stress, Strain and Fatigue loading which were

introduced in studying the effect of impact loading on bridge pier.

Keywords: Impact Load, ANSYS18, Piers, Vehicle, Bearings.

1. INTRODUCTION

A bridge is a structure providing passage over an obstacle without closing the way beneath. The

required passage may be for a road, a railway, pedestrians, a canal or a pipeline. The obstacle to be crossed

may be a river, a road, railway or a valley. In other words, bridge is a structure for carrying the road traffic or

other moving loads over a depression or obstruction such as channel, road or railway.

There are various component in the bridges. The behaviour of various components of bridges under

cyclic loading has been studied experimentally and analytically by various researchers. It has been observed

that due to continuous loading and unloading, the bridge component like deck, girders are subjected to

failure because of fatigue loading. Further small bridges or box type culvert, where there is no provision of

bearing, the abutments and piers are subjected to same type of fatigue loading. In the present study, the

behaviour of pier under fatigue has been studied.

The analysis of a structure has been carried out by modelling it, using suitable software. This paper

includes detailed description of modelling of bridge Pier and describes a geometrical and physical properties

of bridge pier. Also describe placing of load at different position on face. Finally the analysis of bridge pier

has been carried out by using ANSYS 18 workbench. The bridge pier along with cap has been modeled

using FEM.

76



2. MODELLING AND ANALYSIS

Modelling and analysis of bridge pier has been carried out by using ANSYS 18 workbench. The bridge

pier along with cap has been modeled using Finite Element Method.

The FEM is a numerical technique for solving differential equation. This method translate partial differential equation into a set of linear algebraic equation. The structure to be analysed using FEM got discretized into a finite element & connected to each other at joint. These joint are called as nodes

2.1 PROCRDURE

The procedure adopted for modelling and analysis has been explained as shown below.

2.1.1 ANSYS WORKBENCH ANSYS Workbench, which is used to perform various types of structural, thermal, fluid, and

electromagnetic analysis. Workbench is a project management tool in which different tool are available which are helpful to solve the problem.

2.1.2 STATIC STRUCTURA LANALYSIS

It is the structure analysis which is used deformation, stress strain. In this analysis loading system do

not create any inertia and damping effects in body. In the static analysis the load and structure response are

very slow with respect to time i.e. zero to one, zero to two etc.

2.1.3 ENGINEERING PROPERTIES

Material is concrete and steel

Grade is concrete is M35 and Steel is Fe415

Density of Material 2400 kg/m3

Young Modulus of Material is 3e+10 Pa

Poisson Ratio of Material is 0.18

Shear Modulus of Material is 1.2712e+10 Pa

Bulk Modulus of Material is 1.5625e+10 Pa

Tensile Ultimate Strength of concrete (M35) is 4.14e+6 Pa

Compressive Ultimate Strength of concrete (M35) 35MPa

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2.1.4 GEOMETRICAL PROPERTIES

The various Geometrical properties and loadings used for modeling of “Hammerhead Pier” are as given below in Table 2.1

Table 2.1 Geometry Properties

Geometry Properties

Length X 11.5 m

Length Y 2. m

Length Z 7.3 m

Volume 19.43 m³

Mass 44689 kg

Centroid X 5.73 m

Centroid Y 0.5 m

Centroid Z 1.597 m

3. RESULTS

The Bridge Pier is one of the most important component of a bridge. It is a part which is connected with

substructure and semi connected with superstructure being bearing in between superstructure and

substructure. In the present study the analysis of bridge pier has been carried out by using ANSYS 18. The

bridge pier along with pier cap has been modeled using FEM.

The pier is used in the analysis is hammerhead pier and various properties have been discussed in the

previous chapter. The behaviour of hammerhead bridge pier subjected to cyclic loading has been studied.

The various structural characteristic like deformation, stress- strain has been calculated and given below.

Figure 3.1 Loading Pattern

78



3.1 LOAD V/S DEFLECTION CURVE

The load-deflection behaviour has been shown in the Fig. 3.2. It has been observed that there are three

zone in curve i.e. Elastic, Plastic, Failure zone. As per IS 456-2000[8] the behaviour of concrete is also

divided into these three zone. Fig3.2 shows that curve follows a linear variation up to a point (a) and it

change linear to nonlinear due to exceeding strain value of 0.002 at a loading of 3750KN (total loading on

pier is 15000KN), and similar change has been observed at point (b) subjected to a loading of 4500KN. The

strain at that point is .0032. But after the point (b), it can be easily seen in the curve, major change have taken

place. The change is occur due to presence of steel because at strain value of 0.0035 concrete would fail,

and load would be transfer to the steel.

The same observation has also been verified by IS456-2000[8] where the stress- strain variation is

linear up to a strain value of 0.002 and after which suddenly change occur up to a 0.0035, after this value

of strain concrete would fail.

Figure 3.2 Load v/s Displacement

3.2 BEHAVIOR OF PIER AFTER CYCLIC LOADING

The various researchers found that some engineering component of machine, bridges, aircraft, ships are

fail due to fatigue. The major responsible forces for the fatigue failure is the repetitive load or cyclic

loading.

It has also been observed that there are two type of fatigue i.e. high fatigue cycle and low fatigue cycle.

If the material sustain loading more than 100000 cycle, it is called high fatigue cycle and below this value

it is called low cycle fatigue.

It has been observed from Fig. 3.3 that, up to point (a) there is no change in the behavior of cycle. After

the point (a) there is sudden change in the curve observed because after point (a), with the increasing of

loading there will be decrease in the value of cycle.

It is observed in the figure that up to point (a) where 1500KN load is act and till that point no change in the cycle is observed but after the 1500KN and up to a 4500 there is major change occur at every interval of loading. But after the 4500 there are slight change has been observed in the cycle and could be observed from curve.

79



Figure 3.3Cycle (log) v/s Load

3.3 STRESS- STRAIN BEHAVIOR UNDER THE COMBINATION OF ALL LOADING

Fig. 3.4 showed that the behaviour of stress-strain of under various loads considered. It has been

observed from the Fig. 3.4 that the stress-strain value is linear up to strain value of 0.0018 nearly to 0.002

at a stress value of 25.15 MPa and then follow a straight line up to a strain value of 0.0035 after that

concrete would fail and transfer the load to the steel.

The same observation has also been verified by IS456-2000[8] where the stress- strain variation is

linear up to a strain value of 0.002 and after which suddenly change occur up to a 0.0035. After this value

of strain concrete would fail.

Figure 3.4Stress Strain Behaviour at Combination of All Loading



4. CONCLUSION

On the basis of analytical studied the following conclusion has been made.

1. There is no failure due to cyclic designed loading (i.e. 850KN in present study) of pier, irrespective of the number of cycle.

2. The overloading and its repetition caused failure after few cycles, which show that there is no

sudden failure because of overloading.

3. In stress-strain curve, concrete would fail at a value of strain 0.0035 and transfer of stresses from

one material to other i.e. concrete to steel would take place.

REFERENCES

[1] Byung Hwan Oh, (1986), “Fatigue Analysis of Plain Concrete in Flexure”, Journal of Structural Engineering, Vol. 112, No. 2, February 1986.

[2] Young J.Park, (1990), “Fatigue of Concrete under Random Loadings”, Journal of Structural

Engineering, Vol. 116, No. 11, November 1990. [3] Wing-Pin Kwan and Sarah L. Billington, (2003), “Unbonded Post-tensioned Concrete Bridge Piers.

I: Monotonic and Cyclic Analyses”, Journal of Bridge Engineering, Vol. 8, No. 2, March 2003. [4] Tetsuhiko Aoki, K. A. S. Susantha, (2005) “Seismic Performance of Rectangular-Shaped Steel Piers

under Cyclic Loading” Journal of Structural Engineering, Vol. 131, No. 2, February 1, 2005. [5] Sri Sritharan, Justin Vander Werff, Robert E. Abendroth, Wagdy G. Wassef and Lowell F.

Greimann, (2005), “Seismic Behavior of a Concrete/Steel Integral Bridge Pier System”, Journal of Structural Engineering, Vol. 131, No. 7, July 2005.

[6] Y. Edward Zhua, (2006), “Assessment of Bridge Remaining Fatigue life through Field Strain

Measurement” Journal of Bridge Engineering, Vol. 11, No. 6, November 2006. [7] Mucip Tapan, Riyad S. Aboutaha, (2008), “Strength Evaluation of Deteriorated RC Bridge

Column” Journal of Bridge Engineering, Vol. 13, No. 3 June 2008. [8] Indian Standard IS456-2000 “Plain and Reinforced Concrete Code of Practice”. [9] Indian Road Congress IRC: 6-2000 “Standard Specification and Code of Practice for Road

Bridges”.

81



Transport of contaminants through subsurface and its

modeling Muskan Mayank

1, P. K. Sharma

2 and Himanshu Sharma

3

1Trainee Teacher, Department of Civil Engineering, NIT Uttarakhand-246174,India

2Associate Professor, Department of Civil Engineering, IIT Roorkee-247667,India

3Assistant Professor, Department of Civil Engineering, NIT Uttarakhand-246174,India

Email: [email protected]

ABSTRACT

As contaminated ground water causes hazards to public health through the spread of disease, the practice

of groundwater remediation has been developed to address these issues. The transport of contaminants

through the subsurface is affected by the preferential flow of water and solute particles through the soil.

These involves the study of various physical non-equilibrium flow models for water movement and chemical

non-equilibrium for solute transport within the soil media. The transport process may be considered with the

effect of hysteresis in retention curve and conductivity for better understanding. This paper discusses

different models of contaminant transport and corresponding results were obtained using HYDRUS-1D

software. The solute concentration profiles have been presented using uniform and mobile-immobile

transport models. The modeling results indicate behavior of contaminants transport through BTC and earlier

arrival of solute in porous media. Higher values of mass transfer coefficient lead to reduce the solute

concentration and also higher value of sorption coefficient retards the solute.

Keywords: preferential flow, hysteresis, mobile-immobile transport models

1. INTRODUCTION

The unsaturated zone contains partially air as well as water in their pores and is rich in clay or organic

matter, promoting sorption, biological degradation and transformation of contaminants. The contaminants

transfer through the vadose zone is complex and there prediction is difficult. Various contaminants get reacts

with soil sediments, and other geologic materials and they usually travel with different path flow formation

and at velocity of variable magnitude. The various factors that affects the contaminant transport in the

unsaturated zone is listed in Table 1. A detailed mechanism of solute and water transport in the subsurface

and their simulations through HYDRUS-1D is required. This program solves by numerical method, the

Richards' equation for variably saturated water flow and that in case for heat and solute transport, the Fickian

advection-dispersion type equations is used. There is a sink term in the flow equation, which considers for

root water uptake by plants. In liquid phase flow process, the transport equations for solute consider

advective-dispersive process, and diffusive type transport process is considered for gaseous phase. As

contaminated ground water causes health hazards, the practice of groundwater remediation has been

developed to address these issues. The simulation by HYDRUS of various models involving single and dual

porosity or permeability models with consideration of hysteresis in retention curve and conductivity for the

flow to be transient or non-transient depending upon the nature of simulation, is performed.

The issue of groundwater contamination is a serious concern as the concentration of solute is increasing day-

by-day depending upon the source activities which is difficult to control. The drinking wells is also become

contaminated, so we have to take measures to keep some clearance to the drinking wells from the source of

contamination. These polluted environment under the subsurface zone also causes various health hazards.

Therefore, the complete knowledge of the transport process and its flow mechanism for unsaturated zone is

required to control and reduce somehow the effect of concentration. In this paper, modeling of solute

transport and behavior of solute concentration profiles using HYDRUS software is presented.

82



Table 1 Factors affecting transport of contaminants in unsaturated zone.

Factors Survival Influence Migration Influence

Temperature Long term survival at lower

temperature -

Moisture Content Moist soils have longer retention

period

Migration increases under the

saturated condition

Soil Properties Adsorption affects inflences on

survival

Migration process retards for

clayey soils.

Aggregation of particles Enhances Survival Slow movement.

2. BASIC CONCEPTS OF SOLUTE TRANSPORT MODELING

As the movement of solutes is associated with the movement of water fluxes in soils, the need of detailed

analysis of transport mechanism and contaminant behavior with respect to time must first consider and

evaluate with the water fluxes through the subsurface zone.

The equations that governs partially saturated water flow in subsurface are based on the basic Richards

equation, which utilizes uniform flow process and is combines with the Darcy–Buckingham equation for the

flux transport with the mass balance equation in their general form. The Richards equation have been

macroscopically developed by considering variability in space of hydraulic properties of soil , for example,

soil horizon having variability in lateral directions. The generalized mixed-form of Richard’s equation can be

expressed as;

Sz

hhK

zt

h

])1)[((

)( (1)

2.1 Specific models for water flow

Uniform flow model Richard’s Equation is the basis for solving the numerical models for water flow. The

solution requires better understanding of the soil hydraulic functions which results from the retention curve,

the variation of water content with the unsaturated soil hydraulic conductivity function (Simunek and

Genuchtan, 2008).

Sz

hhK

zt

h

])1)[((

)( (2)

where, K represents hydraulic conductivity, [LT-1

], h represents pressure head, [L], represents moisture

content [L3 L

-3], S represents the Source/Sink term [T

-1]

83



Fig 1: Conceptual physical Non-equilibrium models showing transport process (Ref: Simunek & Van Genuchtan, 2008)

(θ = water content in uniform model , θmo and θim = respective mobile/immobile water contents for mim or

dp model, θm and θf = corresponds to matrix/fracture water content for dpb model)[Ref: Simunek & Van

Genuchtan,2008]

Dual-Porosity Model This model has two different regions of flow and transport models as given by (van

Genuchten et. al, 1976) which divides the liquid phase into mobile (movable, inter-aggregate), that dissolved

solute and water transport and immobile (static, intra-aggregate), regions.

immo (3)

wmomo

mo

mo

momo hSz

hhK

zt

h

)(])1)[((

)( (4)

wimim

imim hSt

h

)(

)( (5)

where, K represents hydraulic conductivity, [LT-1

], h represents pressure head,[L],immo , represents

moisture contents in mobile and immobile regions [L3 L

-3], S represents the Source/Sink term [T

-1],

w

represents water transfer rate between inter and intra-aggregate pore domains.

Dual-Permeability Model The flow equations in this model is explained by the fracture (f ) in the macro pore

region and for transport due to inter particle pores and matrix (m) for intra particle pore systems (Simunek &

Genuchtan, 2008).

mfmf )1( (6)

w

ff

f

ff

ffhS

z

hhK

zt

h

)(])1)[((

)( (7)

1)(])1)[((

)( w

mm

m

mm

mm hSz

hhK

zt

h (8)

Where, mf , represent water content in fracture and matrix domain [L3 L

-3], represents the ratio of

volumes of the macro pore or fracture domain and the total soil system[L3 L

-3].

Fig.2. Conceptual physical nonequilibrium models for water flow and solute transport (Ref: Simunek & Van Genuchtan, 2008)

2.2.1 Specific models for solute transport

The advection–dispersion type equation helps to describe the solute transfer process (Simunek & Genuchtan,

2008);

z

qc

z

cD

zt

s

t

c )(

(9)

84



Where, c is the solute concentration [ML-3

], s is the sorbed concentrations [MM-1

], is the bulk density of

soil [ML-3

], D is the dispersion coefficient [L2T

-1], q is the volumetric flux density [LT

-1], is source/sink

term [ML-3

T-1

].

Physical Nonequilibrium: Transport Models

The linear form of the adsorption equation is considered in which the sorbed concentration is distributed

proportionally to the solute concentration.

cKs d (10)

Where, sis the sorbed concentration [M M-1

], c is the solute concentration [ML-3

],dK is the distribution

coefficient[L3M

-1]

For the Dual-Porosity Model,

mocc *, if Γw> 0 otherwise it equals cim for Γw< 0. (11)

smo

momomo

momo

mo

mo

momo

z

cq

z

cD

zt

sf

t

c

)(

)( (12)

ims

im

mo

imim

t

sf

t

c

)1(

)( (13)

*)( ccc wimmos (14)

where, immo cc , represents solute concentrations in mobile/immobile regions,[ML

3],

immo , represents

source or sink term,[ML-3

T-1

],s represents solute mass transfer rate between inter and intra-aggregate pore

domains [ML-3

T-1

].

For the Dual Permeability Model, analogous to the flow model of water, the dual-permeability model form

for solute transport is described by standard advection–dispersion type equations for both matrix or fracture

transport regions (Simunek & Genuchtan, 2008).

s

f

fff

ff

fff

z

cq

z

cD

zt

s

t

c

)( (15)

1)(

)( s

m

mmm

mm

mmm

z

cq

z

cD

zt

s

t

c (16)

*)()1( ccc wmfmdps (17)

Where, f and m represents the respective terms in fracture and matrix regions, dp represents the solute mass

transfer coefficient for dual-permeability system. [T-1

]

3. MODELING OF SOLUTE TRANSPORT USING HYDRUS-1D

HYDRUS-1D is a software package helps to simulate flow of water, heat, and solute in case of one-

dimensional variably saturated porous media. It also used to determine results of carbon dioxide and major

ion solute transfer coming under UNSATCHEM. Basically, the Richards equation used for variably-

saturated water flow and advection-dispersion type equations (CDE) for heat and solute transport are solved

numerically. Several modification are implemented to flow equation to for changing in the properties of soil,

a sink term is incorporated to account for root uptake. The involved process data processing, soil profile

discretization and graphical presentation of the results are all presented by a graphical-based interface in

results of HYDRUS modelling. The program can deal with different water flow and solutes transport

boundary conditions as per the desired conditions. (Šimůnek, et al., 2008).

Basically, the program is equipped with a project manager and data for pre-processing and post

processing. The Unsaturated Soil Hydraulic Properties are studied and implemented by using van Genuchten

[1980], Brooks and Corey [1964] and that by modified VG-M type analytical functions. Sevearal

modifications were made to improve the description of hydraulic properties near saturation. The wetting and

drying hysteresis effect is also incorporated by using the empirical model introduced by Scott et al. [1993]

85



and Kool and Parker [1987]. This model assumes that initial scanned curves for drying condition measures

from the main drying curve, and that for wetting curves from the main wetting type curve.

3.1. Modeling solute transport for non-equilibrium dual permeability model in a heterogeneous soil

The water flow and solute transport modeling by considering non equilibrium dual permeability model in a

heterogeneous soil having two different materials is performed using HYDRUS-1D. The governing equation

used, initial and final boundary conditions taken, selected soil hydraulic and transport parameters and the

results obtained from post-processing are shown in this section;

Governing Equations:

The flow equations for the macropore fracture (subscript f) and matrix (subscript m) pore systems in their

approach are given by;

+ + (18)

(19)

(20)

For the Dual Permeability Model, analogous to the water flow model, the dual-permeability formulation for

solute transport is based on advection–dispersion type equations for transport in both the fracture and matrix

regions.

(21)

(22)

+ (23)

c* is equal to cmo for Γw > 0 and cim for Γw < 0.

86



Fig 3: Material distribution and observation points by considering typical heterogeneous soil showing initial and

boundary conditions for water flow & solute transport for a dual permeability model

Soil Hydraulic Parameters:

Assuming, the default value of soil parameters as:

Material 1:

= 0.078, =0.43, α= 0.036 cm-1

, n= 1.56, = 0.0002889 cm/sec, I (Tortuisity Factor)= 0.5 , = 0,

= 0.8, 0.08 cm-1

, n= 2.00, = 0.0289, ω= 10-5

a= 0.1 cm 2.88*10-7

Material 2: [Ref: Ahmet Karagunduz et. al., wrr, 2015]

= 0.035, =0.34, α= 0.024 cm-1

, n= 2.99, = 0.0007689 cm/sec, I (Tortuisity Factor)= 0.5, = 0,

= 0.8, 0.08 cm-1

, n= 2.00, = 0.0289, ω= 10-5

a= 0.1 cm 2.88*10-7

Transport Parameters:

= 1.5 g/cm3

D = 10 cm2/s

Results from Post-Processing:

87



Table 2: List of some variables to be used in different models (Ref: Simunek & Genuchtan, 2008).

Model Variables Definition

Uniform model

Water Content

Pressure Head H

Resident concentration C

Flux Concentration

dz

c

q

Dc

Total solute mass (uniform transport) )( dKc

Dual-porosity

Water Content

Pressure Head

Resident concentration

Mobile zone flux Concentration

z

c

q

Dc mo

mo

momo

mo

Dual-permeability

Water Content

Pressure Head

Fracture concentration

Flux concentration

CONCLUSION

Behaviour of solute concentration profiles along with the inter-relation amongst the various hydraulic

functions are plotted as obtained through HYDRUS-1D modeling results.

The dual-permeability modeling result shows that the water in the fracture domain reached full

saturation quickly under high intensity precipitations and in the matrix there is gradual increase in the water

content as the time proceeds. The closure of the fracture is much more significant for low-intensity rainfall

and causing delay in the effluent solute arrival.

The dual-permeability model is demonstrated with two different soil material layers for a 100-cm deep

soil profile. The pressure head of –150 cm is set to be initial condition. The transfer of water mass is

assumed to be proportional to the effective saturation gradient in the two domains, and the mass transfer

constant ω set at 0.00001 s-1

. The higher values of mass transfer coefficient lead to reduce the solute

concentration and also higher value of sorption coefficient retards the solute. For simplicity, we consider the

precipitation as the time-variable boundary condition with having the tortuisity factor value of 0.5. While the

88



case for ponded surface, water in the fracture domain reached full saturation quickly, and there is gradual

increase in the water content in the matrix as time proceeds. Consequently, the total moisture content, which

becomes the contribution of the water contents of both matrix and fracture domains, will also shows gradual

increase.

REFERENCES

[1] Brooks, R.H., and Corey, A.T. (1964), Hydraulic properties of porous media: Hydrology Papers, Colorado State

University, 24 p.

[2] Kool, J. B., Parker, J. C., & Van Genuchten, M. T. (1987). Parameter estimation for unsaturated flow and transport

models—A review. Journal of hydrology, 91(3-4), 255-293.

[3] Quisenberry, V. L., Smith, B. R., Phillips, R. E., Scott, H. D., & Nortcliff, S. (1993). A soil classification system for

describing water and chemical transport. Soil Science, 156(5), 306-315.

[4] Šimůnek, J., & van Genuchten, M. T. (2008). Modeling nonequilibrium flow and transport processes using

HYDRUS. Vadose Zone Journal, 7(2), 782-797.

[5] Van Genuchten, M. T. (1980). A closed-form equation for predicting the hydraulic conductivity of unsaturated

soils. Soil science society of America journal, 44(5), 892-898.

[6] Van Genuchten, M. T., & Wierenga, P. J. (1976). Mass transfer studies in sorbing porous media I. Analytical

solutions. Soil Science Society of America Journal, 40(4), 473-480.

Improvement in Properties of Silt Soil using Egg-Shell Powder

Vaangmayaa Singh 1 and V.K. Arora 2 1PG student, Department Of Civil Engineering, NIT Kurukshetra, Haryana

[email protected] 2Professor, Department Of Civil Engineering, NIT Kurukshetra, Haryana

[email protected]

ABSTRACT

Soil stabilization is the process of reinforcing soil with suitable materials to improve desired properties

of soft and weak deposits of soil. Proper investigation of soil profile beneath the proposed structure as well

as proper designing of structure on the basis of shear strength and settlement criteria is mandatory.

Normally stabilization of soil is carried out by expensive additives like lime, cement, bitumen etc. and hence

requires an economic alternative. There are huge stockpiles of industrial and domestic waste materials but

absence of an effective waste disposal. Egg-shell is a domestic waste material which is rich in lime (>90%

usually), calcium, protein and hence can effectively replace industrial lime as a stabilizer. Eggshells have

been already used for the stabilization of cohesion-less soils in Japan. In the present work, the suitability of

egg shell powder (ESP) as a possible stabilizing material to improve the properties of locally available silt

soil is analyzed by laboratory experiments. Soil samples are collected from Kurukshetra, Haryana and mixed

with eggshell powder in proportions of varying % of weight of dry soil. The laboratory tests are carried out

to determine the strength and index properties to study the behaviour of soil blended with eggshell powder.

From the Atterberg limit tests, it is observed that the ESP is able to control the liquid limit and so the

plasticity index of the soil. The Maximum dry density (MDD) and Optimum Moisture Content (OMC) values

are obtained by conducting Standard Proctor Compaction tests. The Unconfined Compression tests (UCS)

are conducted on silt soil samples with optimum % of egg shell powder and an increase in strength with

increase in ESP is observed. A better understanding of these characteristics will enhance the usage of egg

shell powder in geotechnical works, thereby making silt soil suitable for foundation purpose.

Key Words: Soil stabilization, Silt soil, Egg Shell powder, Unconfined compressive strength.

1. INTRODUCTION

Structural design on land depends mainly on foundation and surrounding soil properties. So a detailed

proper investigation of sub-soil strata is necessary. Many soil deposits are found to be unfit to bear the

required load. We can change the site location but improving the required properties of soil is a more



89



suitable and practical option. Soil stabilization is the process of improving soil by adding suitable additives

in the soil. Due to high population, the world has two major problems of land scarcity and waste disposal

leading to pollution. These can be tackled a bit by using civil engineering in an economic and eco-friendly

manner. Industrial and domestic wastes can be used for improving soil properties and hence replacing

traditional and expensive soil stabilizing agents like lime, cement etc. & reducing stockpiles of waste

products.

Silts are low or negligible plasticity fines having low strength and compressibility characteristics which

can cause severe damage to structure resting on it. They are initially unstable in the presence of moisture. So

these soils require stabilization before construction period to get desired properties.

Chicken Egg-Shell is a waste product from restaurants and poultry farms. Lime and Egg shells share

same chemical composition and hence can be used as a replacement for lime. Literature study has revealed

that Egg Shell Powder mainly contains CaO(90-99%) and also SiO2, Al2O3, Cr2O3, Cl, MnO and CuO.

In this study, we will check the effect of Egg-Shell Powder on silts and suitability of Egg-Shell Powder

(ESP) as soil stabilizer by performing various laboratory experiments such as Atterberg’s limit tests,

90



Standard Proctor Compaction tests and Unconfined Compressive Strength (UCS) tests on parent soil

and soil mixed with various percentages of Egg-Shell Powder.

2. LITERATURE REVIEW

Amu et al (2005) studied Egg-Shell powder (ESP) effect on the stabilizing potential of lime in an

expansive clay soil. Observations concluded that 7% lime stabilization effect was better than the stabilization

effect of 3% lime + 4% ESP. They also concluded that the combination of ESP and lime can be used where

high sub-grade performance is not needed.

Muthu Kumar and Tamilarasan (2014) studied the possible use of chicken eggshell waste as a suitable

soil stabilizing agent. Unconfined Compressive Strength tests were performed with and without delaying

compaction on soil samples. Addition of eggshell powder in soil sample led to significant increase in

unconfined compressive strength of soil. The maximum value of unconfined compressive strength was

observed at 3% eggshell powder - soil mix. The unconfined compressive strength was observed more in

delayed compaction case than without delayed compaction case.

Okonkwo, Odiong and Akpabio (2012) studied the effect of eggshell ash on the strength characteristics of

cement-stabilized lateritic soil. The increase in eggshell ash content led to increase in the Optimum Moisture

Content value but reduced the Maximum Dry Density value of the soil-cement-eggshell ash mix samples. The

increase in eggshell ash content also considerably increased the strength characteristics of the soil-cement-

eggshell ash mix up to 35%.

Arashet al (2012) investigated the egg shell powder effect on plasticity of clay and expansive soils. ESP

addition in expansive soil reduced the plasticity of soil. ESP addition in soil resulted in decrease of the liquid

limit of soil hence the decrease was sharper in the plasticity index.

Olarewaju et al (2011) studied effect of egg shell powder on lateritic soil. Compaction test result

indicated that both cement and egg-shell powder significantly increased maximum dry density and optimum

moisture content of the soil. Lateritic soil stabilized with 8% eggshell powder possessed same optimum

moisture content and maximum dry density properties as lateritic soil stabilized with 2% cement. California

Bearing Ratio (CBR) test results indicate that lateritic soil stabilized with 8% egg-shell compared favourably

with lateritic soil stabilized with 2% cement while compressive strength test results indicated that eggshell

powder possesses low binding property.

The study on the influence of Egg-shell Powder and Fly-Ash on Engineering Properties of Al-Umara

Soil, by Najwa Wasif Jassim (2012) indicated that as the percentage of fly–ash and egg-shell powder in the

soil increases, the value of reduction in the plasticity index in soil samples increased at different rates. From

the literature review, it can be concluded that Egg-Shell Powder as soil stabilizing agent increase the strength

properties of soil and reduce the plasticity index of the soil and hence improve the properties of soil. The soil

used in this study is low plasticity silt in nature which leads to low bearing strength problem. The plasticity

and compaction characteristics of soil-eggshell mix & strength characteristics of cured and uncured samples

of soil and Egg-Shell Powder mix are studied in this paper.

3. MATERIALS USED

3.1. Soil

Soil was collected locally from Kurukshetra region of Haryana. Sample was obtained from 4m depth

below ground surface. Soil sample was oven dried and stored in sacks at normal temperature. Soil is

identified as Silt of low plasticity (ML) as per IS 1498-1970. It’s light brown in colour. Its properties are

described in Table 1 as below:

91



Table 1: Properties of Soil

S. No. Properties Of Soil Test Results

1. % Silt 67

% Sand 25

% Clay 8

2. Specific Gravity 2.56

3. Liquid Limit (%) 25.48

4. Plastic Limit (%) 18.83

5. Plasticity index 6.64

6. Maximum Dry Density (g/cc) 2.007

7. Optimum Moisture Content (%) 12.43

3.2. EGG-SHELL POWDER

Chicken Egg shells were collected as a domestic waste material from poultry farms and restaurants. They

were washed and air-dried for 48 hours in sunlight. Egg shells were then grinded in laboratory heavy duty

grinder and sieved through IS 425 micron sieve. Specific gravity of Egg-Shell powder used is 1.34.

4. METHODOLOGY

a) Perform laboratory tests to determine index properties of parent soil b) Study on properties of soil mixed with varying percentages of Egg-Shell Powder by performing

following tests:

ATTERBERG LIMIT TESTS:

Liquid limit (LL) and Plastic limit (PL) tests of the sample were performed. Plasticity Index of

sample was evaluated.

STANDARD PROCTOR COMPACTION TEST:

Compaction characteristics i.e. Maximum Dry Density (MDD) and Optimum Moisture Content

(OMC) were observed from resulting compaction curve.

UNCONFINED COMPRESSION STRENGTH TEST:

Unconfined Compressive Strength (UCS) of samples was evaluated with no curing and 14 days

curing.

(All tests were performed by standard procedures as per IS 2720 Part III-VIII.)


Following results were obtained by performing various laboratory experiments on soil-eggshell mix

samples:

5.1 ATTERBERG LIMITS

The effect of Egg-Shell Powder on the plasticity of silts was studied by plastic limit and liquid limit tests.

Tests were performed on parent soil mixed with 2%, 4%, 6%, 8%, and 10% Egg Shell Powder. The observed

atterberg limits values (LL, PL) and plasticity index (PI) values are listed in Table 2.

Table 2: Variation of Atterberg’s limits and plasticity index with %ESP

%ESP LL (%) PL (%) PI (%)

0 25.48 18.83 6.64

2 22.01 16.20 5.81

4 20.31 15.33 4.98

6 19.44 15.12 4.32

8 19.18 16.27 2.91

10 16.68 14.65 2.03

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Fig 1: Variation of Atterberg Limits with %ESP Fig 2: Variation of Plasticity index with %ESP

It’s observed that liquid limit and plastic limit values are decreasing on increase in percentage Egg- shell

powder as shown below in Figure 1. Plasticity index value is also observed to decrease with increase in

percentage Egg shell powder as shown below in Figure 2.

So it can be concluded that plasticity characteristics gradually decrease with increase in % ESP. At 10%

egg shell powder, maximum reduction in PI as 69.43% is observed.

The reduction in plasticity of soil sample may be due to similar composition of egg shell powder and

lime,which also share the tendency to coarsen the particles and thus resulting in reduction in liquid limit and

hence plasticity index of the soil.

5.2 COMPACTION CHARACTERISTICS

Compation characteristics were observed by performing Standard Proctor Compaction tests on soil

samples mixed with 2%, 4%, 6%, 8%, 10% egg shell powder to plot compaction curve and thus determining

the Maximum Dry Density (MDD) and Optimum Moisture Content (OMC).

The observed maximum dry density and optimum moisture content values with increase in egg shell

powder are listed in Table 3.

Figure 3 and Figure 4 show the variation of MDD and OMC values with addition of varying %ESP in

soil. It’s observed that maximum dry density significantly drops initially at 2% ESP and altogether show an

decreasing trend with increase in % egg shell powder. The reduction in maximum dry density values may be

due to comparatively lesser specific gravity of egg shell powder than soil.

The OMC value increase significantly at 2% ESP and show an increasing trend with increase in

percentage of egg shell powder. Hence it’s observed that addition of egg shell powder in silt soil result in an

increase of optimum moisture content and a decrease in maximum dry density of soil.

Table 3: Variation of Maximum Dry Density and Optimum Moisture Content with %ESP

%ESP MDD (g/cc) OMC (%)

0 2.007 12.43

2 1.942 14.32

4 1.955 11.89

6 1.976 12.02

8 1.941 13.67

10 1.928 15.12

7

6

5

4

3

2

1

0

0 2 4 6 8 10

% ESP

PLA

STIC

ITY

IN

DEX

(%

)

LIQUID LIMIT (%)

PLASTIC LIMIT (%)

30

2

5

2

0

1

5

1

0

5

0

0 2 4 6 8 10

% ESP

93



Fig 3: Variation of Maximum Dry Density with %ESP Fig 4: Variation of Optimum Moisture Content with %ESP

5.3 UNCONFINED COMPRESSIVE STRENGTH (UCS) TESTS

UCS tests were performed to find unconfined compressive strength of soil mixed with various

percentages of Egg shell powder. UCS tests were performed on both samples without curing and after 14 days

curing. The soil samples were prepared mixing with 2%, 4%, 6%, 8% egg shell powder at their respective

optimum moisture content and maximum dry density.

Soil was mixed with optimum quantity of water and kept for 24 hours in an airtight condition for moisture

equilibrium. Samples of required wet soil as per MDD value were compacted in a split mould to diameter 38

mm and length 76 mm. The samples were also kept in vacuum desiccators for 14 days for enhanced chemical

reactions.

Table 4 and Figure 5 show the angle of failure plane and unconfined compressive strength observed for

both uncured and cured samples with varying proportion of egg shell powder. The addition of 2%, 4%, 6%,

8% egg shell powder result in increase of UCS values of soil by 15%, 25%, 61%, 88% respectively.

Fig 5: Variation of Unconfined Compressive Strength with % ESP



Table 4: Variation of Unconfined compressive strength and angle of failure plane with %ESP

%ESP UCS (kg/cm2 ) (no curing) UCS (kg/cm2) (14 days curing) Angle of failure plane(degree)

0 0.80 0.80 64.54

2 0.92 1.68 63.43

4 1.00 2.80 58.11

6 1.28 3.68 56.32

8 1.52 4.08 53.13

Study shows that curing has a positive effect on UCS values of soil and increases it by two to three times

than that of samples without curing and about three to four times than that of parent soil.

6. CONCLUSIONS

Following conclusions can be made from above discussion:

a) A decreasing trend is observed in value of maximum dry density with increase in percentage of egg-

shell powder. This may be due to low specific gravity of egg-shell powder.

b) An increasing trend is observed in value of optimum moisture content with increase in percentage of

egg-shell powder.

c) A decrease is observed in value of liquid limit and plastic limit with increase in percentage of eggshell

powder. This may be due to water repellent behaviour of egg-shell powder.

d) A significant drop in plasticity of soil occurred when mixed with egg-shell powder. Hence it’s concluded

that egg shell powder considerably effect liquid limit and hence the plasticity of soil.

e) A significant increase in unconfined compressive strength of soil is observed with increase in eggshell

powder proportion.

f) A drop in angle of failure plane of soil to horizontal is observed with increase in egg-shell powder in

soil.

g) By review of literature it’s concluded that egg shell powder is comparatively better soil stabilizer than

fly-ash in case of silt soil. This may be due to small particle size of fly-ash.

h) So it’s concluded that Egg-Shell Powder can be used as a soil stabilizing agent for silt as it can increase

its strength by three to four times. It may be used as a replacement of lime for silts.

REFERENCES

[1] Olarewaju, A.J., Balogun, M.O. and Akinlolu, S.O., (2011), Suitability of Eggshell Stabilized Lateritic Soil as Subgrade Material for Road Construction, Electronic Journal of Geotechnical Engineering, 16, Bund.H, pp.899-908.

[2] Muthu Kumar M, Tamilarasan V.S, (2014), Effect of Eggshell Powder in the Index and Engineering Properties of Soil, International Journal of Engineering Trends and Technology, 11(7), pp.319-321

[3] Amu, O.O., (2005), Effect of Egg Shell Powder on the Stabilizing Potential of Lime on an Expansive Clay Soil, Research Journal of Agriculture and Biological Sciences, 1(1), pp.80-84.

[4] Okonkwo, U.N., Odiong, I.C and Akpabio, E.E. (2012), The effects of eggshell ash on strength properties of cement stabilized lateritic, International journal of sustainable construction engineering and technology, 3(1), pp.18-25.

[5] Anu Paul, Anumol V.S, Fathima Moideen, Jiksymol K Jose &Alka Abraham, (2014), Studies on Improvement of Clayey Soil Using Egg Shell Powder and Quarry Dust, International Journal of Engineering Research and Applications, 4(4), pp.55-63

[6] Najwa Wasif Jassim., (2012), Influences of Fly-Ash and Eggshell Powder on Some of Engineering Properties of Al-Umara Soil, Journal of Engineering and Development, 16(2), pp.211-219.

[7] ArashBarazesh, Hamidreza Saba, Mehdi Gharib&MoustafaYousefi Rad, (2012), Laboratory Investigation of the effect of eggshell powder on plasticity index in clay and expansive soil, European Journal of Experimental Biology, 2(6), pp.2378-2384.

95



A Case Study of Pedestrian-Vehicle Conflict at Midblock

Crosswalk in Srinagar, Uttarakhand Shivangni Khandelwal

1, Apoorv Prasad

2, Abhinav Kumar

3

1. M.Tech student, Department of Civil Engineering, National Institute of Technology, Uttarakhand- 246174, Email ID – [email protected]

2. B.Tech student, Department of Civil Engineering, National Institute of Technology, Uttarakhand- 246174, Email ID – [email protected]

3. Trainee Teacher, Department of Civil Engineering, National Institute of Technology, Uttarakhand- 246174, Email ID – [email protected]

ABSTRACT

Pedestrians at uncontrolled midblock crossing locations in the mixed traffic conditions face serious

threat for conflict with vehicles. Due to increase in motor vehicle growth there is an increase in the

regulation of motor vehicles only and the regulation of pedestrian movement is completely neglected.

Pedestrian-vehicle conflict is still an open research topic in the traffic safety and planning. This paper deals

with the pedestrian-vehicle conflict at midblock crosswalk in Srinagar (Garhwal), Uttarakhand on NH-58

which is a two lane single carriageway road. It is an important route connecting Chota Char Dham including

Kedarnath Temple and Badrinath Temple. Buses and Vehicles packed with pilgrims throng the highway

during pilgrim season. It doesn’t have proper facilities for pedestrians to walk so pedestrians are forced to

use the carriageway for their movement, also there are no proper signage and markings indicating speed

limits for vehicles which increases the possibility of conflict when pedestrians cross the road.

The study aims to investigate pedestrian related safety aspects by estimating Post Encroachment Time

(PET) and waiting time for pedestrian during crossing. The data collected from videography survey at two

locations in Srinagar (Garhwal) are examined.

Keyword: conflict, pedestrian safety, Post Encroachment Time, mid-block crosswalk

1. INTRODUCTION

Mixed traffic flows are becoming more common in urban areas all over the world, especially in

developing countries such as India. In mixed traffic flow, motor vehicles, non-motorized vehicles (such as

bicycles and tricycles), and pedestrians share the same facilities (roads and intersections), and therefore

vehicle-vehicle conflicts, bicycle-vehicle conflicts, and pedestrian-vehicle conflicts frequently occur. Many

papers in the literature have defined traffic conflict. For example, Zheng, Ismail, and Meng (2014) pointed

out that almost all operational definitions of traffic conflict can be grouped into two types: those based on

evasive actions, and those based on temporal (and/or spatial) proximity. A representative definition of

evasive action-based traffic conflict is ‘‘an event involving two or more road users, in which the action of

one user causes the other user to make an evasive maneuver to avoid a collision’’ (Parker & Zegeer, 1989).

According to the literature, a pedestrian-vehicle conflict occurs if the oncoming vehicle has to brake

abruptly, if the vehicle has to swerve to avoid colliding with the pedestrian, or if the pedestrian has to take

sudden evasive action, such as jumping back to avoid a collision. This definition is based on evasive actions

taken either by the driver or by the pedestrian. A representative definition of proximity-based traffic conflict

is ‘‘an observable situation in which two or more road users approach each other in space and time to such

an extent that there is a risk of collision if their movements remain unchanged’’ (Amundsen, 1977). This

means that the closer the road users are to each other, either in time or in space, the nearer they are to a

collision. This is more of a conceptual (theoretical) definition, and it is operational because the time and

space parameters are quantitative and can be measured by traffic detectors.

Pedestrian-vehicle conflicts are hard to formulate because of the unpredictable behavior of both drivers

and pedestrians which depends on many uncertain factors. Traffic accidents involving pedestrians are a

common phenomenon in many cities (Li, 2014). Pedestrians are among the most vulnerable road users

(VRUs) because they lack the physical protection to reduce accident consequences (European Conference of

Ministers of Transport, 1998).

A number of published studies have dealt with pedestrian-vehicle conflict, but they were limited to

studying the factors influencing conflict, such as personal characteristics, traffic conditions, and

environmental factors at crosswalks. From the first perspective, personal characteristics like age, gender, and

disability have been studied. For example, Liu and Tung (Liu & Tung, 2014) found that elderly pedestrians

exposed themselves to higher risk of road crossing than young pedestrians due to their decline in walking

ability. Yagil (2000) found that men are less aware than women of their conflicts with vehicles when they

cross the street. Tom and Granié (2011) explored gender differences in pedestrian rule compliance both at

signalized and unsignalized crossroads. From the second perspective, traffic conditions such as traffic

volume and vehicle speed have also been studied. For example, Cheng (2013) proposed that higher vehicle




96



volume might lead to more serious pedestrian-vehicle conflict because pedestrians’ waiting time will

increase and exceed their tolerance limits; higher vehicle speed resulted in a higher collision probability

between pedestrians and vehicles. Cheng also modelled the relationships between pedestrian waiting time

and vehicle volume, pedestrian-vehicle conflict time, vehicle speed, traffic delay, and pedestrian volume.

Himanen and Kulmala (1988) analyzed 799 events of pedestrian-vehicle conflict; their results indicated that

the most important explanatory variables included pedestrian distance from the curb, city size, number of

pedestrians simultaneously crossing, vehicle speed, and vehicle platoon size. From the third perspective,

environmental factors such as city size, signal settings, road width, and lane definition have also been widely

studied. Traffic signals are the most important environmental factor because pedestrians and drivers should

obey traffic light restrictions at signalized intersections. At a non-signalized marked crosswalk, pedestrians

have the right-of-way according to traffic regulations, and the vehicles should give the right-of-way to the

pedestrians. However, the vehicles usually do not give right-of-way to the pedestrians and drivers are not

told when to leave the crosswalk (Troutbeck & Brilon, 1997), which makes crossing of a non-signalized

intersection more complex. Most of these studies are based on signalized intersections and road sections, and

therefore they cannot reflect the complex interpenetration between pedestrians and vehicles and evaluate

pedestrian safety at uncontrolled intersections.

PET is the time between the moment that the first road user leaves the potentially occupied conflict

zone and the moment the second road user reaches it. Usually, conflicts between pedestrians and vehicles are

divided into discrete severity levels according to different thresholds of PET. Higher PET values indicate

lower severity. Malkhamah used vehicle deceleration to divide the severity of conflicts into three different

levels: serious, slight, and potential conflict (Malkhamah, Tight, & Montgomery, 2005). In addition, for

lane-based pedestrian-vehicle conflict, severity is divided into three categories: serious, slight, and potential

conflict, according to the PET indicator. Archer (2005) indicated that PET is useful for measuring critical

events where crossing trajectories for road users are involved. PET calculation requires capturing a static

conflict point rather than a dynamic point. In cases where a driver performs evasive action, the potential

conflict point will be dynamic.

This paper quantifies pedestrian-vehicle conflicts over a non-signalized marked crosswalk using the

proposed safety indicator, Post encroachment time (PET). Also evaluating the influence of pedestrian

waiting time on pedestrian-vehicle conflict.

2. METHODOLOGY

2.1 Study area and data

Two non-signalized marked crosswalk area with bidirectional traffic, was chosen as the study area

shown in Fig.(1) and Fig.(2). The size of the marked crosswalk area is approximately 7.5m×3m.

This study used camera to record the marked crosswalk area for 120min between 10 a.m. and 12 a.m. on

a weekday. Camera was placed on the roof of an office building. There were 890 cars (29.02%), 91 buses

(2.97%),1920 motorcycles (62.60%), 56 LCV (1.83%), 666 pedestrians and 30 other types of vehicles

(0.98%). Their locations were recorded according to the time tags in the video frames.

Fig 1.The study area (Site-1) Fig 2. The study area (Site-2)

97



Fig 3. Vehicular composition

2.2 Data extraction

Based on the collected videos, this study identified the movements of vehicles and pedestrians in the

video by manual means. The videos were run at a frame rate of 25 fps in Kinovea Software and different

parameters were identified.

Fig.3 Grid placement in Kinovea software

Waiting time of the pedestrian is defined as the time difference between the arrival time of pedestrians

and their departure from the curb of the lane. The average waiting time of individual pedestrians was

recorded for entire study hours for the pedestrians of both sides.

For Post encroachment time (PET) calculation, the crossing area including crosswalk and some regions

adjacent to the crosswalk was considered as the conflict area. It was divided into conflict zones of square

grids of size 2.5 m x 2.5 m and was placed in the video with the help of Kinovea Software (Figure 3).

2.3 Estimation of PET between pedestrian and vehicles

At a non-signalized crosswalk, pedestrians have to cross lane by lane without guidance from any special

signaling facilities to arrive at the opposite side of the road. Theoretically, for each pedestrian during his/her

crossing, the conflict zone is a ‘‘common area used by road-users/vehicles approaching from different

trajectories’’ (Archer, 2005). In this zone, the pedestrians are exposed to the risk of oncoming vehicles.

Archer pointed out that PET has good performance in analyzing conflicts during pedestrian street-

crossing behavior.

98



For each pedestrian P = 1, 2, 3,…… N, the time of his/her approach to the conflict zone was recorded as

T0 and the time of his/her leaving the conflict zone as T1 (as shown in Fig. 4, where the pedestrian is leaving

the conflict zone). Therefore, the time for a pedestrian to cross the conflict zone can be calculated as:

∆TP = T1 – T0 (1)

For each vehicle, the time of its approach to the conflict zone was recorded as T2 (as illustrated in Fig. 4,

where the vehicle is just approaching the conflict zone). Therefore, when a pedestrian approached the

conflict zone, the time for the vehicle to reach the conflict zone could be calculated as:

∆TV = T2 – T0 (2)

Then PET can be calculated as:

PET = ∆TV - ∆TP = T2 - T1 (3)

PET is the time difference between when the pedestrian leaves the conflict zone and when the vehicle

approaches the conflict zone.

The PET value can be negative; if a vehicle passes the conflict zone before the pedestrian has finished

crossing the lane, ∆TP will be greater than ∆TV. In this situation, the pedestrian has passed the vehicle, but

has not yet finished crossing the lane.

Fig 4.Procedure analysis for PET calculation

To identify traffic conflict severity, threshold values should be established. These values vary across

studies (Zheng et al., 2014). In (Ismail, 2010), thresholds values varied from 1.0 s to 5.0 s. Peesapati, Hunter,

and Rodgers (2013) tested different thresholds varying from 1.0 s to 10 s and selected 1.0 s as the one that

produced the best results. This variation of threshold values may be caused by the heterogeneity imposed by

type of road, type of vehicle, involved road users, and weather on traffic conflicts (Svensson, 1998).

In this paper, PET threshold values were 1.0 s and 5.0 s, and descriptions of the different severity levels

of pedestrian-vehicle conflict are shown in Table 1.(according to Almodfer et al.)

Table 1 Description for different severity of pedestrian-vehicle conflict.

PET value in sec Severity conflict Description

PET ≤ 1 Serious conflict

In this situation, the pedestrian rushed to cross the lane in

the presence of a coming vehicle, and the vehicle is so near

to the pedestrian

1 > PET ≤ 5 Slight conflict

In this situation, the pedestrian crossed the street in the

presence of a coming vehicle, and the vehicle is far from the

pedestrian

PET ≥ 5 Potential conflict

In this situation, the pedestrian crossed the street in the

presence of a coming vehicle, and the vehicle is far enough

from the pedestrian

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3. STATISTICAL RESULTS

3.1 Number of conflict of different levels of severity over two direction

For 330 selected individual pedestrian, there were 666 crossing in total. 336 crossing occurred in the

absence of vehicles and therefore did not result in a conflict situation. Details about conflict number are

summarized in table 2.

Table 2 conflict number of different levels of severity over two directions

Direction Site Serious

conflict

Slight conflict Potential

conflict

Total

Near end

Site - 1 23 43 25 91

Site - 2 21 37 12 70

Subtotal 44 80 37 161

Far end

Site – 1 10 39 9 58

Site – 2 41 32 21 94

Subtotal 51 71 30 152

In general, conflicts occurred most frequently on site – 1(91) in near end, while for pedestrian in far end,

conflicts occurred most frequently on site – 2(94).

3.2 Percentage of serious conflict

The percentage of conflict occurrence, Pconflict is calculated as

PConflict =NSlight+NSerious

NCrossing (4)

Where Nslight is the number of slight conflicts, Nserious is the number of serious conflicts, and Ncrossing is the

total number of pedestrian crossings.

Accordingly, the percentage of serious conflict occurrence, Pserious can be calculated as:

PSerious = NSerious

NSlight+ NSerious

(5)

PSerious is an important parameter in conflict analysis because it represents the percentage of potentially

dangerous pedestrian-vehicle accidents. PSerious was 35.48%(near end) and 41.80%(far end) respectively, and

for all pedestrians Pserious was 38.62%.

Both pedestrians and drivers caused serious conflicts in far end.41.80% serious conflicts were recorded

in the far end.

Crossing without watching for oncoming vehicles is one aspect of unsafe pedestrian crossing behavior

(Yagil, 2000). 34.84% of serious conflicts happened in site -1 for pedestrians in near end, and 20.41% of

serious conflicts happened in site-1 for pedestrians in far end. These serious conflicts in near end occurred

when the pedestrian entered the crossing immediately before a vehicle approached; the driver decelerated,

and the pedestrian walked faster or started running. This is one of the most dangerous conflict situations

because the driver does not anticipate the presence of pedestrians.

3.3 Influence of waiting time on pedestrian vehicle conflict

Pedestrian waiting time reflects pedestrian delay at intersections. Traditionally, urban network

management has paid more attention to minimizing vehicle delays and little attention to delays encountered

by pedestrians (Vallyon, Turner, & Hodgson, 2011). An in-depth understanding of the influence of waiting

time on pedestrian-vehicle conflict would alert designers, planners, and managers to pedestrians’ needs, thus

making urban areas more pedestrian-friendly (Vallyon et al., 2011).

Waiting time is the time difference between the pedestrian’s arrival time and the pedestrian’s departure

time at the same crossing area. Generally, pedestrians preferred crossing actively rather than waiting

passively (Zhuang & Wu, 2011). However, waiting time can significantly deter people from walking or can

lead to unsafe crossing behavior (Vallyon et al.,2011).

100



Table 3 The number of serious, slight and potential pedestrian-vehicle conflict by waiting time

Waiting time (s) less than 3sec 3 – 30 sec More than 30 sec

Serious conflict 54 24 5

Slight conflict 103 56 11

Potential conflict 43 14 3

Total 200 94 19

On the basis of data from Table 3, waiting time less than 3 sec caused 200 conflicts. The number of

conflicts decreased as waiting time increased. It is worth noting that the number of serious conflicts

decreased significantly when the waiting time went from 3 sec to 30 sec. There were also five pedestrians

who crossed in very risky situations after waiting for more than 30 sec. similar results were suggested in

Martin (2006) and Li (2014).

CONCLUSION

This study evaluating pedestrian – vehicle conflicts at a non-signalized marked crosswalk in Sringar,

Uttarakhand. Pedestrian – vehicle study classify conflict in three different levels: serious, slight and potential

conflicts, based on PET threshold value. Based on above evaluation, conclusion were drawn which is as

follows:

1. The severity of pedestrian – vehicle conflict was evaluated. Result showed that the far end recorded a

higher percentage of serious conflict than the near end, the slight conflict were the most frequently

occurring conflicts for both directions.

2. The relationship between pedestrian waiting time and pedestrian-vehicle conflict was also investigated.

Analytical results showed that shorter waiting time caused 200 conflict situations between pedestrians

and vehicles. As pedestrian waiting time went from 3 to 30 sec, serious conflict decreased significantly.

When pedestrian waiting time went beyond 30sec, five pedestrian crossed in very risky situations.

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19. Zheng, Lai, Karim Ismail, and Xianghai Meng. "Traffic conflict techniques for road safety analysis: open questions

and some insights." Canadian journal of civil engineering 41.7 (2014): 633-641

20. Zhuang, Xiangling, and Changxu Wu. "Pedestrians’ crossing behaviors and safety at unmarked roadway in

China." Accident analysis & prevention 43.6 (2011): 1927-1936.

102



Morphometeric Analysis Of Kaushalya River Basin (Haryana)

Khushbu Choudhary 1, Baldev Setia

2

1 PG student, Department Of Civil Engineering, NIT Kurukshetra, Haryana(136119), [email protected]

2Professor, Department Of Civil Engineering, NIT Kurukshetra, Haryana(136119), [email protected]

ABSTRACT

Watershed Management study is essential not only in understanding a particular terrain but also for

sustainable utilization of the natural resources of that particular area. GIS has proven to be one of the most

efficient tools in this watershed analysis. Present study deals with the study of the geometry of a basin by

calculation of the various parameters like stream order (U), stream length (Lu), bifurcation ratio (Rb),

drainage density (D), stream frequency (Fs), texture ratio (T), elongation ratio (Re), circulatory ratio (Rc),

form factor ratio (Rf) etc. For this study - River Kaushalya, geographically located at - longitudes 77°05'15"

E and latitude 30°41'14" N, has been considered. ArcGIS (ver 10.3) has been used for estimation of these

morphometric parameters. The river covers a course of approximately 20 km through the Mourni Hills of

Panchkula district of Haryana. The GIS based morphometric analysis has revealed the Kaushalaya basin to

be of dendritic nature indicating lack of structural control. Aspect and slope map that has been prepared for

the watershed, gives an insight of the terrain morphology. This study would greatly help in understanding

the drainage of watershed and in proper planning of the region.

Key Words: DEM, Morphometeric Analysis, ,GIS, River Basin

1. INTRODUCTION

Morphometry is defined as analysis and measurement of the configuration of the earth’s surface and

dimension and shape of its land forms. Detection and calculation of morphometric parameters of any

watershed gives description about the hydrologic aspect which gives detail about the geology of the area.

Horton, Thornbury, Strahler are the great researchers who investigated the various drainage parameters

which proves to be very helpful in understanding the geomorphological, lithological, structural and other

aspects of a watreshed.While carrying out analysis through software such as (Arc GIS ,ERDAS, QGIS)

digital elevation data is required for generating the elevation model of a landscape to any extent. Thus a

detailed study of morphometeric analysis of a basin is of great help as it provides information about

influence of drainage morphometry on land forms and their characteristics. In this study, morphometric

analysis and prioritization of watershed are carried out for Kaushalya River Catchment in Panchkula district

of Haryana, India.

2. Study Area

The study area is located in panchkula district, of Haryana state of India. It is geographically located at

latitude 30˚41’14” and longitude 77˚05’15”.The Kaushalya river rises in the Shivalik hills on the border of

Haryana and Himachal Pradesh state and flows through Panchkula district and confluences with Ghaggar

river Pinjore just downstream of Kaushalya Dam. Basin is classified into two parts, khadir and Bangar, the

higher area that is not flooded in rainy season is called Bangar and low flood prone area is called Khadar.

103



Fig1 Map Of Study Area

3. MATERIAL AND METHOD

For deliniation of kaushalya river watershed, and for preparing the drainage map information regarding the

topography is needed. For this present paper ASTER GOLBAL DEM from USGS

(https://earthexplorer.usgs.gov) public data base of 30m resolution was downloaded for the Mourni hills

region, and then this DEM was used to delineate the watershed. Then this delineated watershed is processed

in Arc GIS 10.2.2 and following procedure is followed for morphometric analysis of kaushlay watershed is

as follows:-

1. Firstly, ASTER Global DEM was downloaded from USGS site and primary processing was performed.

2. Watershed of kaushlaya river is extracted from downloaded DEM using Arc GIS 10.2.2.

3. Extracted watershed is again processed in Arc GIS 10.2.2 for calculating the parameters like number and

length of streams of each different order, basin parameter then stream frequency, stream order, drainage

pattern, bifurcation ratio, circulation ratio, relief ratio were calculated.

Fig2: DEM of Kaushalya River Basin

https://earthexplorer.usgs.gov/

104



Fig3: Flow Accumulation Map of Kaushalya Watershed

FIG 4: Flow Direction Map

4 RESULT

Morphometric analysis of basin :

4.1 Linear Aspects:

The linear aspects of morphometric analysis of basin include stream order, stream length, , stream length

ratio and bifurcation ratio.

a) Stream Order (U):

The first step of geomorphological analysis of any drainage basin is designating the stream order based

on its ranking. Concept of stream order was introduced by Horton and latter it was modified by Strahler.

According to Strahler’s concept of stream ordering when two streams of 1st order meet it give rise to a

channel segment of 2nd

order, when two 2nd

order stream meet, it give rise to a channel segment of 3rd

order

and so on. Order of the basin is designated on the basis of the order of the highest stream. After analysis, it

was found that Kaushalya river catchment is of 5th oder and drainage is dentritic in nature.



b) Stream Length(Lt) and Stream Length Ratio(Rl):

According to Horton(1945), stream length refers to the total length of the stream in each order. Mean

stream length is the average length of the stream in each order and is calculated by dividing the total length

of all streams in each order and is calculated by dividing the total length of all streams in each order by the

number of streams in that particular order. Total length of the Kaushalya river basin 1st order is 129.152 km,

2nd

order 77.974 km, 3rd

order 21.037 km, 4th order 26.993 km, 5

th order 10.258 km.

Fig5: Stream Order of kaushalya Watershed

Fig 6: Graph Between Stream Order And No. of Stream

According to Horton, stream length ratio is the ratio of the mean stream length of a given order to the

mean stream length of the next lower order. Stream length ratio of kaushalya river basin is 0.603, 0.269,

1.283, 0.379. This variation indicates late youth stage of geomorphomic development which indicates an

important relationship with the run-off and erosional status of watershed.

c) Bifurcation Ratio(Rb):

Bifurcation ratio may be defined as the ratio of no. of stream of given order to the no. of stream next

higher order. Bifurcation ratio shows small range of variation for different region or for different

environment except in areas where geological control dominates. The Rb values of kaushalya river basin

varies from 2.394 to 1.117 with a mean Rb of 1.736

4.2 Aerial Aspect:

It includes Drainage Density, Texture Ratio, Stream Frequency, Form Factor, Circulation Ratio, Elongation

Ratio.

a) Drainage Density(Dd):

It indicates the closeness of spacing of channels. It is the average length of the channel per unit area of

the basin. Lower value of drainage density (dd) gives an indication that region is underlain by a highly

permeable material with vegetative cover and low relief. Kaushalya river basin has a Dd value of 0.758 which

indicate well drained basin.



b) Form Factor(Ff):

According to Horton (1932) form factor is a dimensionless ratio of area of basin (A) to the square of the

length of the basin (Lb). form factor should lie within the range of 0.1- 0.8.Smaller the value of form factor

more elongated will be the basin. If value of form factor is high then basin will have high peak flow for

shorter duration and if value of form factor is low then basin will will have low peak flow for longer

duration. Form factor value of Kaushalya river basin is 0.301 which indicates basin is elongated.

c) Circulation ratio(Rc):

According to Muller (1953) circulatory ratio (Rc) is defined as the ratio of the area of the basin to the area of

the circle having the same circumference as the perimeter of the basin. Circulatory ratio indicates the

dentritic stage of watershed. Low, high, medium values of Rc indicates the young, mature, and old stage of

life cycle of tributary watershed. Rc value of Kaushalya watershed is 0.429.

d) Elongation Ratio(Re):

Elongation ratio is defined as the ratio of diameter of a circle having the same area as that of the basin to

maximum length of the basin. Value of elongation ratio should lie in the range of 0.6- 0.8. Value close to 1,

indicates the region of very low relief and values of elongation ratio lying in the range of 0.6-0.8 indicates

high relief and steep slope. Elongation ratio of Kaushalya river basin is 0.619 which indicate high relief and

steep slope.

e) Stream Frequency(Fs ): Stream Frequency is the ratio of stream segment of all order to area of basin. It describe the no. of

streams per unit area. Stream frequency of kaushalya river basin is 1.108.

f) Texture Ratio (Tr ):

Tr is calculated as the ratio of total no. of stream of 1st order(N1) to the perimeter of the watershed.

Texture ratio of kaushalya watershed is 1.955.

Fig7: Aspect Map of Kaushalya Watershed

Fig 8: Slope Map of Kaushalya Watershed



5. CONCLUSION

In the present paper GIS technique has been used for the analysis of morphometric parameters of the

basin and it proved to be an efficient tool in drainage delineation. Bifurcation ratio, length ratio and stream

order of the basin indicated that kaushalay river basin is fifth order basin with dendritic type of drainage

pattern. Drainage density, texture ratio, showed that basin has moderate texture and it is almost elongated.

The complete morphometric analysis indicated that area has ground water prospect.

TABLE 1 : Morphometric Parametres

Sr. no. Parameter Value

1 Area 350.116

2 Perimeter (Km) 101.229

3 Basin order 5

4 Drainage density(Dd) (Km/Km2) 0.758

5 Stream frequency (Fs) (Km)2 1.108

6 Texture ratio(Tr) (Km) 1.955

7 Mean Bifurcation ratio (Rb ) 1.736

8 Form Factor (Rf) 0.301

9 Circulatory ratio (Rc) 0.429

10 Elongation Ratio (Re) 0.619

REFERENCES:

[1]Chaitanya B. Pande1, Kanak Moharir (2015),” GIS based quantitative morphometric analysis and its

consequences: a case study from Shanur River Basin,Maharashtra India”, springer

[2]Mohd Yousuf Khanday,Akram Javed(2017),” Hydrological investigations in the semi-arid Makhawanwatershed,

using morphometry”, springer.

[3]Karamat Ali1,2*, Roshan M. Bajracharya1, Bishal Kumar Sitaula3, Nani Raut1, Hriday Lal

Koirala4,(2017),”Morphometric Analysis of Gilgit River Basin inMountainous Region of Gilgit-BaltistanProvince”,

Northern Pakistan, Journal of Geoscience and Environment Protection.

[4]Abhishek Banerjee1, Prafull Singh1,Kamleshwar Pratap(2015),” Morphometric evaluation of Swarnrekha

watershed, MadhyaPradesh, India: an integrated GIS-based approach”, springer

[5] Shah, K.C., Pranay, R. Pali(2017),” Morphometric Characteristics of Sub-Watershed (P-17) in Paras Region, Akola

District, Maharashtra, India – using Remote Sensing & GIS”, International Journal of Advanced Earth Science and

Engineering.



Effect of Pervious Concrete in Sub-base/Base Coarse for

Highway Construction

Hitesh Patel1, Dr. Aditya Kumar Anupam

2

1M. Tech., Transportation Engineering, Department of Civil Engineering, NIT Uttarakhand,

[email protected] 2Assistant Professor, Department of Civil Engineering, NIT Uttarakhand, [email protected]

ABSTRACT In highway construction we require a Granular Sub-Base Coarse (GSB)/drainage layer to allow drainage

of water which is sometimes not as efficient in draining stormwater in tropical countries like India. So to

cope up with drainage issues of road payments, we tried to put pervious concrete to all new application and

use it as a replacement of both Granular Sub-base (GSB) Coarse and Dry Lean Concrete (DLC) Sub-Base

Coarse. Pervious concrete not only has better permeability and porosity but also gives better compressive

strength as compared to GSB Layer and DLC layer respectively. It is believed as an effective material for

controlling storm-water in an economical and environmental friendly way. Permeable concrete is normally

made of single-sized aggregate bounded together by Portland cement. Because of its insufficient structural

strength its application is restricted as a low traffic pavement material only, but we are aiming at developing

the pervious concrete with enhanced structural strength. Various mix designs at different aggregate

gradations, different cement content and different water cement ratios were attempted and their effects on the

compressive strength and permeability of pervious concrete were investigated in this research. To support

our study, we performed various tests to identify physical properties of aggregates. Also to determine the

feasibility of pervious concrete, compressive strength test and water permeability tests on all test samples

were performed. During our study, we found that the strength of pervious concrete did not satisfy the

strength requirements as per IRC: SP: 049 with the used cement type and adopted aggregate gradation.

Further studies will be done to obtain the appreciable strength of concrete to use it in sub base layer of road.

INTRODUCTION

Rapid urbanization is greatly affecting the hydrological characteristics of stormwater runoff. Over the

years various strategies, methodologies and planning were tried to counter the adverse effects of urbanization

and to replenish the easy and smooth recharge of stormwater into the natural water system. But not many of

them were quite efficient in solving the problem. One of those methods was the use of pervious concrete[1].

Pervious concrete or porous concrete or sometimes called green concrete is a macro-porous concrete that

was initially used as building material in 1800 in Europe as paving surface and load bearing walls but due to

lack of knowledge and construction technology advancement its use was discontinued. After World War II,

during 1950s it again gained popularity[2]. A lot of research work was carried out to use it as a construction

material. Recent studies have shown that it can be used as a successful payment layer (surface course) but

pervious concrete as a payment layer has some drawbacks which limit its application only in light and low

traffic areas, parking lots and pedestrian pavements[3].

Pervious Concrete Pavements allow stormwater to filter through the voids in its surface into the

underlying rock reservoir where it is temporarily stored and infiltrated into the surrounding materials[3].

Pervious concrete is same as conventional concrete in many aspects but it contains large amount of voids,

approximately 15% to 35% voids of the total volume of concrete[4]. In pervious concrete fine aggregates are

totally or partially eliminated from the design and the coarse aggregate gradation is kept very narrow so that

the aggregates can lock by the binding material. Cement paste is used as a binding material which is formed

by the mixture of water and cement to create a paste to form a thin coat around the aggregates so much that it

leaves the voids between the aggregates to allow easy flow of water. Water cement ratio is generally kept

between 0.25 and 0.30 with the addition of chemical admixtures. Pervious concrete with water cement ratio

0.30 to 0.40 has also been tested successfully. The relationships between pervious concrete strength and

various mix design considerations like water cement ratio, aggregate gradation, cement content and type,

effects of admixtures and fillers have not been studied and not much experiments and research work has been carried out on the pervious concrete, especially in Indian environmental and traffic conditions[5].

Fresh mixture of pervious concrete is a stiff mix with very low workability. It is a zero slump concrete

and slump value is kept less than 20 mm. These specifications produce a hardened concrete with

dramatically large amount of inter connected voids[6]. A typical value of pervious concrete permeability





ranges from 0.025 cm/s to 0.6 cm/sec[7]. The conventional design of the pervious pavement consists of a

pervious concrete pavement layer with a thickness of 100 mm (4 inches) to 150mm (6 inches), a permeable

base layer with a thickness up to 457mm (18inches), and a layer of permeable subgrade. If the subgrade

permeability is low, then drainage pipes can be installed to drain water, but drainage pipes makes the system

costly[8].

Pervious concrete has usually less compressive strength as compared to conventional concrete due to

presence of large amount of voids which limits its application in many areas majorly as a payment material

due to its very low strength. Also its use in cold climate regions is discouraged due to various stresses caused

due to freezing and thawing action. According to EPA, about 75% of pervious pavements have failed due to

wear and tear[2]. Pervious concrete strength hugely depends on the method of compaction used and the

compaction energy applied to the wet mix of concrete. Low compaction energy as compared to standard

method of compaction affects the pervious concrete properties by reducing compressive strength, split

strength, unit weight, and increasing permeability[9]. Being a brittle material, the behavior of pervious

concrete is influenced by the crack propagation, or fracture behavior. Under repeated traffic and

environmental loads, concrete pavements often fail under fatigue cracking. Detailed understanding of

fracture and fatigue behavior of pervious concrete can help to improve the pervious concrete design

procedures[10]. Pervious concrete shows superior strength when mixed with polymers. They retard

hydration process of cement and shows better strength at lower compaction efforts to achieve desired voids

content[11].

AIM OF STUDY

With the aim of preserving the environment and recharge groundwater also with making the urban space

more safe and user-friendly we are hereby suggesting the design requirements for the pervious concrete to be

able to use as Granular Sub Base Layer and Drainage Layer conforming to specifications laid down by

MoRTH[13], IRC: SP: 49[14].

EXPERIMENTAL PROGRAM

Material

Cement

To obtain sufficient strength, OPC of grade 43 is preferred for pervious concrete. But Portland Pozzolana

Cement (Fly Ash Content 34%) confirming to IS 1489-1 was used in the initial phase of my study in order to

make concrete more environmental friendly.

Aggregate

In order to have sufficient voids, aggregates of size range 19mm to 9.5mm are very suitable. However

aggregates of smaller size 9.5mm to 2.36mm are also used to increase strength. But use of smaller sized

aggregates decreases the permeability of concrete[12].

Fine aggregate and coarse aggregates of specific gravity 2.74 each were obtained from crusher. Water

absorption of aggregates is less than 3% according to IRC: SP: 49. All aggregates gradation adopted

confirms to IRC: 44–2008[15].

Admixture

In the initial phase of the study no chemical admixtures were used in the design.

Methodology

The design mix is based on Absolute Volume concept. Initially we have specified water cement ratio and

water content and then we calculated cement content and total volume of aggregates. Then volume of coarse

aggregate was determined from the total volume of aggregates as per IS: 10262-2009[16]. After that we

specified total volume of voids which is required for a design. Then we obtained the volume of fine

aggregate by subtracting the determined volume of fine aggregate minus the volume of voids. Gradation of

coarse aggregate was varied accordingly to identify the optimum design proportion.



Table 1 Mix Design Data MIX Mix 1 Mix 2 Mix 3

Water (liter) 136 136 180

Water Cement Ratio 0.32 0.32 0.4

Cement (Kg/m3) 425 425 450

Total Aggregate (m3) 0.73 0.73 0.68

Coarse Aggregate (% volume of total aggregate) 80 80 85

Percentage Voids (% volume of total aggregate) 15 5 10

Fine Aggregate (% volume of total aggregate) 5 15 5

Aggregate Cement Ratio 3.99 5.40 4.32

Slump (mm) 10 0 25

Aggregate Gradation

Coarse aggregate gradation was done using the stack of sieves according to IS: 2386 (Part I) – 1963[17]. The

result of gradation is shown in table and corresponding to graph as shown below.

Table 2 Coarse Aggregate Requirements For Single Sized Aggregates

Fig 1 Coarse Aggregate Gradation Curve

Method of Mixing and Compaction

Concrete mixing was done in transit/drum mixture. Total mixing time was kept as 4 minutes initially

dry mix was mixed for 2 minutes. After adding water mixing was done for 2 more minutes. Cubes casting

-20

0

20

40

60

80

100

120

110100

% P

ass

ing

log (sieve size)

% Passing Minimum % Passing Maximum Mix 1

Mix 2 Mix 3

IS Sieve Size (mm) % Passing Minimum % Passing Maximum % Passing Adopted

Mix 1 Mix 2 Mix 3

40 100 100 100 100 100

20 85 100 100 100 100

16 0 85 75 50 80

12.5 0 85 25 0 25

10 0 20 0 0 5

4.75 0 5 0 0 0



time was kept less than 30 minutes from the time of addition of water. Compaction was done by hands only

confirming to IS: 516-1959[18]. Five cubes of 150mmx150mmx150mm were casted for each mix.

TESTS

Compressive Strength Test

The aim of the study is to determine the compressive strength of pervious concrete. In the initial

phase of the project we tested 3 cubes for Compressive Strength at specific age of 7days of submerged

curing confirming to IS: 516-1959 for all mixes.

Water Permeability Test

The water permeability of concrete was determined using air and water permeability test apparatus.

Two cubes were tested for Water Permeability of concrete confirming to IS: 3085-1965 [19] for all mixes.

𝐾 =𝑄

𝐴𝑇𝐻

𝐿

Equation 1

K = Coefficient of permeability (cm/sec)

Q = Quantity of water percolating over the entire period of test after steady state has been reached.

(500mL)

A = Area of specimen face (15x15 cm2)

T = Time over which Q is measured (sec) 𝐻

𝐿= Ratio of pressure head to thickness of specimen, both expressed in same units (

100𝑐𝑚

15𝑐𝑚).

RESULTS

Table 3 Compressive Strength Test Results

M

ix

Water/Cement

Ratio

Aggregate/Cement

Ratio

7 days Compressive Strength

(MPa)

Water

Permeability

(x10-2

cm/sec)

M

1 0.32 3.99

3.35 2.76

3.82 2.83

2.67 -

M

2 0.32 5.40

4.36 1.52

3.80 1.81

4.51 -

M

3 0.40 4.32

6.74 2.14

5.71 2.06

6.13 -

Fig 2 Percentage Voids VS Water Permeability

CONCLUSION With very high percentage of coarse aggregate we will achieve higher water permeability.

Mechanical strength decreases if we increase the water permeability of concrete.

At high water cement ratio pervious concrete show shear failure on performing slump cone test for

determining workability.

Pervious Concrete is very prone to failure near edges and sides by impact load.

The failure occurs at cement paste and aggregate interface. Aggregates were unaffected when cubes were

tested for compressive strength.

1

1.2

1.4

1.6

1.8

2

2.2

2.4

2.6

2.8

3

0 5 10 15 20

Wa

ter

Per

mea

bil

ity

Percentage Voids

M2

M1

M3



Vibrating the pervious concrete wet mix is not suitable method of compaction as the cement matrix and

aggregates separates out which severely affects its strength.

Strength of pervious concrete increases if percentage of fine aggregates is increased.

From the observations and data obtained from the tests carried out on the samples of pervious concrete,

it is evident that it is quite difficult to design an economical and good quality of pervious concrete for

GSB/DLC sub base layer by using PPC (34% fly ash) and 80% quantity of coarse aggregate due to low

values of mechanical strength.

REFERENCE

[1] A. K. Chandrappa and K. P. Biligiri, “Comprehensive investigation of permeability characteristics of

pervious concrete : A hydrodynamic approach,” Constr. Build. Mater., vol. 123, pp. 627–637, 2016.

[2] “Portland Cement Pervious Pavement Manual.” Florida Concrete and Products Association, Inc.

[3] Caltrans, “Caltrans Storm Water Quality Handbook Pervious Pavement Design Guidance,” no. May,

2016.

[4] D. H. Nguyen, N. Sebaibi, M. Boutouil, L. Leleyter, and F. Baraud, “A modified method for the

design of pervious concrete mix,” Constr. Build. Mater., vol. 73, pp. 271–282, 2014.

[5] P. D. Tennis, M. L. Leming, and D. J. Akers, Pervious Concrete Pavements. 2004.

[6] A. Joshaghani, A. A. Ramezanianpour, O. Ataei, and A. Golroo, “Optimizing pervious concrete

pavement mixture design by using the Taguchi method,” Constr. Build. Mater., vol. 101, pp. 317–

325, 2015.

[7] R. Sriravindrarajah, N. D. H. Wang, and L. J. W. Ervin, “Mix Design for Pervious Recycled

Aggregate Concrete,” Int. J. Concr. Struct. Mater., vol. 6, no. 4, pp. 239–246, 2012.

[8] V. Schaefer, K. Wang, M. Suleiman, and J. Kevern, “Mix design development for pervious concrete

in cold weather climates,” Cent. Transp. Res. Educ. Iowa State Univ., no. February, p. 67, 2006.

[9] M. Suleiman, J. Kevern, V. R. Schaefer, and K. Wang, “Effect of compaction energy on pervious

concrete properties,” Concr. Technol. Forum-Focus Pervious Concr. Natl. Ready Mix Concr. Assoc.,

no. January, pp. 1–8, 2006.

[10] Y. Chen, K. Wang, X. Wang, and W. Zhou, “Strength , fracture and fatigue of pervious concrete,”

Constr. Build. Mater., vol. 42, pp. 97–104, 2013.

[11] F. Giustozzi, “Polymer-modified pervious concrete for durable and sustainable transportation

infrastructures,” Constr. Build. Mater., vol. 111, pp. 502–512, 2016.

[12] A. K. Chandrappa and K. P. Biligiri, “Pervious concrete as a sustainable pavement material-Research

findings and future prospects: A state-of-the-art review,” Constr. Build. Mater., vol. 111, pp. 262–

274, 2016.

[13] Ministry of Road Transport and Highways, Specifications for Road and Bridge Works, Fifth

Revision, New Delhi, India, 2013.

[14] Indian Road Congress, IRC: SP: 49-2014, Guidelines for the Use of Dry Lean Concrete as Sub-Base

for Rigid Pavement, New Delhi, India, 2014.

[15] Indian Road Congress, IRC: 44-2008 Guidelines For Cement Concrete Mix Design For Pavements,

New Delhi, India, 2008.

[16] Indian Standards, IS: 10262-2009 Concrete Mix Proportioning- Guidelines, New Delhi, India, 2009

[17] Indian Standards, IS: 2386-1 Methods of Test for Aggregates for Concrete, Part 1: Particle Size and

Shape, New Delhi, India, 1963.

[18] Indian Standards, IS: 516-1959 Method of Tests for Strength of Concrete, New Delhi, India, 1959.

[19] Indian Standards, IS: 3085-1965 Method of Test for Permeability of Cement Mortar And Concrete,

New Delhi, India, 1965.



Performance Enhancement of the Marshall Mix using

Epoxy Resins Jaspreet Singh, Maninder Singh, Neena Garg

ABSTRACT

In this study, the bitumen was doped with various percentages of epoxy resin and comparison was made

between Base bitumen and Epoxy asphalt. The dosages of epoxy resin was taken as 4.5%, 5% and 5.5%,

which was added to the optimum binder content and the Marshall Stability and the flow of the mixes were

determined using Limestone aggregates. The effects of epoxy content on bitumen were examined by using

Marshall Stability and ITS test. Marshall Stability of the epoxy asphalt at 5% dosage of epoxy resin was

found to be maximum compared to 4.5% and 5.5% resin content. The Indirect Tensile Strength of epoxy

asphalt was also maximum at 5% epoxy resin compared to 4.5% and 5.5%. It was finally concluded that best

results were obtained by using 5% of epoxy resin.

Keywords: Base asphalt, Limestone Aggregates, Epoxy asphalt, Stability, ITS, Marshall stability

1. INTRODUCTION

Epoxy asphalt is an excellent material to be used in the wearing course. It has better mechanical

properties and high temperature stability than that of the base binder [1, 2]. Epoxy resin when mixed with

asphalt it improves the Rutting Resistance, Fatigue Life of the mix and the flow at the higher temperatures.

During its processing temperature, it does not make mix too viscous and brittle at low temperatures [3 – 5].

Epoxy asphalt gives better heat resistance and provides high strength. Epoxy asphalt has a promising future

as compared to that of the other polymer modified asphalt binder.

In the last two decades, epoxy asphalt binder is widely used as a wearing course of Orthotropic steel

deck bridges and also in the Runway Pavement and Asphalt concrete pavements. It is a two – phase chemical

system in which thermosetting epoxy is in continuous phase and the conventional asphalt is in dispersed

phase. It is stored in typically 2 different components before mixing Epoxy as a component A and curing

agent/Blended asphalt as a component B. As epoxy is polar in nature, while asphalt is non – polar therefor to

mix these 2 compounds a crosslinker is used to form a proper bond between component A and B, so that

during curing there should not be a phase separation of the blend. There are different types of chemical

modifiers such as furan, Thiourea, sulfur, Maleic Anhydride (MAH) etc are used for the chemical

modification of the bitumen.

The first application of epoxy asphalt was on bridge decks was in mid – 1960’s by California Bay Bridge

Authority to pave the San Mateo – Hayward Bridge in 1967, which is in service today and servicing

extremely well after 47 years [6]. The epoxy asphalts was firstly used only in Runway Pavements. The

engineers from Shell oil company in 1960’s done an overlay of 12.5 – 25 mm on an old pavement, using 6.3

mm maximum size aggregates, the material exhibits excellent performance[7]. In 1986 first full – scale trial

was placed in Staffordshire (UK), which is exhibiting excellent performance[8] In 1990’s china develop first

formula for asphalt and exhibits a first test on small section of urban road in shanghai[8,9].

1. Jaspreet Singh Gill: Student ME Civil Infrastructure, Thapar University, Patiala.

2. Mr. Maninder Singh: Assistnat Professor, Punjabi University, Patiala.

3. Ms. Neena Garg: Assistant Professor, Thapar University, Patiala.

2. MATERIALS



Polymers are used to improve the performance of the pavement. There are different types of polymers

used in the modification of the bitumen. In this study the polymers used are PMB 40 and Epoxy resin are

used and test under Marshall testing machine and the viscosity of the epoxy asphalt is checked. PMB 40 is

taken from the Hincol polymers manufacturing company. The properties of the polymers are described

below in Table1. Epoxy resin is taken from a resin manufacturing company GO Green from Tamil Nadu.

The epoxy resin is in liquid form and in two separate compounds one is epoxy and the second one is hardner.

The properties of the epoxy resin and the hardener is shown Table 2. Limestone aggregates used in this

research is taken from quarry Paonta sahib situated in district sirmour. The gradation of Aggregates is taken

from MoRTH specification Grade – 2 and the gradation adopted is mid-point gradation for the mix design.

The test on aggregates was done to check the workability and strength of the aggregates. The results were

discussed in Table 1. Bitumen is the binding material used in road construction from past decades to bind the

aggregates together and to give smooth riding quality. The bitumen used in this study is VG 30, taken from

Iran’s Jeyoil manufacturers with following properties as described below in Table 2.

Table 1: Properties of PMB 40

Properties Results Test Method

Penetration at 25C 30 – 49 IS 1203 : 1978

Softening Point oC, min 59 IS 1205 : 1978

Ductility at 27 oC, min 50 IS 1208 : 1978

Table 2: Properties of Epoxy and Hardener

Properties Specification Units Araldite (LY556) Aradur (HY951)

Viscosity at 25oC ISO 12058 mPa.s 10,000 – 12000 10 – 20

Density at 25oC ISO 1675 gm/cc 1.15 – 1.20 0.97 – 0.99

Flash Point ISO 2719 oC > 200 >180

Table 3: Testing on Aggregates: Limestone

Testing on Aggregates Results Permissible Values

Impact Test 23.6 Maximum 27

Crushing Test 22 Maximum 45%

Specific Gravity of Coarse Aggregates 2.62 Maximum 3.0

Specific Gravity of Finer Aggregates 2.7 Maximum 3.0

Abrasion 20 Maximum 35

Table 4: Bitumen testing

Tests performed Results Permissible value

Penetration Test at 25oc/100 gm/5 sec, mm 65 60 – 70

Viscosity Test at 60oC 3050 >2400

Ductility Test at 25oC, cm 100 40 cm, minimum

Softening Point Test, oC 52 40 – 55

Specific Gravity 1.01 0.99 – 1.02



3. EXPERIMENTAL

The Marshall Stability and the flow of the mix is calculated using Marshall Stability Machine. The

Aggregates and the bitumen mixing should be done thoroughly so that the surface of aggregates should be

fully coated with bitumen. After placing the preheated mixture into the mould the sample is compacted with

a preheated mechanical hammer of 4.54 kg, approximately 75 number of blows at a rate of 30 number of

blows per minute to be given on both sides of the sample face. Load is applied perpendicular to the axis of

the specimen at a constant deformation rate of 51 mm per minute. The indirect tensile test results are also

used to determine the possibility of the moisture damage for both conditioned and unconditioned samples. In

this test, a compressive load of 50 mm/min is applied in a direction vertical to its diametrical plane. The

proving ring dial gauge gives the reading of failure load at its peaks and is known as the Indirect Tensile

Strength (ITS) of the mix. This value of ITS is used to calculate the rutting and fatigue potential of the

bituminous mix. The results are used to determine the field pavement potential for moisture damage under

both conditions.

Equation used to calculate the Indirect Tensile Strength:

ST =

4. Results and Discussions:

Marshall Stability test was conducted to calculate the Stability and Flow of the three binders as per

ASTM D6927 – 06. First, the Optimum Binder content is calculated using river bed aggregates and VG 30

bitumen. The OBC is calculated on the behalf of the volumetric analysis. The further investigation is done

using the OBC by varying the content of different Polymers on which the Stability and Flow of the mix is

calculated. Indirect Tensile Strength test is conducted to determine the possibility of the moisture damage of

the unconditioned sample. Volumetric properties of different binder at Optimum Binder Content.

The Volumetric analysis of the Base asphalt, PMB 40 and the Epoxy asphalt are discussed below in the

table 5.

Table 5: Volumetric Analysis of the 3 Binders

Volumetric properties Using BC grade – 2

VG 30 PMB 40 EPOXY RESIN

OBC, (% by weight of mix) 5.30 5.1 5.1

Stability, kg 2313.15 2538.24 2767.56

Flow, mm 3.4 3.39 2.5

VFB % 72.19 73 75.18

VMA % 15.81 16.12 16.69

Air Voids % 4 4 4

Density, g/cc 2.43 2.42 2.42



Figure 1: Stability Results of three Binders

Figure 2: Flow Results of three binders

From the above results the value of stability is maximum for epoxy as compared to the other binders

and the flow value is minimum for the epoxy asphalt binder. The epoxy asphalt shows excellent results in

comparison with the conventional asphalt and the Polymer Modified Asphalt (PMB 40). The flow of the

above mention all three binders epoxy asphalt has the minimum flow at the optimum binder content.



Figure 3: ITS results for different binders

The ITS performed on the all three binders at an optimum binder content of 5% the above results shows

the value of epoxy asphalt is maximum among the three other binders

5. CONCLUSION:

The thermosetting epoxy resins shows excellent properties when mixed with bitumen. It improves the

physical properties of the mix. The Marshall Stability and Flow of the epoxy asphalt at 5% shows better

results as compared to that of the PMB 40 and the Base asphalt. The value of ITS at Optimum Binder

Content is 73.5% and 37.6% higher than that of the base binder and PMB 40, respectively.

References:

1. Gonzalez, O; Munoz M.E; Santamarfa, A; Garcia – Morales, M; F.J.; Partal, P. Eur Polym J 2004, 40,

2365.

2. Issac, C.A; Debs, P.; Constr. Building Material 2007, 21, 157.

3. Herrington, P.; Albaster, D.; Arnold G.; Cook, S.; Fussell, A.; Reilly, S. Epoxy modified open – graded

porous asphalt. Economic evaluation of long – life pavement: Phase II, Design and testing of long – life

wearing courses. Land Transport New Zealand land Research Report; New Zealand, wellington, 321,

2007.

4. Simpson, W.C.; Summer, H. J.; Griffin, R. L.; Miles, T.K. ASCE J Airport Div 1960, 86,55.

5. OECD. Economic evaluation of long – life Pavements Phase, J. Europeon Confrence of Ministers of

Transport. OECD Publishing; 2005. Printed in France.

6. 6 Hicks R.G., Dussek I.J., Seim C. 2000. Asphalt surfaces on steel bridge decks. Transportation

Research Record: Journal of the Transportation Research Board, No. 1740. TRB, National Research

Council, Washington D.C Paper no 000389.p.135 – 142.

7. 7 Simpson R.L, Griffin W.C, Sommer H.J., Miles T.K Design and Construction of Epoxy Asphalt

Concrete Pavements. Presentation at the meeting of the Highway Research Board of the Council,

Washington, D.C., 1960.

8. 8 Lu WM, Guo ZY. Compounding of the High Strength Asphalt Concrete and its Properties. Chin J

Highway Transport 1996;9(1): 8 – 13.

9. 9 Lu W.M., Guo Z.Y., Wang X.L., Li J.H Characteristics Performance and Application of Cold Mix

Epoxy Asphalt. East China Highway. No.2, 1996.p.64 – 68.



Design of Storm Water Drainage Network for Urban City

Neeraj Kumar 1, Malvee Singla

2 and Harsh Dalal

3

1 Trainee Teacher, National Institute of Technology, Uttarakhand, Email-

[email protected] 2

M.Tech Scholar, Indian Institute of Technology, Roorkee, [email protected] 3 Harsh Dalal, Former UG Student, SVNIT Surat, [email protected]

ABSTRACT

The urban development and population growth can create a huge impact on urban water management. Proper functioning of storm water drainage system is a key parameter in preservation and improvement of

urban water environment. Paved roads, construction of houses and commercial buildings, parking lots, etc. increases the imperviousness of ground. As impervious surface area increases, the storm water coming off

increases velocity and volume of runoff. This paper presents the key design of storm water drainage for Dholera Special Investment Region TP-4

(part). The design is based on the rainfall data of 41 years (1961--2002) which have been taken for study. Using the rainfall data of 41 years, Intensity duration curve has been derived. The total area of TP-4(part) of

Dholera Special Investment Region is 5.36 km2 .This areas consist of industrial zone (52.77%), residential

zone (30.35%), recreation, sports and entertainment zone (8.64%) and tourism: resorts (8.23%). Here, Kirpich method has been used for estimation of storm water runoff. The design is carried out as per the guidelines given in CPHEEO (2013) manual of Ministry of Urban Development, Govt. of India. The outfalls of system are directed to proposed site.

Similar Approach can be adopted for the city located in Hilly Region with some correction factors for efficient and effective storm water drainage network.

Key Words: Storm Water, Storm Water Drainage Network.

1. INTRODUCTION

A storm drain is defined as that portion of storm drainage system that receives runoff from inlets and

conveys the runoff to some point where it is then discharge into the channel, water body, and pipe system. It consists of one or more pipes connecting one or more inlets. Storm drain may be closed conduit or open conduit or combination of the two.

The complete system is referred as storm drain system and will normally consist of curbs and/or gutters, inlets or catch basins, laterals or leads, trunk lines or mains, junction chambers, manholes, and ponds. The purpose of the storm drain is to collect storm water runoff from the roadway and convey it to an outfall.

Storm drain design generally consists of three major parts:

System planning which includes data gathering and outfall location

Pavement drainage which includes pavement geometrics and inlet spacing

Location and sizing of mains and manholes.

Damage to surrounding and adjacent property, resulting from water overflowing the roadway curbs and

entering such property, risk and delay to traffic caused by excessive ponding in sags or excessive spread along the roadway, increased potential for accidents and weakening of base and subgrade due to saturation from frequent ponding of long duration are most serious effects of inadequate roadway drainage system.

The principles of Sustainability, Level of service and Cost-effectiveness are to be satisfied for the

selection of design criteria of storm water drainage network for effective drainage of storm water to prevent any type of flooding in the region.







Figure 1: Layout of Storm Water Drainage Network

2. STUDY AREA

Dholera Special Investment Region: DHOLERA is situated in Ahmedabad district in the Gulf of

Khambhat. Dholera is in proximity with the coastal line. It is covered by water faces on three sides, namely, on the east face by Gulf of Khambhat, on the north side by Bavaliari creek and on southern side by Sonaria creek.

Figure 2: Dholera City Plan



The project is spread over an area of 35,000 hectares. The processing area which is proposed is 14,000 hectares and rest is non- processing zone. Project goals are to double the employment potential, triple industrial output and quadruple exports from the region in next five years.

Total area of Dholera TP-4(part), taken for our problem is 495.058 hectares. Dholera TP-4(part) Plan consists of following zone:

1. Industrial zone= 261.242 hectares

2. Residential zone= 150.250 hectares

3. Recreational, Sports and Entertainment Zone= 37.822 hectares

4. Tourism: Resorts = 35.793 hectares

.

Figure 3: Land Use Map for DSIR

3. DESIGN METHODOLOGY

Runoff Coefficients The proportion of the total rainfall that will reach the storm drains depends on the percent

imperviousness, slope, and ponding character of the surface. The runoff coefficient is also dependent on the

character and condition of the soil. The infiltration rate decreases as rainfall continues, and is also influenced by the antecedent moisture condition of the soil. The infiltration rate decreases as rainfall continues, and is

also influenced by the antecedent moisture condition of the soil. Field inspection and aerial photographs are useful in estimating the nature of the surface within the drainage area.

Table-1: Runoff coefficient for different land use

Sr. No. Type of Area Percentage of Imperviousness

1 Commercial and Industrial Area 70 to 90

2 High density Residential Area 60 to 75

3 Low density Residential Area 35 to 60

4 Parks and other Undevelopment Area 10 to 20

The weighted average imperviousness of the drainage basin may be estimated as:

C Ai Ii Where A= Drainage area contributory to section of drain

I= Imperviousness factor of respective area



Figure 4: Design Procedure for Storm Water Drainage Network

Time of Concentration The rainfall intensity i is the average rainfall rate in millimetre per hour for a particular drainage basin

or sub-basin. The intensity is selected on the basis of the design rainfall duration and return period. The

design duration is equal to the time of concentration for the drainage area under consideration. The time of

concentration in a storm drainage system is the sum of the inlet time to (the time it takes for flow from the

remotest point to reach the sewer inlet), and the flow time t f in the upstream sewers connected to the outer

point:

t c t o t f

The inlet time, or time of concentration for the case of no upstream sewers, can be obtained by Kirpich Formula:

tc 0.0195L0.77

S 0.385

Where L = length of channel/ditch from headwater to outlet, m

And S = Average catchment slope, m/m.

Intensity Duration Frequency Curve (IDF) IDF curve is a graph with duration plotted as abscissa, intensity as ordinate and a series of curves, one

for each return period. An IDF curve gives the expected rainfall intensity of a given duration of storm having desired frequency of occurrence. Here, Gumbel Extreme Value Distributions (Type-I), also known Gumbel method’s parameter are used for IDF Curve.



Return Period

The probability of occurrence of an event of a random variable (e.g. rainfall) whose magnitude is equal to or in excess of a specified magnitude X is denoted by P. So, return period (also known as recurrence interval) is defined as

𝑇 =1

𝑃

Shorter Duration Storm The design duration shall be taken equal to the time of concentration. For roadside drains, the time

of concentration is generally of the order of 5, 1 0, 1 5, 20, 30 or 40 minutes and it is a general practice in India to collect and measure accumulated rainfall and record values once or twice in 24 hours.

A general equation given in IRC:SP:13, may also be used for deriving intensity for shorter

duration.

𝑖 =𝐹

𝑇(

𝑇 + 1

𝑡 + 1)

The equation is: Where, i = Intensity of rainfall within a shorter period of ' t ' hours within a storm F = Total rainfall in a storm in cm falling in duration of storm of T hours. t = Smaller time interval in hours within the storm duration of T hours.

T = Duration of total rainfall (F) in hours

5. RESULTS

Intensity Duration Curve The duration of rainfall which we have taken for our design purpose are 1 hour, 2 hour ,3 hour, 6

hour, 12 hour and 24 hour. For every year, from the data we have been supplied from Dholera rainfall

station, we have calculated maximum rainfall for various duration and year.

The return period which we have taken for our problem are 1.5 year, 2 year,5 year,10 year,20

year, 50 year and 100 year.

Computation of Gumbel distribution parameters

Table 2: Gumbel Parameter Valuation

Duration Mean Standard deviation

6s

Rainfall X S

u x 0.5772

1 Hour 68.08 46.54 36.30 47.13

2 Hour 90.68 62.00 48.36 62.77

3 Hour 101.92 69.68 54.35 70.55

6 Hour 116.81 79.86 62.29 80.86

12 Hour 125.44 85.76 66.89 86.83

24 Hour 131.71 90.04 70.23 91.17



Table 3: Gumbel Parameter Valuation

Duration Rainfall

𝑌𝑡 = − ln (ln (

𝑇

𝑇−1))

100 Year 50 Year 10 Year 5 Year 02 Year 1.5 Year

1 Hour 4.60 3.90 2.25 1.50 0.37 -0.09

2 Hour 4.60 3.90 2.25 1.50 0.37 -0.09

3 Hour 4.60 3.90 2.25 1.50 0.37 -0.09

6 Hour 4.60 3.90 2.25 1.50 0.37 -0.09

12 Hour 4.60 3.90 2.25 1.50 0.37 -0.09

24 Hour 4.60 3.90 2.25 1.50 0.37 -0.09

The rainfall intensity after the 24 hour correction for different time period and different rainfall

duration is computed as below:

Table 4: Maximum Intensity for Different Time Duration And Different Return Period

Duration X t u Yt

Rainfall

100 Year 50 Year 10 Year

05 Year 02 Year 1.5 Year

1 Hour 278.34 245.30 167.44

132.05 78.72 57.01

2 Hour 370.76 326.76 223.04

175.90 104.87 75.95

3 Hour 416.71 367.25 250.68

197.69 117.85 85.35

6 Hour 477.62 420.94 287.32

226.59 135.08 97.83

12 Hour 512.87 452.00 308.53

243.31 145.05 105.05

24 Hour 538.53 474.61 323.96

255.48 152.30 110.30



1. The IDF Curve, using Gumbel parameter, for DSIR TP-4(Part) is obtained as follows:

Figure 5: IDF Curve for DSIR TP-4 (Part)

2. As the city is of national importance, hence Return period for 2 years is selected for our design purpose. 3. Total length of trunk of stormwater drainage network is 29.22 km and highest size of conduit is 29.22

km. 4. The estimated cost of designed network is approx. 40 crores.

REFERENCES

1. Baleva, S.N. and Mishra, K.R., 2016. Overview of Storm Water Network Of East Zone Of Ahmedabad

City. Development, 3(2).

2. Chow, V.T., Maidment, D.R. and Larry, W., 1988. Mays, Applied Hydrology. International edition,

MacGraw-Hill, Inc, p.149.

3. Echols, S. and Pennypacker, E., 2008. Stormwater as amenity. The application of artful rainwater design.

In Proceedings of the 11th International Conference on Urban Drainage, Edinburgh, Scotland, UK.

http://cws. msu. edu/documents/Echols_Stormwaterasamenity. pdf (dostęp: 16.05. 2010).

4. Jain, R 2007, ‘Storm Water network Design Of North Zone of City of Ahmedabad’, M. Tech thesis, 5. Indian Roads Congress SP 042-2014, Guidelines on Road Drainage (First Revision), Indian Roads

Congress, New Delhi

6. Kaltenbach, A.B., 1963. Storm sewer design by the inlet method. Public Works, 94(1), pp.86-89.

7. Kashefizadeh, M., bin Yusop, Z. and Hekmat, A., 2013. Advancing stormwater management practice in

Iran using water sensitive urban design approach. International Journal of Water Resources and

Environmental Engineering, 5(9), pp.515-520.

8. Kellagher, R.B.B. and Udale-Clarke, H., 2008. Sustainability criteria for the design of stormwater drainage

systems for the 21st century.

9. Ministry of Urban Development- 2013, Manual on Sewerage and Sewage Treatment- Central Public Health and Environmental Engineering Organisation, New Delhi

10. Roesner, L.A., 1974. Impact of Stormwater Runoff on Receiving Water Quality. Short Course Proceedings: Applications of Stormwater Management Models August 19-23, 1974, Amherst,

Massachusetts, University of Massachusetts, Amherst, p 159-176. 9 fig, 5 tab, 10 ref. 11. Subramanya, K., 2013. Engineering Hydrology, 4e. Tata McGraw-Hill Education.

https://archive.org/details/gov.in.irc.gov.in.sp.042.2014



Performance based seismic design of structure: A review

Sumit1, Dr. S.M.Gupta

2

1 M.Tech Student (Structural Engineering), N.I.T Kurukshetra,Haryana,India, [email protected]

2 Professor,Civil Engineering Department, N.I.T , Kurukshetra, Haryana,India, [email protected]

ABSTRACT

A detailed review of previous available analysis of structures using performance based seismic

design (PBSD) method is presented in this paper. The seismic analysis and design of reinforced concrete

(RC) frames is still an unresolved issue due to its complex behaviour. Seismic design codes are traditionally

based on the force-based approach wherein structures are designed with a certain minimum lateral strength.

However, it has been observed that such an approach, which relates to the elastic response, does not produce

consistent inelastic response in terms of the amount and distribution of damage in structural elements. In

views of the above, the displacement based approach, also known as performance based design (PBD)

approach, has been explored in recent times.

The basic concept of Performance Based Seismic Design is to provide engineers with the capability

to design buildings that have a predictable and reliable performance during earthquake. Performance based

design, comparatively an advance method having advantages over the strength based design as suggested by

many researcher, is a design procedure followed to achieve realistic behavior of the structure and eventually

resulting in a reliably accurate earthquake resistant design.

The performance based design utilizes “static nonlinear pushover analysis” as a tool to estimate the

nonlinear capacity of the structure. The Graph of pushover curve has been plotted in terms of base shear -

roof displacement.

Key Words: Performance Based Seismic Design, Performance Objective

INTRODUCTION

Amongst the natural hazards, earthquakes have the potential for causing greatest damages. According to

the existing standard code of practice IS: 1893(part-1)-2002, more than 60% of existing land is vulnerable to

different kinds of earthquakes. Since earthquake forces are random in nature & unpredictable, the

engineering tools needs to be sharpened for analyzing structures under the action of these forces. Following

the 1989 Loma Prieta and 1994 Northridge earthquakes, structural engineers in the United States began

development of structural design procedures that changed emphasis from strength to performance. The

resulting criteria and methodologies came to be known as “performance based design.” Interest in these

procedures has spread throughout the international earthquake engineering community.

Performance based design is gaining a new dimension in the seismic design philosophy wherein the near

field ground motion (usually acceleration) is to be considered. Earthquake loads are to be carefully modeled

so as to assess the real behavior of structure with a clear understanding that damage is expected but it should

be regulated. In this context pushover analysis which is an iterative procedure shall be looked upon as an

alternative for the orthodox analysis procedures. In pushover analysis of building, subjecting them to

monotonically increasing lateral forces with an invariant height wise distribution until the preset

performance level (target displacement) is reached.

The Performance Based Seismic Design (PBSD) is a rapidly growing idea that is present in all

guidelines in all recent guidelines in USA like Vision 2000 (SEAOC, 1995), ATC40(ATC, 1996),

FEMA273(FEMA, 1997), and SAC/FEMA350 (FEMA, 2000a). This PBSD of buildings has been practiced

since early in the twentieth century. Developed countries like England, New Zealand, and Australia had their

performance based building codes in place for decades. The International Code Council (ICC) in the United

States had a performance code available for voluntary adoption since 2001 (ICC, 2001).

What makes performance-based seismic engineering (PBSE) different and more complicated is that in

general this massive payoff of performance-based design is not available. That is, except for large-scale

developments of identical buildings, each building designed by this process is virtually unique and the

experience obtained is not directly transferable to buildings of other types, sizes, and performance objectives.

Therefore, up to now PBSE has not been an economically feasible alternative to conventional prescriptive

code design practices. Due to the recent advances in seismic hazard assessment, PBSE methodologies,



experimental facilities, and computer applications, PBSE has become increasing more attractive to

developers and engineers of buildings in seismic regions.

LITERATURE REVIEW

Dilip J. Chaudhari and Gopal O. Dhoot (2016) presented a four-storey RC building is modelled and

designed as per IS 456:2000 and analyzed for life safety performance level in SAP2000 v17. Analysis is

carried out as per ATC 40 to find out storey drift, pushover curve, capacity spectrum curve, performance

point and plastic hinges as per FEMA 273 in SAP2000 v17. Concluded that by this design building lies in

between immediate occupancy and life safety range. So, the required performance objective of design is

achieved.

S Monish and S Karuna (2015) presented the analysis of two types of plan irregularities namely

diaphragm discontinuity and re-entrant corners in the frame structure. These irregularities are created as per

clause 7.1 of IS 1893:2002(part1) code. Various irregular models were considered having diaphragm

discontinuity and re-entrant corners which were analysed using ETABS to determine the seismic response of

the building. The models were analysed using static and dynamic methods, parameters considered being

displacement, base shear and fundamental natural period. It was found that the model which is most

susceptible to failure under very severe seismic zone, modelling and analysis is carried out using ETABS.

Concluded that the response spectrum method are accurate, when compared with equivalent static

method, since the method is based only on empirical formula. The performance of model D1 (H shaped) and

L3 was more vulnerable to earthquake than rest of the models.

S.P Akshara (2015) studied the displacement based approach known as performance based seismic

design (PBSD), which evaluated how building system are likely to perform under different potential hazards

events, by using of non-linear static pushover analysis a five storey residential RC building analysed for

seismic performance using dual requirement of life safety under design basis earthquaken (DBE) and

collapse prevention under maximum considered earthquake (MCE) and it was found that to satisfy the

strength requirement but failed to satisfy one of the displacement requirement at maximum considered

earthquake (MCE). Also storey drift requirement specified by IS 1893:2002 is not satisfied.

Dimpleben P.Sonwane and Prof. Dr. Kiran B.Ladhane (2015) presented an effective computer based

technique that incorporates pushover analysis together with pushover drift performance design of RC

buildings is carried out. The study begins with the selection of performance objectives, followed by

development of preliminary design, an assessment whether design meets performance objectives or not,

finally redesign and reassessment, if required, until the desired performance level is achieved. Studied a

(G+4) storied unsymmetrical(L-Shaped) reinforced concrete building designed according to IS 456:2000,

analysed using pushover analysis in SAP2000. The building is considered as special moment reisisting frame

(SMRF) and find the performance of building.

Determined best possible combination of reinforcement which was economical, effective and having

minimum damage to enable immediate occupancy and termed as performance based design. Reinforcement

of various elements of the structure i.e. the beams and the columns was increased in different combinations

and their effect on the performance of the structure was studied. The design of reinforcement was done in

STAAD.Pro and analysis was carried out using SAP2000 nonlinear software tool. The effect of shear wall on

the performance of the structure is also studied in this work and concluded that Performance increases on

increasing reinforcement of columns only resulting into an appreciable decrease in the maximum roof

displacement and increase in the base shear. Decrease in roof displacement is maximum interior column and

for corner and mid-face columns it is comparable. Performance of the building decreases when the sectional

sizes of beams and columns are reduced while keeping same reinforcement. Provision of shear wall results in

a huge decrease in base shear and roof displacement in unsymmetrical building.

Arvind. S. Khedkar et al.(2014) carreied out a comparison between Performance based Seismic design

and conventional design method (using I.S 1893; 2002) for irregular RC building frames (10 storeys) and

evaluates performance using pushover and Time History analysis. Following points are observed during

whole design process; The Performance Based Seismic Design method is based on the “strong column weak

beam” concept in which the beams are designed as per plastic moments calculated and columns are designed

which ensures larger life safety of the structure. Basic difference between regular and irregular frame design

is for upper storey the calculations for base shear decreases due to asymmetry. Performance point of the

frames designed by PBSD method is enhanced than for all frames designed by conventional method. For the

irregular frame with two step setback when designed by conventional method (I.S 1893;2002) method

displacement is maximum than other two frames after performing time history analysis. For the irregular



frame with two step setback when designed by PBSD method the displacement is lowest after time history

analysis compared to the irregular frame with one step setback and regular frame. This proves the degree of

reliability of Performance based seismic design method. Time period is one of the effective means to check

the reliability of PBSD method.

Q. Xue et al. (2013) Presented that methodology of performance-based seismic design and evaluation

has been studied for several years. The result has been applied in developing the seismic design draft code in

a commissioned research project sponsored by the Architecture and Building Research Institute at Ministry

of the Interior. In addition, a generalized numerical method for displacement-based seismic evaluation and

direct displacement-based seismic design is also developed. Sensitivity study on the design parameters are

carefully carried out to find the optimal setting in order to increase the design efficiency. It has been found

that the design procedure based on the yielding displacement estimated through proper empirical formula is

more efficient for ordinary buildings because of the resulted non-minimum strength.

Ms. Nivedita N. Raut and Ms. Swati D.Ambadkar (2013) studied pushover analysis under strong ground

motion , effect of the layout of masonry infill panels was investigated over the elevation of masonry infilled

RC frames on the seismic performance and potential seismic damage of the frame based on realistic and

efficient computational models and compared base shear vs. displacement in bare frame, infill wall frame

and ground. It was seen that displacement was more than other two frames at roof level in bare frame and at

ground floor in weak story displacement was more than other two frames. Hinge formation in the beam is

more than column.

Wei Li and Li Qing-Ning (2012) presented the advantages and disadvantages of the current seismic

design code in China. Suggesting the tall building structures beyond the code specification (TBBC), applying

PBSD method due to its many advantages of PBSD and aiming TBBC characteristics, a PBSD flowchart is

presented and the proposed code is described. Structural seismic performance objectives, performance levels

and the main method to implementation of PBSD have been presented. Site feasibility requirements,

conceptual design scopes and basic rules have been proposed. Performance objective-oriented procedures for

preliminary design and seismic performance evaluation have been presented. Suggestions on seismic

performance criteria and the evaluation of new TBBC have been made. In order to verify the feasibility of

PBSD for application of TBBC, a typical case study has also been conducted. It is believed that PBSD

methodology will bring a new era to engineering practices with increased confidence in, and reliability on,

seismic performance and safety.

Dalal Sejal P et al. (2011) observed that for various other different types of structures more research

work is needed, especially for development of PBPD (PerformanceBased Plastic Design) method. It is

important to note that in the PBPD method, control of drift and yielding is built into the design process from

the very start, eliminating or minimizing the need for lengthy iterations to arrive at the final design. Other

advantages include the fact that innovative structural schemes can be developed by selecting suitable

yielding members and/or devices and placing them at strategic locations, while the designated non yielding

members can be detailed for no or minimum ductility capacity.

P. Poluraju and P. V. S. Nageswara Rao (2011) studied the performance of reinforced concrete frames

using the pushover analysis, they concluded that the behavior of properly detailed reinforced concrete frame

building is adequate as indicated by the intersection of the demand and capacity curves and the distribution

of hinges in the columns and the beams. Hinges were mostly developed in the beams and few in the columns

but with limited damage.

Yousuf Dinar et al. (2014) evaluated and compared the performances of bare, different infill percentage

level, different configuration of soft storey and Shear wall consisting building structures with each other and

later depending upon the findings, suggests from which level of performance shear wall should be preferred

over the infill structure and will eventually help engineers to decide where generally the soft storey could be

constructed in the structures. Above all a better of effects of pushover analysis could be summarized from

the findings. Masonry walls are represented by equivalent strut according to pushover concerned codes. For

different loading conditions, the performances of structures are evaluated with the help of performance point,

base shear, top displacement, storey drift and stages of number of hinges form. The results lead to a decision

that infill, shear wall and soft storey configuration significantly affects the performance of the structures of

rigid joint. Under performance based analysis which is pushover, increasing infill increases the performance

overall while shear wall has maximum resistance against any lateral loads. The comparison of performance

of all soft storey cases under pushover analysis reveal that shipment of soft storey in each floor upward or

downward has a significant effects.

Y.Fahjan et al. (2012) examined the consistency of different approaches for nonlinear shear wall

modeling that are used in practice. For this purpose, 3, 5 and 7-story reinforcement concrete (RC) frames



with shear walls are analyzed using nonlinear two dimensional nonlinear finite element method under

constant gravity loads and incrementally increased lateral loads. The analysis results for these models are

compared in terms of overall behavior of the structural system. Besides, definition of the plastic hinge

properties which strongly affects the prediction of the capacity curve of RC wall in the pushover analysis is

also discussed. This paper focused on nonlinear modeling of RC shear walls for low and mid-rise buildings.

In current practice, plastic hinges develop along the critical height of the shear walls. In this study, the shear

walls were assumed to be a part of a RC frame so that different nonlinear modeling techniques would fully

be investigated, and concluded that

1. The shear wall with two layers of longitudinal and transverse reinforcement bars could be modeled with

multi-layer shell and mid-pier frame with plastic hinges to reflect the material nonlinearity. The plastic

hinge properties of the shear wall could be defined using FEMA 356 recommendation or fiber-based

hinge property. The pushover analysis based on FEMA 356 model and fiber model produced identical top

displacement-base shear curves for the sample frames. These curves are approximately similar except

multi-layer shell model for all cases.

2. Number of plastic hinge locations is a major key for the accurate representation of the inelastic

phenomenon for the RC shear walls.

Yernagula.Pratap and P.V S. Neelima (2015) presented a procedure and methodology adopted in

performance based design and its implications to achieve an earthquake resistant design. A G+4 storey

commercial building, assumed to be situated in seismic zone IV (according to IS: 1893(part1)-2002), is

considered for the case study. Static nonlinear pushover analysis is performed to estimate the capacity of the

building represented in the form of a pushover curve. Five performance levels, based on the criteria for

earthquake resistant design, are defined for the building. The hinge mechanism obtained in each step of the

pushover, is studied to obtain a desired performance level. From the analysis was found that the hinges

where developed in the beams and few in columns but with limited damages. Performance increase on

increasing reinforcement of columns & beams and it resulting into an appreciable decrease in the roof

displacement. To increase the reinforcement of columns found the maximum increase in base shear.

Performance of building decreases when the sectional size of beams and columns are reduced with same

reinforcement. Concluded that the performance based seismic design satisfied the acceptable criteria of

Immediate occupancy, life safety of the building under various intensity of earthquake.

S.R. Satish Kumar and G.Venkateswarlu (2008) examined the influence of parameters such as the

strength, stiffness, energy dissipation capacity and detailing such as percentage of reinforcement and amount

of confining steel on the local and overall damage is considered. Over 700 RC regular frame of two, four and

eight storeys designed and detailed as per indian seismic codal provisions are analyzed by varying the time

period, response reduction factor and percentage of longitudinal reinforcement. Non-linear time history

analysis for six different earthquake accelerograms are carried out using the pivot hysteretic model.

Variation in response parameters with time period, percentage of reinforcement with response reduction

factor are presented. It was found that the percentage of reinforcement plays a major role in the seismic

performance. Based on the study a simple design procedure to implement performance based design is

suggested.

Vipul Prakash (2004) presented the prospects for performance based engineering (PBE) in India. Based

on an extensive damage survey of the region, comparisons among old and new seismic codes in India and

Performance Based Engineering (PBE) based draft codes in US, documents produced under National

Programme on Earthquake Engineering Education (NPEEE), and the unabated popularity of seismically

vulnerable constructions in India. It lists the pre-requisites that made the emergence of PBE possible in

California, compares the situation in India and discussed the tasks and difficulties for implementing PBE in

India.

Considerable effort will still be required to translate the modeling guidelines and evaluation criteria

available in PBE draft codes (ATC, 1996; FEMA 1997b, 2000) for use in India, because the system of units,

testing procedures and construction practices in India are different from those in USA.

Continued economic growth is likely to result in shorter design life of existing seismically vulnerable

buildings, and provide the needed funds for their replacement by seismically more robust buildings.

Therefore, in spite of the present shortcomings, the future of Performance Based Engineering in India is

bright!



CONCLUSION

Several approaches for the PBSD method proposed by researchers have been reviewed in this paper. The

promise of performance-based seismic engineering (PBSE) is to produce structures with predictable seismic

performance. To turn this promise into a reality, a comprehensive and well-coordinated effort by

professionals from several disciplines is required. It is also noted that the PBSD for any structure is greatly

influced by the conditions such as the shape and size, material, objectives etc which are unique for each

structure which has to be taken care.

It is safe to say that within just a few years PBSE will become the standard method for design and

delivery of earthquake resistant structures. In order to utilize PBSE effectively and intelligently, one need to

be aware of the uncertainties involved in both structural performance and seismic hazard estimations.

REFRENCES

[1] S. Monish, S. Karuna, A study on seismic performance of high rise irregular rc framed buildings, IJRET:

International Journal of Research in Engineering and Technology eISSN: 2319- 1163 | pISSN: 2321-7308, 2015

[2] Arvind. S. Khedkar, Rajkuwar. A. Dubal and Sandeep. A. Vasanwala, Performance Based Seismic Design of

Reinforced Concrete Moment Resistant Frame with Vertical Setback, IJERT ISSN: 2278-0181 Vol. 3 Issue

2,February – 2014.

[3] Q. Xue, W. C. Hsu, C. Jhang, M. J. Tsai and C. W. Wu,Performance-Based Seismic Design and Evaluation of

Irregular Building Structures Sinotech Engineering. Vol.118. January 2013.

[4] ASCE, 2000,Prestandard and Commentary for the Seismic Rehabilitation of Buildings, FEMA 356 Report, prepared

by the American Society of Civil Engineers for the Federal Emergency Management Agency, Washington, D.C.

[5] ATC, 1997a, NEHRP Guidelines for the Seismic Rehabilitation of Buildings,FEMA 273 Report, prepared by the

Applied Technology Council for the Building Seismic Safety Council, published by the Federal Emergency

Management Agency,Washington, D.C.

[6] SEAOC, 1995, Vision 2000: Performance Based Seismic Engineering of Buildings, Structural Engineers

Association of California, Sacramento, California.

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Researches in Engineering Civil And Structural Engineering Volume 13 Issue 4 Version 1.0 Year 2013

Type:Double Blind Peer Reviewed International Research Journal Publisher: Global Journals Inc. (USA)

[8] Li Wei and Li Qing-Ning, Performance-based seismic design of complicated tall buildingstructures beyond the code

specifi cation, Struct. Design Tall Spec. Build. (2012). Published online in Wiley Online

Library(wileyonlinelibrary.com). DOI: 10.1002/tal.637.

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International Journal of Civil And Structural Engineering Volume 1, No 4, 2011.

[10] P. Poluraju, P. V. S. Nageswara Rao, Pushover analysis of reinforced concrete frame structure using SAP 2000,

International Journal of Earth Sciences and Engineering ISSN 0974-5904, Volume 04, No 06 SPL, October 2011,

pp. 684-690

[11] Qiang Xue, et. Al.,The draft code for performance-based seismic design of buildings in Taiwan, Civil and

Hydraulic Engineering Research Center, Sinotech Engineering Consultants Inc., Taiwan, 2 October 2007.

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Hall,1995.

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Pushover Analysis, International Research Journal of Engineering and Technology (IRJET) ISSN: 2395 -

0056,Volume: 02 Issue: 03 | June-2015.

[14] Dilip J. Chaudhari, Gopal O. Dhoot, Performance Based Seismic Design of Reinforced Concrete Building, Open

Journal of Civil Engineering, 2016, 6, 188-194,March 2016.

[15] Yousuf Dinar, Md. Imam Hossain, Rajib Kumar Biswas, Md. Masud Rana, Descriptive Study of pushover

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ISSN: 2278-1684,p-ISSN: 2320-334X, Volume 11, Issue 1 Ver. II (Jan. 2014).

[16] Dimpleben P. Sonwanw, Prof. Dr. Kiran B. Ladhane, Seismic Performance based Design of Reinforced Concrete

Buildings using Nonlinear Pushover Analysis, International Journal of Engineering Research & Technology

(IJERT) ISSN: 2278-0181, Vol. 4 Issue 06, June-2015.

[17] Yernagula.Pratap, P.V S. Neelima, Performance Based Design:Case Study, International Journal of Engineering

and Management Research, ISSN (ONLINE): 2250-0758, ISSN (PRINT): 2394-6962, Volume-5, Issue-3, June-

2015.

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frames with shear walls, 15 WCEE Lisboa 2012.

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The 14th

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Paper No. 447, Vol. 41, No. 1, pp. 201-222, March 2004.



A Review on the Free Vibration Analysis of Laminated

Composite and Sandwich Shells

Javed Ahmad Khan1, H. D. Chalak

2

1M.Tech. Scholar, Dept. of Civil Engineering, NIT Kurukshetra, [email protected] 2Assistant Professor, Dept. of Civil Engineering, NIT Kurukshetra, [email protected]

ABSTRACT

With the widespread use of laminated and sandwich structures in various industries such as automobile,

aerospace, civil, marine etc. A number of theories had been put forwarded by various researchers. Each

theory has its own assumptions which are reflected as their limitations while analysing the sandwich shells.

A review of recent research carried out on the free vibration analysis of multi-layered laminated shells is

presented in this paper. The strengths and weaknesses of these theories is also discussed in detail regarding

their applications. The review also includes the analytical analysis of laminated shells using different finite

element software packages. A discussion was also carried out in last regarding the application of analytical

software for analysis.

Key Words: Laminated shells, Composite shells, Free vibration.

1. INTRODUCTION

Shells made of laminated composites are gaining popularities in several industries such as automobile,

naval, aerospace, armed vehicles, nuclear containments etc. Though composite structures exhibits best

properties such as high strength to stiffness ratio, light weight, insulation properties etc. But to bonding of

layers of materials, problems of stress concentration at the interfaces, delamination, matrix cracking comes

into picture.

Because of material and structural irregularities and the application of composites in such important

industries, it is important to analyse the strength of structures made from these materials. A lot of work has

been done for the analysis of composites and FGMs shells using different theories and methods in static,

vibration, buckling mode (linear as well as non-linear ranges) with different shapes and material properties

subjected to different kinds of loadings (sinusoidal, point, udl, thermal, hygrothermal or electrical loads etc.).

Each theory as well as method has its own strength and weaknesses. Under this article, an extensive review

of different theories with different methods proposed by the researchers for the analysis of Laminated

Composite shells is presented.

The theories available for the analysis of laminated composite and FGM shells are broadly classified into

two categories:

a. 3D Elasticity Theories b. Equivalent Single Layer (ESL) Theories By making suitable assumptions in the 3D Elasticity Theories for kinematics of deformation or stress

state along the thickness of the shell, ESL models can be worked out.

Classical Lamination/Shell Theory (CST), First-order Shear Deformation Theory (FSDT), Higher-order

Shear Deformation Theory (HSDT) and Zig-Zag Theories (ZZT) come under ESL category. The basic

concept of ESL is modelling the entire lamination scheme into an equivalent single layer with the help of

homogenization technique. Using Love-Kirchhoff assumptions in addition to ESL models are known as CST

theories. Due to negligence of transverse-shear deformations in CST formulations, accurate stress and

displacements cannot be predicted. This gives rise to analysis of LC shells using FSDT.

Analysis of composites using FSDT does not give zero shear stresses at top and bottom of shells which

leads to introduction of factor known as shear correction factor. This was due to linear assumption of

transverse shear stress along the thickness. But the main problem with the shear correction factor is that it

depends upon a number of factors such as material properties, lamination scheme, geometrical properties etc.

However, using FSDT, thin (ratio of thickness to representative dimensions is 1/20 or less) and moderately





thin (thickness smaller by at least one order of magnitude when compared with other shell parameters say

radii or thickness is at least 1/10 of smaller length of shell) laminated shells can be analysed satisfactorily.

However, exact behaviour of composite shells cannot be predicted using FSDT especially in case of thick

shells. Therefore, taking higher order variation of transverse stress along the thickness of shell (as linear in

case of FSDT) good interpretation of behaviour can be worked out.

Using FSDT and HSDT theories, analysis of thin and moderately thick composite shells can be

performed well. Results of FSDT when compared with those of HSDT, only a slight variation in results was

observed at the expense of increase in computational efforts. This increase in computational efforts led the

researches to develop further theories which can able to predict stresses more accurately especially in the

vicinity of the interfaces, surfaces, geometrical non-linearities etc.

Using ESL theories, analysis of LCs can be worked out in a good manner but the same model cannot be

used in different cases. This gave need to the development of layer-wise theories (LWT) or discrete layer

formulations. This was done by including some enhanced warping functions during the formulations. Such

approach can be justified by removing the use of C1 requirements in both ESL and LWT models. The main

advantage of this is that the rapid change in slopes at the interfaces can easily be plotted along the thickness.

In ESL models, the number of unknowns does not depend upon the number of layers as in case of LWT

models because in LWT formulations, each layer is treated individually. Some researchers used LWT

models along with ESL models.

2. MATERIAL PROPERTIES

The behaviour of laminated composites, sandwich (singly layered, multi layered) and FGMs can be

studied only by assuming the variation of material properties across the thickness of the structure. The

formulations used by the researchers for modelling the variation of material properties are:

Single layered homogenous shells In case of homogenous shells, material properties are independent of the coordinates and are constant throughout the thickness. The material property, mij(ζ), for homogenous structure can be written as

ijm constant

where, ζ is the mid surface global thickness coordinate.

Multi-layered homogenous shells Material properties for multi-layered homogenous shells are assumed to be layer wise Heaviside functions and given by

( )

1

1

( ) ( )lN

m

ij ij m m

m

m m H H

where, H(ζ) is Heaviside function, ζm-1 and ζm are global thickness coordinate measured from mid surface to top and bottom surface of mth layer.

3. REVIEW OF SHELL THEORIES

Brischetto [1] carried out the analysis of the approximation of the curvature term in the free vibration of

one-layered and multi-layered isotropic, composite and sandwich cylindrical and spherical shell. He used the

three-dimensional exact solution for shell structure in the framework of layer-wise approach. He solved the

differential equation by means of exponential matrix method and found that the curvature approximation is

valid for thin and shallow shells. Structures including sandwich configuration show bigger error because of

their bigger transverse anisotropy. In the case of cylindrical and spherical shell panels, there is dependence

on the half-wave number but it is not a priori predictable. The error also depends on the considered vibration

mode. In general, the approximation of the curvature terms does not give important errors in the case of in-

plane vibration modes.

Wali et al. [2] presented a numerical model for the free vibrations of 3-dimensional functionally graded

material shell structure based on a discrete double directors shell element. They adopted the higher order

shear deformation theory to formulate the theoritical model. They also assumed that the material properties

are changing in thickness direction according to general four-parameter power law distribution in terms of

volume fractions of constituents. The accuracy and the efficiency of 3D-shell model to predict the free

vibration behavior of FGM shell structures were demonstrated by comparing the present results with those

available in literatures. By examining the Eigen-frequencies of the FGM shell structures, they explained that



the lumped mass matrix is tested for their ability to accurately model the free vibration behavior. The

performance and the accuracy of the present DDDSM with lumped mass matrix will be checked in dynamic

behavior of FGM shells in future works. Xiang et al. [3] presented an efficient solution method based on the Haar wavelet to study the free

vibration behaviors of composite laminated conical and cylindrical shells under different boundary

condition. They adopted the first order shear deformation shell theory to formulate the theoritical model.

They expressed the displacement and rotational fields as the products of Fourier series along the

circumferential direction and Haar Wavelet series and their integral for the meridional direction. In this study

they found that the numerical results are very close with previously published results in literature. The

boundary conditions are satisfied exactly and the computational cost is comparatively low. The advantages

of this method are its simplicity, fast convergence and excellent accuracy. Jin et al. [4] proposed a simple accurate numerical procedure based on Haar wavelet discretization

method (HWDM) to the free vibration analysis of composite laminated cylindrical shells subjected to various

boundary condition. They adopted the Reissner-Naghdi's shell theory to formulate the theoretical model.

They first converted the initial partial differential equation into system of ordinal differential equation and

then the discretization of governing equations and corresponding boundary conditions are implemented by

using HWDM. By comparison and convergence studies excellent accuracy and low computational cost were

found. The effect of several aspects including boundary conditions, length to radius ratios, lamination

schemes and elastic modulus ratios on the natural frequency parameter were found.

Gao et al. [5] studied the vibrations of composite laminated structure elements of revolution subjected to

general elastic restraints including cylindrical, conical and spherical shells. They studied and analyzed the

model based on the first order shear deformation theory and a modified Fourier series method. In this study

regardless of boundary condition, each displacement and rotation components of the structure elements were

expressed as the superposition of a standard Fourier cosine series and two supplementary function. The

general elastic restraints of the structure element were accounted for by using the artificial spring technique

in this analysis. The accuracy and convergence of the modified Fourier series solution are presented by

numerical example. In the comparison, good agreements were obtained.

Mohammadi and Sedaghati [6] presented the nonlinear vibration analysis of sandwich shell structures

with a constrained electro rheological fluid core layer. They used the finite element modeling based on

assumed strain functions for discretizing the sandwich shell structure. According to the experimental data

available in literature, the results show that the developed finite element modeling approach leads to more

accurate results compared with the hierarchical finite element modeling for large displacements. Their

parametric studies on nonlinear vibration damping behavior for different boundary condition showed the

hardening type in the nonlinear behavior of the sandwich panel in which the natural frequency increases with

increase in amplitude.

Nguyen-Van et al. [7] proposed the numerical analysis of free vibration of laminated composite shell

structures of various shapes, span to thickness ratios, boundary conditions and lay-up sequence. The method

was based on novel four-node quadrilateral element within the framework of first order shear deformation

theory. In this study several numerical investigations were conducted and the results obtained were in

excellent agreement with other available numerical and analytic solutions. They also found that the present

element is relatively simple but yields slightly better accuracy for thin to thick laminated shells with various

boundary conditions, modulus ratios and stacking sequence. Since in this method the integration is done on

the element boundaries for the bending and membrane term, the present element remained accurate even

when it is highly distorted.

Zheng et al. [8] presented the spectral collocation method based on integrated orthogonal polynomials

rather than conventional differentiation was applied to the free vibration analysis of coupled axisymmetric

laminated shell structures with arbitrary elastic support boundary condition. They firstly divided the coupled

shell structure into its multiple components (cone, cylinder and sphere) at the location of junction in the

meridional direction. Then they applied the Hamilton’s principle and the equations of motion for all the

individual shell segments were derived on the basis of first order shear deformation theory. They proved that

the numerical results are in high agreement with existing solution in the literature and very good accuracy

and stability have been found. It was also found that the accurate natural frequencies of a coupled laminated

shell can be obtained by using a small number of collocating points and the computation cost is considerably

low.

In this paper Kouchakzadeh and Shakouri [9] deal with vibrational behavior of two joined cross-ply

laminated conical shell. They investigated the natural frequencies and mode shapes. Using thin walled

shallow shell theory of Donnell type and Hamilton’s principle, the governing equations were developed.



They solved the equations assuming trigonometric response in circumferential and series solution in

meridional directions. The first natural frequency of joined shell and corresponding circumferential mode

number increases when two semi –vertex angles get close to each other. It was observed that the fundamental

frequency of joined shell increases with increase in shell thickness. By increasing the number of layers in

constant thickness they got the small effect on natural frequencies of joined shell when more than four layers

were used in cross-ply lamination. They obtained the maximum value of first natural frequency when the

first semi-vertex angle is slightly greater than the second one.

Kumar et al. [10] presented the free vibration response of laminated composite and sandwich shell by

using an efficient 2-D finite element model based on higher order zigzag theory. They proposed finite

element model satisfies the inter-laminar shear stress continuity at the interfaces in addition to higher order

theory features, hence most suitable to model sandwich shells along with composite shells. All the three radii

of curvatures were included in present formulation. Numerical results of vibration responses for different

features of laminated composite and sandwich shells such as boundary condition, ply orientations, thickness

radius and curvature shows that the proposed 2D FE model is capable to predict results very close to 3D

elasticity solutions for laminated composite and sandwich shells. The presented model was more accurate,

especially for sandwich shells, as it incorporates trans-verse shear stress continuity at each layer interface

besides higher order theory features.

A solution of the free vibration problem formulated for the cantilever composite cylindrical shell has

been obtained by Lopatin and Morozov [11]. The governing variation equations have been derived based on

the Hamilton’s principle and solved using the generalized Galerkin method. They derived the two variants of

the formula (with and without taking into account the axial component of the inertia force) for calculations

of the fundamental frequency. The calculations based on the analytical solution were verified by the finite-

element analysis. Parametric analysis had been performed for the shells with various geometry and elastic

characteristics. It has been demonstrated that the approach developed in this work can be successfully

employed for rapid, reliable and accurate calculations of the fundamental frequency and would facilitate

design analysis of the shells under consideration.

A unified formulation was developed for free vibration analysis of circular cylindrical double-shell

structure with general boundary condition by Zhang et al[12]. In this study the displacement components

were expanded using Fourier series for cylindrical shell regardless of boundary condition. To improve the

accuracy and convergence of Fourier series several supplementary functions were added. The numerical

results of free vibration analysis of the double-shell structure coupled with annular plates were conducted to

check the convergence and accuracy of the present method. A variety of extra vibration results for the

double-shell structure with various boundary conditions were given, which may serve as benchmark results

for validating new computational methods in the future.

Mai-Duy et al. [13] explained the free vibration analysis of composite shell structure of various shapes,

modulus ratios, span to thickness ratios, and boundary condition and lay-up sequence by a novel smoothed

quadrilateral flat element. They developed the element by incorporating a strain smoothening technique into

a flat shell approach. The flat element formulation was adequately accurate and stable in all test cases which

was in contrast to general trend to use curved higher order finite elements analysis of shells. The results

obtained were in excellent agreement with those present in literature. In this it was observed that the present

element is relatively simple but yields good accuracy for many thin to moderately thick laminated shells

without shear locking. The present element remained accurate even with badly-shaped elements because

integration was done on the element boundaries for bending membrane and geometric terms.

The generalized Differential Quadrature Method was used to study the free vibration analysis of

functionally graded conical and cylindrical shells by Tornabene [14]. He adopted the first order shear

deformation theory for analysis. He developed the treatment within the theory of linear elasticity when the

materials were assumed to be isotropic and homogeneous through thickness direction. He obtained the

vibrational results without the modal expansion methodology. After analysis, the complete revolution shells

are obtained as special cases of shell panels by satisfying the kinematic and physical compatibility. The

GDQ method provided converging results for all the cases as the number of grid points increases. The

analysis provides information about the dynamic response of conical, cylindrical shell structures for different

proportions of the ceramic and metal. It can be pointed out from the analysis that the frequency of vibration

of functionally graded shells and panels of revolution depends on the type of vibration mode, thickness,

power-law distribution, power-law exponent and the curvature of the structure.

Panda and Singh [15] proposed a nonlinear finite element model for geometrically large amplitude free

vibration analysis of doubly curved composite spherical shell panel using higher order shear deformation



theory. They introduced the nonlinearity in Green-Lagrange sense. They used the variation approach to

develop the governing equation of the vibrated spherical shell panel. The effects of the curvature, thickness,

and vibration amplitude, modular ratio, stacking sequence, lamination scheme and different support

conditions on the frequency ratio were examined. The frequency ratio is more pronounced when amplitude

ratio increases. It was also observed that the spherical panel depicted hardening type spring behavior with

increase in amplitude ratio and softening type spring behavior with stacking sequence, thickness and

curvature ratios.

Kumar et al. [16] presented the free vibration analysis of laminated composite skew hyper shells using a

finite element formulation based on higher order shear deformation theory. They included the effect of cross

curvature in the formulation. They used the isoparametric finite element in the model consists of nine node

with seven nodal unknown per node. The results employing present formulation were in excellent agreement

with those available in literature. They observed that for a given skew angle, the fundamental frequencies

exhibit very less change when symmetric units of angle and cross ply laminations were repeated. It was also

observed that with increase in skew angle, the fundamental frequency increases except in CFCF shell. With

increase in skew angle, the amplitude of mode shape corresponding to fundamental mode decreases.

Lopatin and Morozov [17] formulated the free vibration problem for the composite laminated circular

cylindrical shell with clamped-clamped ends. They solved the problem on the basis of theory of laminated

shells taking into account the transverse shear strains averaged over the wall thickness. The equations of

motion were solved using the Galerkin method and the formula was derived for fundamental frequency. The

results of the calculations obtained for the shells with various geometry parameters and laminated structure

of composite material were verified by comparison with finite-element solutions. In the design analysis of

laminated composite cylindrical shells, the analytical approach developed in this work can be used for rapid

and accurate calculations of fundamental frequency.

CONCLUSION

After the detailed literature survey, following points were noted down:

1. Use of shear correction factor in FSDT and HSDT theories limit their applications in the analysis. ZZT

and HOZT does not include the use of such correction factors.

2. Frequencies worked out using FSDT and HSDT were more accurate for thin shells as compared to thick

shells whereas, HOZT was able to predict good results even for thick shells.

3. The element type should be carefully chosen during the finite element analysis of the shells.

4. During the use of finite element softwares, the convergence study should be carried out accurately.

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A Review on the Analysis of Beam-Column Joint under

Seismic Load

Mohammad Firoz Khan 1, H. D. Chalak

2

1 M. Tech. Scholar, Department of Civil Engineering, NIT Kurukshetra, [email protected] 2 Assistant Professor, Department of Civil Engineering, NIT Kurukshetra, [email protected]

ABSTRACT

Beam-column joint is one of the most critical element in the structural especially when it is subjected to

any kind of lateral load (seismic load, impact load, wind load etc.). The failure of beam-column joint results

in the progressive failure of the structure if not designed properly. This factor is prominent for the buildings

located in high seismic zones.

In this paper, an outline is presented on the past work carried out on the failure analysis of a beam-

column joint both experimentally and analytically. It was seen that the reinforcement and its placement near

the face of joint plays an important role in determining the strength. However, the work on the analysis of

the strengthened joint under lateral conditions is very limited. Also, it depends upon the type of method

chosen for carrying out the strengthening of the joint.

Key Words: Beam-Column joint, Seismic analysis, Strengthened beam-column joint.

1. INTRODUCTION

Over the past some decades, intensive research work has been done on the reinforced concrete beam

column joints. In high seismic areas, strengthening of improper designed reinforced concrete structures

represent critical issue involving technical as well as social aspects. Because such type of RC structures were

designed for gravity loads only not for seismic load or high wind load/lateral loads. Basically, columns

having minimum cross sectional area and longitudinal reinforcement not able to satisfy shear demand and

flexure generated during earthquake.

Construction of strong beam-weak column under seismic loads, it may lead to formation of local

hinges in the column. The associated failure mode characterized by catastrophic and brittle structural failure.

Inadequately detailed reinforced concrete beam-column joints, especially exterior joints, identified as the

most critical joints that may fail due to excess amount of shear stresses. In interior joints where rebar are

improperly anchored, bond failure in longitudinal reinforcement has been also observed.

Strength of reinforced concrete beam-column joint plays a significant role in the performance of any RC

frame structure especially for the cases when subjected to large lateral load. The lateral loads can be in the

form of the seismic load, wind load, impact or ballast load. Inadequately detailed reinforced concrete beam-

column joints, especially exterior joints, identified as the most critical joints that may fail due to excess

amount of shear stresses. In interior joints where rebar are improperly anchored, bond failure in longitudinal

reinforcement has been also observed. In RC building, portion of columns that are common to beams at their

intersection are called beam-column joints. Beam-column connection may be crucial regions in RC frames

designed for high seismic attack. Beam-column connection can be classified as follows:

According to geometrical configuration Interior beam-column joints

Exterior beam-column joints

Corner beam-column joints

According to loading condition and structural behavior

For static loading

Earthquake and blast loading





In philosophy of seismic design, the designing of beam- column joint is basically based on strong

column and weak beam concept, the formation of plastic hinge are expected to form in the beam near about

the column face and develop the flexural strength beyond the design strength. The development of high

internal forces at plastic hinges may lead to critical bond condition in the reinforcing longitudinal bars

passing through joint region and required high shear demand at the joint core. The development of bond

strength may affects the shear mechanism at the joint.

Failure of joint in shear may occur before the formation of plastic hinges in the beam.

Failure of joint in shear may occur after the formation of plastic hinges in the beam.

Due to excess slippage of the bar, bond failure of the longitudinal reinforcement may occur.

LITERATURE REVIEW

1) Uma and Prasad [11] presented a review of the postulated theories related to beam column joint behavior

subjected to seismic loading. They discussed the general behavior of common types of joints in reinforced

concrete moment resisting frames. The mechanisms involved in joint performance with respect to bond

and shear transfer were critically reviewed and discussed in detail. The design of shear reinforcement and

its detailing aspects were also studied. It was reported that the significant amount of ductility could be

developed in a structure with a well-designed beam column joint where the structural members can act

satisfactorily according to guidelines of designing.

2) Somma et al. [8] presented that in the design of new buildings, modern seismic codes prevent the failure

of the beam column joint through design capacity approaches. However the failure of the beams is not the

only kind of failure which jeopardizes the building. Numerous research models have been proposed in the

past representing the seismic behavior of the beam column joint but no clear consensus on identifying their

modes of failure. They provided a complete method of identifying connection failure mechanism. They

concluded that failure of beam column joint is not dependent on one parameter considering shear stress

parameter or percentage of transverse reinforcement, for determination of failure modes, mechanism

contributing to the joint resistance need to be considered.

3) Yan and Du [3] presented the comparison of precast and cast-in-situ beam joints under seismic loading.

To study the seismic performance of both type of joints, four joint specimens were produced including two

specimens of precast joints and two specimens of cast in situ joints. the axial compression ratio of the

joints was adopted as the main variable in their study and analysis was carried out on the basis of observed

joint failure modes and relationship carried out from test data such as hysteresis curve, skeleton curves,

sleeve joint strain curve. It was noticed that precast joint feature a relatively concentrated crack distribution

in which the limited number of cracks was distributed throughout the plastic zone of beam. Cast-in-place

joint feature more evenly distributed cracks in the plastic zone.

4) Melo et al. [5] carried out study on cyclic behavior of interior beam-column joints reinforced with plain

bars. They found out that seismic damage of the beam column joint of the building, built with plain bars

and without proper detailing needs further study of the behavior of the structure. They presented the result

of the cyclic test carried out on six beam column joints built with plain bars, these specimen represented

the existing RCC structures. For comparison an additional specimen was built with deformed bars and

tested. the loading were applied to these specimens and results were carried out for comparison.

5) Bahrami et al. [6] carried out the study of a new moment resisting connection of beam to precast concrete

column during lateral loading through application of nonlinear finite element analysis ABAQUS. The

precast column was connected to the beam with the help of (i) inverted E (bolted connection) (ii) box

section (welded connection). Connection response associated with stiffness, ductility, lateral stiffness were

compared to a reference monolithic connection. Achieved lateral resistance, lateral stiffness and ductility

of the proposed section was approximately 98%, 80% and 80% respectively. This showed that

performance of the proposed connection was nearly of equal performance to that of monolithic connection.

6) Li et al. [7] carried out study on the seismic behavior of the beam column joint strengthened by

ferrocement composites. Ferrocement is proposed to protect the joint region through replacing concrete

cover. Six exterior beam column joints including two control specimen and four strengthened specimens



are prepared and tested under constant axial load and cyclic loading. Experimental results have indicated

that using ferrocement composites as strengthening material has improved the effectiveness of the beam

column joint in terms of peak horizontal load, energy dissipation, stiffness and shear strength.

7) Decker et al. [9] carried out seismic investigation of interior reinforced concrete sand-lightweight concrete

beam-column joints. Sand -lightweight concrete is 20%lighter than its normal weight counterpart based

upon coarse lightweight aggregate. The effect of seismic forces are directly proportional to the mass, so

these prove better under seismic conditions but on the other hand they are quite brittle in nature. The tests

were carried out on the specimens and it was observed that if the designing and detailing were done

according to codal provisions and joint shear was kept within a reasonable limits, these could perform

similar to normal built concrete specimen.

8) Al-Salloum, Almusallam [10] studied the efficiency and effectiveness of carbon fiber reinforced polymers

(CFRP) in upgrading the shear strength and Ductility of seismically deficient beam column joint. For this

purpose they took 4 samples, two baseline specimens (control specimen) and two strengthened specimen

(which were strengthened with CFRP sheets). The CFRP sheets were bonded to the joint and a part of

column region. All these four specimens were subjected to cyclic loading equivalent to severe earthquake

damage. The damaged specimens were repaired by filling epoxy and wrapping them with CFRP sheets.

These repaired specimens again were subjected to cyclic loading and their response were obtained and

compared. The comparison showed that CFRP sheets had increased the shear resistance of joints and

increased the ductility of the beam column joint.

9) Kremmyda et al. [2] did the numerical investigation of the resistance of precast RC pinned beam column

connections under shear loading. In their research, they proposed an analytical expression for prediction of

resistance of precast pinned connections under shear monotonic and cyclic loading. The proposed formula

addressed the case where the failure of the connection occurs with simultaneous flexure failure of the

dowel and compression failure of the concrete around dowel, expected to occur when (i) adequate concrete

cover of dovels is provided or (ii) adequate confining reinforcement is foreseen around the dowels in the

case of small concrete covers. The expression was calibrated against the available experimental data and

numerical results derived from nonlinear numerical investigation.in addition to this, recommendations for

design of precast pinned beam column connections were given, especially when connections were utilized

in earthquake resistant structure.

10) Yang et al. [1] carried out seismic load tests on RCC beam column sandwich joints with strengthening

measures. The study was done to on sandwich joints. 6 specimens were taken and subjected to cyclic

loading to check their performances. Tested specimen were consist of 1 specimen with additional vertical

dowel bar, 2 specimen with additional diagonal bars and 1 specimen with additional later beams, compared

with 2 specimen without strengthening measures. The comparison was carried out in resistance behavior,

deflection, performance, ductility, energy dissipation. Based on these results, the effect and mechanical

behavior of strengthening measures were analyzed.

11) Girgin et al. [4] carried experimental study on cyclic behavior of precast hybrid beam column joint

connections with welded components. This study was carried out to meet up the need of improved beam

column connections to transfer the cyclic load effects between structural elements. Beam bottom

longitudinal rebar’s were welded to beam end plates while top longitudinal rebar’s were placed to

designate gaps in joint panels before casting of topping concrete in the connection. The tests were carried

out on 6 specimens including 1 monolithic and five precast hybrid precast hybrid half scale specimens

representing interior beam-column connections of a moment frame of high ductility level. It was observed

that maximum strain developed in the beam bottom flexural reinforcement plays an important role in the

overall behavior of the connections.



CONCLUSION

From the literature survey, following important points were noted down:

1. The strength of the joint primarily depends upon the placement of the reinforcement in the

vicinity of the joint.

2. The distance between the shear stirrups should be adequate enough so that concrete can be

easily placed between them.

3. The strengthening element used on the joint should be elastic so that the ductility of the joint

can be maintained under lateral loading.

4. The analysis using analytical methods primarily depends upon the mesh choice. Hence, the

element type and its size should be chosen properly.

REFERENCES

[1]Yang. ZH, Li. YM, Liu. JW, Seismic load test on reinforced concrete beam-column sandwich joints with strengthening measures,4

th International Conference On Advances In Experimental Mechanics.

[2] Kremmyda. GD, Fahjan. YM, Psycharis. IN, Tsoukantas. SG, Numerical investigation of the resistance of precast RC pinned beam-to-column joint under shear loading, Earthquake Engineering And Structural Dynamics, Wiley publishing inc.

[3] Liu. HT, Yan. QS, Du. XL, Seismic performance comparison between precast beam joints and cast-in-place beam joints, Advances In Structural Engineering, Sage Publication Inc.

[4] Girgin. SC, Misir. IS, Kahraman. S, Experimental cyclic behaviour of precast hybrid beam-column connections with welded components, Int J Concr Struct M.

[5] Melo. J, Varum. H, Rossetto. T, Cyclic behaviour of interior beamcolumn joints reinforced with plain bars, Earthquake engineering & Sructural Dynamics,Wiley-Blackwell.

[6]S.Bahrami, M.Madhkhan,F.Shirmohammadi, Nima Nazemi, Behaviour of two new moment resisting precast beam to column connections subjected to lateral loading,Eng Struct.

[7] Li. B, Lam. ESS, Wu. B, Wang. YY, Seismic behaviour of reinforced concrete exterior beam-column joint

strengthened by Ferro cement composites, Earthquake And Structures, Techno Press.

[8] Somma.G, Pieretto.A, Rossetto.T, Grant.D.N, A new approach to evaluate failure behavior of reinforced

concrete beam-column connections under seismic loads,15WCEE LISBOA.

[9] Decker.CL, Issa.MA, Meyer.KF, Seismic investigation of interior reinforced concrete sand-lightweight concrete

beam-column joints, ACI STRUCT J.

[10] Al-Salloum.Y.A, Almusallam.T.H, Seismic response of interior RC beam-column joints upgraded with FRP

Sheets, J. compos. constr., 2007.

[11] Uma. S. R, Prasad.A.M, Seismic behaviour of beam column joints in reinforced concrete moment resisting frames,

Document No. :: IITK-GSDMA-EQ31-V1.0, IITK-GSDMA Project on Building Codes.



Seismic Analysis of Water Tank

Tanuja khati1, Smita Kaloni

2, Shashi Narayan

3

1 M.Tech Student, Department of Civil Engineering, National Institute of Technology Uttarakhand-246174,

[email protected], contact number-7500194794 2 Assistant Professor, Department of Civil Engineering, National Institute of Technology Uttarakhand-246174.

Email- [email protected] 3 Assistant Professor, Department of Civil Engineering, National Institute of Technology Uttarakhand-246174.

[email protected]

ABSTRACT

Water storage tank is one of the important lifeline structures in the society. Elevated water tank containing

liquid is a complex problem. Therefore the aim of the designer is to design the structure in such a way that it

should be capable of resisting all forces. To eliminate leakage water tank should be designed as a crack free

structure. However during an extreme loading event like earthquake, water tanks are subjected to lateral

forces and failures may occur in the system due to sloshing effects. The main aim of the study is to

understand the seismic behavior of elevated reinforced concrete water tank in earthquake prone areas. For

analysis Elevated circular water tank has been chosen.

Keywords: Elevated water tank, seismic analysis, impulsive mass and convective mass.

INTRODUCTION

Liquid storage tanks are considered as one of the important lifeline structures in the society. Many

engineering fluids like petroleum, chemicals etc are stored in tanks. In case of event like earthquake safety of

water tank against horizontal ground motion has become a issue of concern. It is important to handle the

seismic demand as the population is increasing day by day, so the water demand does. There are various

types of water tank according to their shape ,use, demand, material, location of water tank. On the basis of

location of water tank, water tank can be underground, ground supported, elevated, overhead etc. On the

basis of shape, water tank can be circular, rectangular, intze, conical dome etc. Horizontal ground motion

due to earthquake causes sloshing in water tank. Sloshing is a phenomenon in which liquid moves irregularly

in the container. For sloshing to occur liquid must have free surface which is only possible in partially filled

tank. In empty or completely full tank sloshing will not occur. Sufficient freeboard is required to prevent the

damage caused to roof or top of wall of tank. In the past decades destruction caused by earthquake to water

tank has increased. During 1960 chile earthquake magnitude 8.5, caused shear failure of beams of water

tank. During 1980 El-Asnam earthquake 7.2 caused torsion failure. In fig.1 water tank in chobri was half full

at the time of earthquake so sloshing could have been as one of the reason of destruction of tank. Water tank

in morbi (Gujrat) with a capacity of 5000 L was empty at the time of earthquake so no sloshing would have

generated.

Fig.1. Damaged water tank in chobri (Gujrat)

Housner (1963) investigated the seismic response of ground supported tank as well as an elevated tank as a

case study. The partially filled tank was modeled as two degree of freedom system and designed empty and

full water tank as single degree of freedom system. The liquid in water tank an be divided in to two types.

Liquid in upper region is called convective mass and generates convective hydrodynamic pressure and

oscillates water in the container. Therefore the liquid in the tank undergoes acceleration. Second is impulsive

mass which is in the lower region mass and generates impulsive hydrodynamic pressure. Praveen K.

Malhotra (2006) presented a method of analyzing sloshing in cone and dome roof tanks. To prevent the

sloshing in water tank sufficient freeboard is required so that roof slab can be prevented from resisting the

pressure. Also in sufficient freeboard causes increases in impulsive mass. Praveen K. Malhotra (2010)





investigated the seismic response on ground supported cylindrical tank. A flexible steel and concrete water

tank was modeled with rigid foundation and analyzed the behavior of impulsive and convective

hydrodynamic pressure in the water tank. Base shear, maximum sloshing height and time period were

calculated mathematically. Thomas (2016) compared the seismic response of elevated circular water tank

and square water tank of same capacity. The performance of water tank with diagonal bracing was also

studied.

In this paper elevated circular water tank is modeled based on method proposed in Indian Standard (IITK-

GSDMA). The hydrodynamic pressure, time period, sloshing effect on the tank is presented in this paper.

The effects of afore mentioned quantity is also presented.

METHODOLOGY

The water tank is modeled as a spring mss model as shown in fig. 2 In an event like earthquake horizontal

ground motion are generated which causes horizontal acceleration and the entire liquid of the water tank

goes under it.The liquid in the upper region of the water tank is known as convective mass which is

connected to tank wall by springs. This convective mass generates convective hydrodynamic pressure in the

tank. Convective hydrodynamic pressure generates oscillations which excites the liquid in the water and that

exert hydrodynamic pressure in tank. If a tank does not have sufficient freeboard then sloshing will occur.

Convective mass is responsible for sloshing due to insufficient freeboard. Convective mass is present only in

partially filled tank.The second one is called impulsive mass which is present in the lower region of the

water tank and rigidly connected to water tank.

Fig.2. spring mass model of elevated water tank

Impuslive mass will generate impulsive hydrodynamic pressure. Empty or full filled water tank will have

only impulsive mass and do not generate convective hydrodynamic pressure.

Step 1: Determine the weight of different components of water tank.

Step 2 : Calculate the c.g of empty container from the bottom of staging.

Step 3: Determine the parameters of spring mass model i.e. (mi, mc, hi, hi*, hc, hc*).

Step 4: Compute the lateral stiffness of staging.

kpanel=12𝐸𝑐𝐼𝑐𝑁𝑐

ℎ3 [𝐸𝑏𝐼𝑏

𝐿𝐸𝑏𝐼𝑏

𝐿+2(

𝐸𝑐𝐼𝑐ℎ

)] for intermediate panels (1)

kpanel=12𝐸𝑐𝐼𝑐𝑁𝑐

ℎ3 [𝐸𝑏𝐼𝑏

𝐿𝐸𝑏𝐼𝑏

𝐿+(

𝐸𝑐𝐼𝑐ℎ

)] for top and bottom panels (2)

Step 5 :Compute the impulsive and convective time period for water tank.

𝑇𝑐 = 𝐶𝑐√𝐷

𝑔 (3)

Cc= coefficient of time period for convective mode

D = Inner diameter of circular tank

g =Acceleration due to gravity



𝑇𝑖 = 2𝜋√mi+ms

ks (4)

Step 6 : Compute design horizontal seismic coefficient for impulsive & convective mode.

(Ah)i=ZI (Sa/g)/ (2R) (5)

Step 7: Compute base shear (V) at the bottom of staging for elevated water tank in impulsive &

convective mode

Vi=(Ah)i ×(mi+ms)×g, Vc=(Ah)c×(mi+ms)×g & 𝑉 = √𝑉𝑖2 + 𝑉𝑐

2 (6)

Step 8: Compute base moment in impulsive and convective mode.

𝑀 = √(𝑀𝑖∗2 + 𝑀𝑐

∗2) (7)

Step 9: Compute hydrodynamic pressure on wall (Pw) and base slab (Pb) in impulsive & convective mode.

Piw = Qiw(y)(Ah)i ϼghcosϕ (8)

Qiw = 0.866 [1 − (y/h)2]tanh(0.866(D/h) (9)

Pib = 0.866(Ah)iρgh sinh(0.866x/h)/cosh(0.866D/h) (10) Pcw = Qcw(y)(Ah)cρghD(1 − 1/3cos2ϕ)cosϕ (11) Qcw(y) = 0.5625cosh(3.674 × y/D)/cosh(3.674 × h/D) (12) Pcb = Qcb(x)(Ah)cρgD (13)

Qcb(x) = 1.125((x

D) − (

4

3) (x/D)3)sech(3.674 × h/D) (14)

where, (Ah)c =Design horizontal seismic coefficient for convective mode

(Ah)i=Design horizontal seismic coefficient for impulsive mode

D= Inner diameter of circular tank

g = Acceleration due to gravity

h= Maximum depth of liquid

hc =height of convective mass above bottom of tank wall ( without considering base pressure )

Ks= Lateral stiffness of elevated tank Staging

L= Inside length of rectangular tank parallel to the direction of seismic force

M* =Total overturning moment at base

Mc* = Overturning moment in convective mode at the base

Mi*= Overturning moment in impulsive mode at the base

pcb= Convective hydrodynamic pressure on tank base

pcw =Convective hydrodynamic pressure on tank wall

pib =Impulsive hydrodynamic pressure on tank base

piw= Impulsive hydrodynamic pressure on tank wall

V =Total base shear

x = horizontal distance in the direction of seismic force, of a point on base slab from the reference axis at the

center of tank.

y =Vertical distance of a point on tank wall from the bottom of tank wall

RESULTS AND DISCUSSION

A RC circular water container of 50,000 L capacity having internal diameter of 6 m and height of 2 m

(including freeboard of 0.5 m) is modeled. It is supported on RC staging consisting of 8 circular columns of

450 mm dia with horizontal bracings of 300x600 mm at three levels. Tank is located on hard soil in seismic

zone III. Grade of staging concrete and steel are M20 and Fe415, respectively. Density of concrete is 25

kN/m3.the size of different components is illustrated in table 1. Dynamic property of the model is given in

table 2. Modal property for partially filled and empty condition of the tank is given in table 3. Hydrodynamic

pressures on wall and base slab for convective and impulsive mode is given in table 4. Pressure during

impulsive mode is higher than that of convective mode. Time period, total base shear and overturning

moment for partially filled tank is larger than that of empty condition.



Preliminary Data:

Table 1 Size and weight of different component of water tank.

Components Size(mm)

Roof Slab 120 thick

Wall 250

Floor Slab 150

Floor Beams 400x600

Shaft 250

Bottom Cone 300

Table 2 Spring mass parameters.

parameters mi(kg) mc(kg) ms(kg) hi(m) hi*(m) hc

*(m) hc(m) hcg(m)

Data 12212.85 28327 101469.24 0.843 0.333 2.367 0.804 14.76

Where, mc=convective mass of liquid, mi=Impulsive mass of liquid, hi=height of impulsive mass above

bottom of tank wall ( without considering base pressure), hs= Structural height of staging, measured from top

of foundation to the bottom of container wall, hc*= height of convective mass above bottom of tank wall

(considering base pressure), hi*= height of impulsive mass above bottom of tank wall (considering base

pressure), hcg=height of center of gravity of the empty container of elevated tank, measured from base of

staging.

Table 3 Comparison of different tank condition

Tank condition Time period(s) Total base shear (kN) Overturning moment (kN-m)

Partially filled 0.20 201 2962.6

Empty 0.19 179.2 2644.6

Table 4 Hydrodynamic pressures on wall and base slab

Hydrodynamic pressure on wall Impulsive mode

Convective mode

2.28kN/m2

0.764kN/m2

Hydrodynamic pressure on base slab Impulsive mode

Convective mode

2.15kN/m2

0.769kN/m2

Pressure due to wall inertia -

0.18kN/m2

Pressure due to vertical excitation At base of wall

At top of wall

1.77 kN/m2

0 kN/m2

Maximum hydrodynamic pressure - 3.12 kN/m2

Hydrostatic pressure - 14.7 kN/m2

Maximum sloshing wave height - 0.4347m



Fig.3. Insufficient freeboard in water tank

Table.5. Different parameters with insufficient freeboard

df(m) xf(m) ϴ mi(kg) mc(kg) Ti(s) Tc(s)

0.01 5.11 0.66 40412.80 127.00 0.36 0.20

0.02 4.37 1.69 39953.50 586.20 0.36 0.43

0.05 3.78 2.59 39311.20 1228.60 0.36 0.62

0.07 3.40 3.26 38727.50 1812.40 0.36 0.75

0.10 3.05 3.94 38063.10 2476.70 0.35 0.88

0.11 2.87 4.17 37750.40 2789.50 0.35 0.93

0.12 2.77 4.39 37501.25 3038.50 0.35 0.97

0.15 2.48 4.93 36834.43 3705.36 0.35 1.07

0.18 2.24 5.38 36211.59 4328.30 0.34 1.16

0.20 2.03 5.82 35608.27 4931.53 0.34 1.24

0.22 1.85 6.17 35066.10 5473.77 0.34 1.31

0.25 1.65 6.58 34401.60 6138.20 0.34 1.38

0.30 1.26 7.23 33212.12 7327.68 0.33 1.51

0.33 1.10 7.58 32572.20 7967.63 0.33 1.58

0.35 0.90 7.84 31946.90 8592.90 0.32 1.64

0.38 0.77 8.19 31343.50 9196.25 0.32 1.69

0.43 0.45 8.71 30156.48 10383.32 0.31 1.80

0.45 0.31 9.02 29492.10 11047.70 0.31 1.85

0.48 0.14 9.22 28915.60 11624.20 0.31 1.90

0.50 0.00 9.46 28327.00 12212.80 0.30 1.95

Where df= actual freeboard, xf= wetted width of roof, θ= angle of freeboard, mi=impulsive mass

mc=convective mass, Ti=impulsive time period, Tc=convective period.

The model is designed with both cases of sufficient and insufficient freeboard to prevent the damages caused

to roof slab due to pressure. In many cases sufficient freeboard is not provided to avoid the unused space in

water tank. Different spring mass parameters have been calculated with reference to IITK- GSDMA. Results

shows the comparison of different parameters like total base shear, overturning moment, time period

between partially filled tank condition and empty condition. Different condition of actual freeboard less than

required freeboard has been taken.Fig,4, Fig5, Fig6, Fig.7 shows the effect of convective and impulsive

hydrodynamic pressure on base slab and wall of water tank. Fig.8 shows the relationship between convective

mass, impulsive mass and time period. Fig .9 shows the relationship between base shear and freeboard.



Fig.4. Convective hydrodynamic pressure on base slab. Fig.5. Convective hydrodynamic on wall

Fig.6. Impulsive hydrodynamic pressure on base slab Fig.7 Impulsive hydrodynamic pressure on wall

Fig.8. Mass v/s time period Fig.9. Base shear v/s freeboard

CONCLUSION The following conclusions have been drawn out from the trend of results.

Total base shear, overturning moment of partially filled is greater than empty tank condition.

Convective hydrodynamic pressure at top of wall is more than convective hydrodynamic pressure at

top of base of slab.

Impulsive hydrodynamic pressure on wall decreases from base of wall to top of wall(excluding

freeboard).

Impulsive hydrodynamic pressure and convective hydrodynamic pressure on base slab increases as

the horizontal distance( in the direction of seismic force), from the reference axis at the center of

tank increases. There will be no convective and impulsive hydrodynamic pressure at the centre of

tank.

Free surface angle θ with respect to horizontal increases as the width of wetted roof slab

decreases. Maximum θ will occur when actual freeboard is greater than or equal to required

freeboard.

The smaller the actual freeboard is, the smaller the convective mass and the larger the impulsive

mass will be. Insufficient freeboard causes increase in impulsive mass.

The smaller the time period is, the smaller the convective mass and impulsive mass will be.



REFRENCES

1. IITK-GSDMA Guidelines for Seismic Design of Liquid Storage Tanks Provision with commentary

and explanatory examples (2007). NICEE, IIT Kanpur.

2. IS 1893-2002 (Part-I) Criteria for Earthquake Resistant Design of Structure – Part-1, General

Provisions and buildings, Bureau of Indian Standards, New Delhi.

3. Housner GW. Dynamic behavior of water tanks. Bull Seismol Soc Am 1963;53:381–7.

4. MALHOTRA, P.; WENK, T.; and WIELAND, M. Simple procedure for seismic analysis of liquid-

storage tanks. J. Struct. Eng. International, IABSE, 10(3), 2000, pp.197–201.

5. Gaikwad, M.V.(2013).“Seismic performance of Circular Elevated Water Tank with Framed

Staging”. International Journal of advanced research in Engineering and Technology, 4(4), 159-167.

6. Jaiswal, O.R., Jain, S.K. (2005). “Modified proposed provisions for a seismic design of liquid

storage tanks: Part I – codal provisions”. Journal of Structural Engineering, 32(3), 195-206.

7. Housner GW. Dynamic behavior of water tanks. Bull Seismol Soc Am 1963;53:381–7.

8. Durgesh C Rai, Performance of elevated tanks in Mw 7.7 Bhuj earthquake of January 26th, 2001.



Effect of Cement Content on Performance of Cold Mix

used for Constructing Flexible Pavement

Shivani Singh Dhriyan1, Yogesh Garg 2,

Sugandh Singh3

Asst. Professor,Graphic Era University, email -

[email protected]

MTech Student, NIT Patna, email-

[email protected]

MTech Student, Graphic Era University, email- [email protected]

ABSTRACT

The use of cold mix technology for the construction of roads is an environment friendly method.

The conventional hot bituminous mix produces huge amount of harmful gases which adversely affect the

nature by creating air pollution and health issues to the workers. Cold mix technology is of great use in

hilly regions as there is a problem of maintaining paving temperature in case of hot bitumen mixture. In

present study, the effect of cement on cold mix design has been studied. Various cold mix samples are

prepared using bitumen emulsion, aggregates and some another admixture. Comparative analysis is carried

out to know the effect of cement as admixtures on the properties such as Marshall Stability value flow

value, air voids of bituminous mixture sample containing different amount of admixture. The test result

indicates admixture i.e. cement increase in the strength of cold mix.

Key words: Sustainable Development, Cold Mix, Bitumen Emulsion, Marshall Stability, Flow value

1. INTRODUCTION

The population of India is increasing at a faster rate and at the same time natural resources are also on

the verge of extinction. Therefore, there is a great need of sustainable development of the country so that

we can gift a better place for our coming generation. The cold mix technology is one of the methods which

can be adopted for the construction of road without harming the nature. As we all know that hot mix of

bitumen produces huge amount of toxic gases which not only degrade the environment but have hazardous

effect on health of workers. Cold mix consists of bitumen emulsion and aggregates. No heating of binding

material is required therefore we can construct the road even during rainy season. There is one more

advantage of bitumen emulsion mixture that it eliminates the problem of maintaining paving temperature in

colder hilly regions. Thus, cold mix technology helps in reducing large amount of fuel which is require

melting the bitumen.

In the present study effect of cement content on cold mix has been investigated. According to Senior

Engineer (Research Institute of Highway, Ministry of Transport, Beijing, China) addition of cement to cold

mix increases the rate of break – up of bitumen emulsion and enhances the binding between aggregates and

asphalt. The strength of mix design and other properties were evaluated when it contains cement as filler

material and when it contains stone dust as filler material.

2. REQUIRED MATERIALS

For present investigation bitumen emulsion, aggregates of different gradation, Portland cement and

stone dust were used for preparing mixture. Bitumen emulsion is a binding material in which bitumen

particles are present in dispersed form in water. Emulsifier is added with water to facilitates breaking of

bitumen into minute particles and keeps it dispersed in suspension.When bitumen emulsion is taken out of

container for using it breaks i.e. aqueous phase and organic phase separates. In this study, slow

setting







bitumen emulsion has been used. Aggregates having good strength have been added to emulsion along with

cement. In other mix samples stone dust was added in place of cement as a filler material.

3. TESTS

Firstly, properties of aggregate used in mixture has been computed through various experiments.

Following tests were performed on aggregates.

(a) Los Angeles Test to find the hardness of aggregates.

(b) Compression test to find the strength of aggregates.

(c) Impact test to find the toughness of aggregates.

(d) Specific gravity test

(e) Shape test

(f) Sieve analysis to find fine and coarse aggregates.

Later, Marshall test was conducted on bitumen cold mix to find the strength, stability and optimum bitumen content for the mix. Numbers of specimen were prepared as per Marshall mix design.

Table 1: Gradation of Aggregate

Sieve Size (mm) Weight Retained (gram) 12.5 mm -10.0 mm 120 10.0 mm -4.75 mm 300 4.75 mm -2.36 mm 270 2.36 mm -600 micron 228 600micron -300 micron 66 300micron -150 micron 72 150micron - 75 micron 60 75micron - PAN 84 Binder content, % by weight of mix 6 - 14 %

Following steps are involved in preparing a Marshall Specimen (Yadav Om Prakash and Manjunath

K.R, 2012):

1) Approximately 1200 gram of the mixed aggregates and filler are taken and heated to a temperature of upto

60 degree centigrade for 10-15 minutes.

2) Now bitumen emulsion is mixed with aggregate and filler and mixing is done thoroughly for 3-4 minutes

as required.

3) Now this mixture is poured into pre-oiled Marshall Mould.

4) Compaction of this mixture is done by Marshall Method.

5) Compaction hammers by giving 50 blows.

6) After compaction, sample is extruded and kept it for 24 hours at room temperature.

7) After 24 hours, measure the dimensions of sample (height) and weight of sample. Now sample is

immersed in water bath at 35-40 degree centigrade for 40-60 minutes.

8) After completion of time sample is taken out and weight of the sample is noted. After taking weight,

keep the sample for 24 hours at room temperature.

9) After completing 24 hours, curing is done for 48 hours and after the curing period, test the sample in

Marshall Apparatus and determine stability value and flow value.



Un

it

Wei

gh

t o

f sp

ecim

e

n(g

/cc)

Flo

w

Val

ue

(mm

)

Mar

shal

l S

tab

ilit

y

Val

ue(

Kg

) A

ir

Vo

ids

(%)

4. RESULT

4.1 Sample with Stone Dust Used as Filler

Table 2: Dimension and Weight of Marshall Specimen of cold mix

Bitumen Emulsion (%)

Dry weight of sample (Kg)

Weight of sample in water (Kg)

Height (h) of sample in cm

Unit weight of sample (g/cc)

9 1.257 0.613 6.8 2.376 10 1.267 0.620 6.7 2.408 11 1.269 0.629 6.6 2.425 12 1.281 0.635 6.7 2.408

Table 3: Various Properties of Cold Mix

Bitumen (%)

Gt Value Gm Value Vv (%) Vb (%) VMA VFB Stability value (Kg)

Flowvalue (mm)

9 2.140 1.951 8.831 20.322 29.153 77.704 2016 8.5

10 2.087 1.958 6.181 24.475 30.656 79.832 2204 14

11 2.038 1.967 3.483 28.685 32.168 89.172 1952 17.3

12 1.993 1.967 1.034 32.783 33.817 96.942 1720 19

a) Gt Value: Theoretical specific gravity b) Gm Value: Bulk Specific Gravity c) VV: Air Voids

d) Vb: volume of bitumen e) VMA: Voids in mineral aggregates. f) VFB: Voids filled

Unit Weight vs Bitumen Content

2.43

2.42

2.41

2.4

2.39

2.38

2.37

4 6 8 10 12 14

Bitumen (%)

Stability Value vs Bitumen Content

2500

2000

1500

1000

500

0

4 6 8 10 12 14

Bitum

en (%)

Figure1: Variations in unit weight of bituminous mix Figure 2: Variations in stability value of bituminous mix

with increase in bitumen content with increase in bitumen content

Flow Value vs Bitumen Content

20

15

10

5

0

4 6 8 10 12 14

Bitum

en (%)

Air Voids vs Bitumen

10

8

6

4

2

0

4 6 8 10 12 14

Bitumen (%)

Figure3: Variation in flow value of bituminous mix with Figure 4: Variation in air voids of bituminous mix with

increase in bitumen content increase in bitumen content



Sta

bil

ity

V

alu

e (K

g)

Air

V

oid

s (%

)

Flo

w

Val

ue

(mm

) U

nit

Wei

gh

t

(g/c

c)

4.2. Bituminous cold mix sample with cement as filler

Table 4: Dimension and Weight Table

Bitumen (%) Dry weight of sample (Kg) Weight of sample in water (Kg)

Height (h) of sample in cm

Unit weight of sample (g/cc)

9 1.255 0.603 6.8 2.429 10 1.269 0.612 6.7 2.425 11 1.256 0.621 6.5 2.458 12 1.329 0.631 6.6 2.448

Table 5: Various Properties

Bitumen (%)

Gt Value Gm Value Vv (%) Vb (%) VMA VFB Stability value (Kg)

Flow value (mm)

9 2.164 1.924 11.090 20.041 31.131 64.376 2178 10.9 10 2.110 1.931 8.483 24.137 32.620 73.994 2885 11.0 11 2.059 1.948 5.390 28.408 33.796 84.052 2364 11.18 12 2.013 1.954 2.930 32.566 35.496 91.754 1980 13

3500

3000

2500

2000

1500

1000

500

0

Marshal Stability Value vs Bitumen

4 6 8 10 12 14

Bitumen (%)

2

.48 2

.46 2

.44 2

.42 2

.4 2

.38 2

.36 2 2

2

Unit Weight vs Bitumen

4 6 8 10 12 14

Bitumen (%)

Figure5: Variation in stability value of bituminous mix Figure6: Variation in unit weight of bituminous with increase in bitumen content mix with increase in bitumen content

Air Voids vs Bitumen

12

10

8

6

4

2

0

4 6 8 10 12 14

Bitumen (%)

1

3.5

13

12.5

12

11.5

11

10.5

Flow Value vs Bitumen

4 6 8 10 12 14

Bitumen (%)

Figure7: Variation in air voids of bituminous mix with Figure 8: Variation in Flow value of bituminous mix

increase in bitumen content with increase in bitumen content



5. CONCLUSIONS

From the above tests and results the following points can be concluded.

(i) The bitumen emulsion can be used as a main binding material in place of bitumen for

constructing flexible pavement.

(ii) The addition of cement content increases the strength of the cold mix.

(iii) The optimum binder content for cold mix with cement content is approximately 11 %.

REFERENCES

[1]Choudhary Rajan, Mondal Abhijit and Kaulgud Harshad, International Conference on Emerging Frontiers in Technology for Rural Area. 2012.

[2]Pundhir N.K.S, Grover Maj Shalinder Grover and Veeraragvan. Cold mix design of semi Dense Bituminous Concrete”, IRC, Indian Highway 2010.

[3]Yadav Om prakash and Manjunath K.R. 2012, Cold Mix Design of Semi Dense Bituminous Concrete, Journal of Mechanical and Civil Engineering. 1(6), 9-16.

[4]Bureau of Indian Standards. Bitumen emulsion for roads (cationic type) Specification (Second Revision). IS 8887:2004, March, 2004

[5] Ministry of Road Transport and Highways. “Specifications for road and bridge works”.

[6]Ministry of Road Transport and Highways (MoRTH 2001), “Specifications for Road and Bridge Works (Fourth Revision)”, Indian Roads Congress, New Delhi, Section 500, Bituminous cold mix, Clause 519.1., pp 227 -232.

[7]S.K. Khanna and C.E.G. Justo, Highway Material testing (Labouratory Manual), Nemchand and Bros, Roorkee 1997

[8]S.K. Khanna and C.E.G. Justo “Highway Engineering” 2005



Local scour around bridge pier in non-uniform

sediments

ANSHUL YADAV 1, BALDEV SETIA

2

1PG student, Department of Civil Engineering, NIT Kurukshetra, Haryana

[email protected] (Mob no. +91 9027418164)

2 Professor, Department of Civil Engineering, NIT Kurukshetra, Haryana

[email protected] (Mob no. +91 94162 20222)

ABSTRACT

Scouring of piers and abutment has been recognized as a major cause of failure of bridges over

waterways. According to a study conducted by Brandimarte et al (2012), it is estimated that 60% of bridge

failures result from scour and other hydraulic related issues. Local scour around bridge pier has been a

topic of research for the past few decades, and researchers have devoted a lot of time and attention towards

this issue as it is one of the major causes of bridge failures. Most of the studies in this direction have been

conducted with uniform sized sediments. However, practically no stream truly comprises of uniformly

sized sediments. Realizing the need for studies on non-uniform sediments so as to find the maximum scour

depth, a laboratory investigation has been planned and carried out in the hydraulics lab of civil engineering

department of NIT Kurukshetra. The effect of non-uniformity of sediments on scour depth and scouring

pattern in the uniform and unsteady flow environment is to be studied. The study deals with estimation and

analysis of local scour around bridge piers in non-uniform sediments. The sediments being used in the

study was collected from river Yamuna and non-uniformity was generated artificially by mixing the sand

collected from different locations in different proportions. The experiments are being conducted in uniform

flow supplemented by some non-uniformity in flow in the form of a hydrographic run. A critical review of

literature and results of initial part of the laboratory investigation are being presented in this paper.

Key Words: scour, pier foundation, non-uniform sediment, unsteady flow

1. INTRODUCTION

Scouring refers to the removal of sediments in a stream due to action of flowing water. When flow

occurs around a bridge pier, it undergoes a 3-D flow separation leading to the formation of horseshoe

vortex near the channel bed, which in turn increases the local shear stress, causing scour hole around the

bridge pier. The correct estimation of scour depth at a bridge pier is essential for efficient and safe design

of bridges. Most of the studies in the past have been carried out to predict the scour depth around the bridge

pier in uniformly sized sediments, while only a few studies pertain to non- uniform sized sediments. As the

sediments in all streams are non-uniformly sized, it is extremely essential to estimate the scour depth and

scouring pattern in non-uniform sediments.

Garde (1996) emphasized that a lot of work had been done on uniformly sized sediments, and the

researchers were required to devote some time towards non- uniformly or well graded sediments. The

current method of estimation of scour depth (Lacey-Inglis method) used in the design of bridge piers

sometimes gives excessive scour depth, which occurs after a very long time of design flow leading to the

uneconomical design of bridge piers. For economical and safe design of bridges, it is very essential to direct

the research towards the natural prevailing conditions of the stream (i.e. non-uniformly sized sediments). The

equilibrium scour depth in case of non-uniform sediments is found to be less than that in case of uniform

sized sediments due to the formation of a protective covering known as armouring. But at high velocities, the





effect of armouring vanishes and the equilibrium scour depth for non-uniform sediments is found to be equal

to that of uniform sediments (Melville and Chiew, 1989). But it has not been well established about the

velocity upto which armour layer becomes stable and act as a protective covering for the underneath

particles. The temporal variation of the stability of the armour layer has also not been well established in the

literature. Baker and RE (1986), Melville and Chiew (1989), Kothyari (2008) have significantly contributed

in the research in non-uniform sediments.

The main feature of non-uniform sediments is to form an armour coat which differentiates the non-

uniform sediments from uniform sediments. The armour layer mainly consists of the coarser particles which

protects the scouring of the underneath particles up to a certain limit. When the amount of sediment

entering in a channel reach is equal to that of leaving, an armour layer may form but is partially covered by

finer fractions of the sediment in transport (Chiew, 1991). Around the threshold condition v/vc≈1,

armouring occurs on the approach flow bed and at the base of the scour hole which helps in the reduction of

scour depth around bridge pier. But at high values where the flow is capable of eroding all sized sediments,

the non-uniformity of the sediments has only a minor effect on the scour depth. At low velocities the

armour layer forms a protective covering over the underlying sediments but at higher velocities armour

layer is destroyed and scour depths starts increasing up to the equilibrium scour depth (Baker, 1986). The

scour depth in live bed and clear water conditions were found to be similar for non-uniformly sized

sediments (Melville and Coleman, 2000). Even though, great studies have been made in understanding the

scour related phenomenon one still has to depend on empirical or semi empirical equations for scour depth,

developed primarily on the basis of laboratory data. To check their validity and accuracy, there is a strong

need to compare them with the field data (Kothyari, 2008)

Upon a review of the literature, there was a gap in the stability of the armour layer; it was found that the

corelation of the velocity and the stability of the armour layer was not well established. Therefore it is

expected that the armour layer becomes stable and acts as protective covering upto a velocity greater than

the incipient velocity of flow. The influence of non-uniformity of sediments on formation and stability of

armour layer has also not been depicted in the literature. Thus, it is essential to direct the research towards

non-uniform sediments with special reference to the armour layer.

2. EXPERIMENTAL WORK AND PROCEDURE

The experimental work reported herein was a part of major experimental programme to study the effect

of non-uniformity of sand on scour depth and scour related phenomenon. The experimentation work of the

present study is being carried out in the Fluid Mechanics laboratory of National Institute of Technology,

Kurukshetra. The experiments are being carried out in a recirculating flume of length(L), width(B) and

height(H) 15m, 0.4m, and 0.5m respectively with circular piers of diameter 25mm, 30mm, 40mm, 50mm,

65mm, 75 mm and 100 mm. The experimental work was divided into 3 phases and in each phase the non-

uniformity of sediments will be increased. In the first phase of the experimentation, the sediments being used

have a standard deviation (σg) of 1.86 with median sediment size (d50) as 0.22mm. The size of non-uniform

sediments varies in the range of 0.075mm to 4.75 mm. The effect of diameter of piers on scour depth was

studied by keeping the depth of flow and velocity as constant for one set of piers. The duration of the

experiment was kept as 5 hours to study the time scale variation of the scour depth. The scour depth readings

were taken after the duration of 5, 10, 20, 30 minutes, and then at 1hour interval up to 5 hours with the help

of a point gauge. The experiments were also carried out in discrete steps unsteady flow and non-uniform

flow in the form of hydrographic run. The results of the initial part of the experimental work are presented in

this paper as it is part of a major study and experimental work is still in progress. A typical definition sketch

indicating the schematic and symbols used in the experimental investigation and study are as shown in

Figure 1.



Figure 1: Definition sketch of symbols

In accordance with the guidelines for good experimentation, some range of the different parameters

has been suggested (Setia, 1997). However, the minimum and maximum values of different parameters

with the available flume of width 0.4m and available sediments of median sediment size d50 =0.22mm are

shown in table 1.

Table 1: Scheme of Experimentation

Parameters Maximum Minimum

Constriction ratio, (B/D) 16 4

Flow depth ratio, (h/D) 3.92 0.98

Diameter of pier/medain sediment size, 454 114

(D/d50)

Figure 2: A photographic view of recirculating Figure 3: A pier installed in the laboratory

flume in lab (Length 15m, width 0.4m, and flume showing scour hole and armouring

height 0.5m) at the base of scour hole

3. RESULTS

The results of the 5-hour test run representing the temporal variation of scour depth are presented in this

paper. The first reconnaissance of the figure shows the strong dependence of scour depth on the diameter or

size of cylindrical pier. The smallest of the diameters has the least scour depth and the biggest, the highest.

The variation of scour depth with the time is shown in Figure 3 and 4. From the graphs shown in Figure 3 &

4, it is observed that for smaller diameter piers rate of scour is more in the beginning and this rate of scour

decreases with time till it attains the maximum scour depth. However in larger diameter piers also rate of

scour decreases with time, but rate of decrease of scour is less as compared to smaller diameter piers. In

larger diameter piers progressive scour occurs, during the 5 hour run of the test it is observed that scour

depth was progressively increasing even after the 5 hours.



Figure 4: Temporal variation of scour depth (v=0.2121m/sec & h= 0.078cm)

Figure 5: Temporal variation of scour depth (v=0.2581m/sec & h=9.8cm)

However the non-dimensionalized values of the scour depth with the diameter of the pier shows reverse

trend with smallest diameter having highest value and largest diameter having the least as shown in Figure 5

and 6. This can be due to the dependence of scour depth on some other parameters also. The reverse nature

of non-dimensionalized scour depth shows that scour depth strongly depends on the diameter of the pier but

it depends on some other parameters also.

Figure 6: Non-dimensionalized scour depth v/s time Figure 7: Non-dimensionalized scour depth v/s time

(v=0.2121m/sec & h=0.078m) (v=0.2581m/sec & h=0.098m)



The 5-hour scour depth can be related with the diameter of the pier and velocity of flow using non-linear

regression by the equation given below:

Hs=18.6657V2.2349

D0.5044

Where,

Hs-scour depth from bed level in cms

D-diameter of the pier in mm

V-velocity of flow in m/sec

Fig.8 observed v/s predicted values of 5-hour scour depth

Unsteady flow

The test run is also done in unsteady flow conditions for 5-hour duration as per the ideal conditions of

the experimentation as suggested by various researchers. As per the ideal conditions of the experiment

40mm diameter pier was selected for the test run, the results of which are as:

0

0.01

0.02

0.03

30 60 90 120150180210240270300dis

char

ge

(cum

ecs)

time (minutes)

0

20

40

60

30 60 90 120150180210240270300

vel

oci

ty (

m/s

ec)

time (minutes)



From the above graphs, it can be concluded that an increase in the velocity of the flow results in the

corresponding increase in discharge, depth of flow, and scour depth. The scour depth continued to increase

as the velocity was increased continuously and sediments in suspension remained in suspension and no

deposition in scour hole took place, and erosion because of the descending flow leads to increased scour

depth.

At low velocity, a very few or no sediments were in motion and the armour bed composed of coarser

sediments and armour bed became stable throughout the channel bed and at the base of the scour hole

during the experimentation. But at higher velocity statistically more sediments were in motion leading to the

formation of ripples and dunes on the channel bed, the armour layer was still acting as a protective layer and

was visible in the troughs of the ripples and dunes on the channel bed. Armouring at the base of the scour

hole consists of coarser particles which helps in the reduction of the scour depth at higher velocities up to a

certain extent, but due to the action of downflow flow the underneath finer particles gets eroded and

transported through the voids of the armouring in the scour hole leading to the continuous increase in the

scour depth.

In accordance with the results reported in literature it is found that at low velocities the equilibrium scour

depth in case of non-uniform sediments is found to be less than that in case of uniform sized sediments. As

the non-uniformity of sediments increases, armouring influences the local scour greatly because the

coarseness of the armour increases. Also, for large gradation of sediment size, the armour peak velocity

increases and equilibrium scour depth is less for a constant relative mean velocity. In case of uniformly sized

sediments, the equilibrium scour depth in clear water is found to be 10% less than live bed scour as

concluded by many researchers as the bed material is transported from the upstream into the scour hole,

which is not true in case of non-uniformly sized sediments.

The properties of the sediments in scour hole were also found to be different from that of the rest of the

sediments. It was found that sediments in scour hole were composed of coarser sized particles with an

increase in the median size of the sediments. The non-uniformity of the sediments in the scour hole was also

observed to decrease and uniform coarser sized particles were found to be deposited in the scour hole.

4. CONCLUSIONS

As a prelude to the main work, a critical review of the literature existing on the subject has been carried

out. It has been found that mainly the investigations have been carried out on uniform sediments using

circular smooth piers having uniform cross-section throughout and under clear water conditions. But scour

on non-uniform sediments under live bed conditions is relatively lesser investigated. Out of the three aspects

of scour, the part dealing with prediction of scour is extensively investigated but the part dealing with

mechanism and protection deserve to be explored further. As evident from open literature, most researchers

have not considered change in temperature of the stream which can have some effect on the scour depth and

scouring pattern. The water qualities like salinity of water becomes important in case of offshore bridge

piers, the effect of which on scour phenomenon is yet to be explored in detail. It is expected that this study

will help in the efficient and safe design of the piers and will contribute towards the development of the

nation.

9 REFERENCES

[1] Chiew and Melville (1989), Local Scour at Bridge Pier with Non-uniform Sediments, Proc. Instn Civ. Engrs, Part 2, 1989, June,215-224.

[2] U.C. Kothyari, K.G. Ranga Raju, and R.J. Garde (1992), Local Scour around Cylindrical Bridge Piers, Journal of Hydraulic Research, 30:5, 701-715.

[3] Luigia Brandimarte, Paolo Paron, Giuliano Di Baldassarre, Bridge pier scour: A Review of Processes, Measurements, and Estimates, Environmental Engineering and Management Journal, May 2012, vol 11, No. 5, 975-989.

[4] Melville and Coleman (2000), Bridge Scour, Water Resources Publications LLC, ISBN Number: 1-887-201-18-1.

[5] B. Setia (2008), Equilibrium Scour Depth Time, Int. Conference on Water Resources, Hydraulics &



Hydrology(WHH ‘08), University of Cambridge, UK, feb.23-25, 2008.

[6] B. Setia (1997), Scour around Bridge Piers: Mechanism and Protection, Ph.D Thesis, Department of

Civil Engineering, Indian Institute of Technology, Kanpur, India.

[7] Y.M. Chiew (1991), Bed features in Non-Uniform Sediments, Journal of Hydraulic Engineering, vol.

117, No. 1, January, 1991.

[8] Baker (1986), Local scour at Bridge Piers in Non-Uniform Sediment, Highway Capacity Manual, 6th

edition, 91P-402.

[9] U.C. Kothyari (2008), Bridge Scour: Status and Research Challenges, ISH Journal of Hydraulic

Engineering, 14:1, 1-27.

[10] Y.M. Chiew & B.W. Melville(1987), Local Scour around bridge piers, Journal of Hydraulic Research,

25:1,15-26.



Numerical Modelling of Partially Restrained RCC

Rectangular Slabs

Priyam Sharma Assistant Professor, Civil Engineering Department, GLA University

[email protected]

ABSTRACT

This paper describes FE analysis of partially restrained rectangular slabs for some selected span ratio to

study the flexural behaviour of slabs. The research is solely based on the assumption that a slab may be

treated as beams spanning in two directions and simply supported edge and fixed edge of slabs are the two

cases of partial fixity. Finally results of FE analyses have been utilized to calculate the middle strip bending

moments which has been converted into bending moment coefficients.

Keywords: Finite element analysis, Rectangular slab

1. INTRODUCTION

In designing the rectangular slabs there are convenient design methods available to users to obtain the

slab strip moments (ACI 1963, IS 456:2000, BS 8110). Such methods are not available for analysis and

design of partially restrained rectangular slabs. From the support conditions rectangular slabs are classified

into nine categories (ACI 1963, IS 456:2000, BS 8110). In this research partial fixity is expressed as

percentage of fixity such that simply supported edge implies 0% fixity and fixed edge implies 100% fixity.

Now, in each of the category from the nine categories at each simply supported edge fixity is increased from

0% to 100% for span ratio 0.5 and for each percentage of fixity slabs are analysed and the obtained nodal

bending moments are converted into middle strip bending moment coefficients. Hence, this paper presents

graphs between coefficients and different percentage of fixity to compute the design moments with above

described support condition and for span ratio 0.5.

2.DESCRIPTION OF SELECTED PARAMETERS Span ratio:

“IS 456: 2000” describes the span ratio as longer span/shorter span but “ACI 318: 1963” describes the

span ratio as shorter span/ longer span. To validate the present model, the span ratio is selected according to

the “ACI 318: 1963” and is kept in the range of 0.5 to 1.0. To make the span ratio 0.5 length of shorter span is 3 m and length of longer span is 6 m.

Modelling of slab: To model slab and reinforcement “shell 63” element of ANSYS 14.0 is used as described by “Ahmed

and Chowdhary” (1999 a, b).

Optimum mesh size: No. of divisions in shorter direction is 6 and no. of divisions in longer direction is 12.

Support Continuity

Based on literature review and as per researcher’s knowledge there is the existence of some continuity at

the supports in masonry wall supported slabs. Since the instrumentation is not designed specifically to

investigate the continuity, the precise degree of rotational constraint cannot be identified, but it has been

estimated that the restraint would not be less than 10% nor more than 35% approximately. There are number

of ways by which the partial fixity or continuity can be incorporated into finite



element models of slabs. To some extent, the choice of a particular scheme is dependent on the finite

element code and whether or not the code will accept certain modifications of rotational spring

elements. Three different techniques could be used for including the effect of continuity and partial

fixity at the supports. The first approach is to add rotational spring elements at the boundary nodes and adjust the spring

stiffness correspond to the degree of fixity desired. Although spring elements are available in some codes (e.g. ANSYS etc.), they are not available in the entire computer codes and, consequently, two alternative schemes are also utilized.

The effect of a rotational restraint can also be simulated by adding three-dimensional beam elements attached at the boundary nodes and extending beyond the boundary point. The effective rotational stiffness can be prescribed by appropriate selection of the parameters E, I and L. This procedure of introducing support continuity is utilized in both the beam/slab and plate/slab models.

A final, convenient method of restraining rotation at the boundary nodes is to prescribe a moment reaction at the nodes where the magnitude of the applied moment is selected as a percentage of the fixed-end moment developed for the particular loading applied.

While any of the three methods for representing support continuity/ fixity can be used, with a

proper assignment of parameters, to yield satisfactory results. Each method has certain undesirable

features. For example, the use of spring elements is possible only with certain codes but it is likely the

most convenient procedure. The use of dummy beams to represent adjacent spans increases the

number of elements and nodes, although this approach seems physically rational and is intuitively

appealing. And the application of end moments to represent partial fixity is a convenient and rational

scheme but first requires the determination of fixed-end moments (complete fixity) for each loading

condition. Each of the three techniques for modelling fixity produced essentially identical effects on

response, and the choice of a particular method depends only on the preference of the analyst and the computer code available for use. Hence, the first approach is used in the present investigation for incorporating semi rigid connections as ANSYS computer program is having the facility for incorporating rotational spring.

3. MODELLING SUPPORT CONDITION AS PARTIAL FIXITY

To incorporate partial fixity at the simply supported edges in all the nine categories of slab (ACI

1963) zero length “combin 14” element is used. Method described by “M.E. Kartal” (2010) to incorporate partial fixity in frames have been followed considering slab as beams of 1000 mm width

spanning in two direction and hence the stiffness of the rotational spring ( ki, j ) is calculated by using the Equation (1) given by “M.E. Kartal” at different percentage of fixity and is presented in table 1

shown below

Where, “νij” is the fixity factor, which represents the connection percentage.

Table 1 Torsional stiffness of spring for different percentage of fixity Stiffness

% FIXITY Short span Long span 0.01 2.08E-11 1.04E-11

5 1.10E+08 54824561 15 3.68E+08 1.84E+08 25 6.94E+08 3.47E+08 35 1.12E+09 5.61E+08 45 1.70E+09 8.52E+08 55 2.55E+09 1.27E+09 65 3.87E+09 1.93E+09 75 6.25E+09 3.13E+09



85 1.18E+10 5.90E+09 95 3.96E+10 1.98E+10

99.99 2.08E+13 1.04E+13 In the developed model one end of the rotational spring is connected to the node at the edge of

the slab and the other end is connected to the node which is fixed also both the node coinsides (not

merged) with each other so that to have a proper working zero length rotational spring.

0

.05

0

.04

AC

I FE

Analysis

0

.03

0

.02

0

.01

0

0

.5 0

.6 0

.7 0

.8 0.

9 1

Figure 1: Ansys Model of Partially fixed slab

Figure 2: Variation of coefficient of short span positive

moment for case 9

Table 2 Input parameters of Numerical model

Modulus of

elasticity of Thickness (mm)

Unit weight Live load Poisson’s ratio

concrete Ec (N/mm3) (N/mm

2)

(N/mm2)

25000 = perimeter/180 =100 23.6×10-6

7×10-3

0.2

4.RESULTS OF FE ANALYSIS Verification of FE model with selected case:

As extensive literature on partially restrained rectangular slabs is not available, it is not possible

directly verify results of FE model for such slabs. But for the purpose of checking the accuracy of

incorporated semi rigid connections simply supported edge is 0.01% restrained in case 9 of ACI 1963 and

compared with the coefficients provided by the code for span ratio 0.5, 0.6, 0.75, 0.9 and 1.0. results are

found to be satisfactory as described by Ahmed and Chowdhary (1999 a).



Mo

men

t co

effi

cie

nt

Case 1

1.40E-02

1.20E-02 long span coeff.

1.00E-02

8.00E-03

6.00E-03

4.00E-03

2.00E-03

0.00E+00

0 10 20 30 40 50 60 70 80 90 100

Case 1

1.40E-02

1.20E-02 long span coeff.

1.00E-02

8.00E-03

6.00E-03

4.00E-03

2.00E-03

0.00E+00

0 10 20 30 40 50 60 70 80 90 100

Case 4 Case 4 0.014 0.1

0.012 0.09

LONG SPAN COEFF.

0.08

0.01 0.07

0.008 0.06

0.05

0.006

0.04

0.004 0.03

0.02

0.002

0.01

0 0

0 10 20 30 40 50 60 70 80 90 100

SHORT SPAN COEFF.

0 10 20 30 40 50 60 70 80 90 100

Case 6 Case 6

0.014 0.08

0.012 LONG SPAN COEFF.

0.07

0.01

0.06

0.008 0.05

0.04

0.006

0.03

0.004

0.02

0.002 0.01

0 10 20 30 40 50 60 70 80 90

0

0 100

Figure 3. Variation of Negative bending moment for different cases

SHORT SPAN COEFF.

0 10 20 30 40 50 60 70 80 90 100



Case 7 1.40E-02

1.20E-02 LONG SPAN COEFF.

1.00E-02

8.00E-03

6.00E-03

4.00E-03

2.00E-03

0.00E+00

0 10 20 30 40 50 60 70 80 90 100

Case 7

0.1

0.08 SHORT SPAN COEFF.

0.06

0.04

0.02

0

0 10 20 30 40 50 60 70 80 90 100

Case 3 Case 5

8.00E-02 1.40E-02

7.00E-02

Short span coeff.

1.20E-02

coef

fici

ent 6.00E-02

1.00E-02

5.00E-02

8.00E-03

4.00E-02

3.00E-02

6.00E-03

Mo

men

t

2.00E-02 4.00E-03

1.00E-02 2.00E-03

0.00E+00

0.00E+00

0 10 20 30 40 50 60 70 80 90

100

Case 8

8.00E-02 short span coeff.

1.40E-02

7.00E-02 1.20E-02

6.00E-02 1.00E-02

5.00E-02

8.00E-03

4.00E-02

6.00E-03

3.00E-02

4.00E-03

2.00E-02

2.00E-03

1.00E-02

0.00E+00

0.00E+00

0 10 20 30 40 50 60 70 80 90 100

Long span coeff.

0 10 20 30 40 50 60 70 80 90 100

Case 9

LONG SPAN COEFF.

0 10 20 30 40 50 60 70 80 90 100

Figure 4. Variation of Negative bending moment

for different cases



5.CONCLUSIONS 1. During the validation of the model with the incorporated semi rigid connections it is found that 0.01% fix

edge may be considered as simply supported edge and 99.99% fix edge may be considered as completely

fixed edge for the study of flexural behaviour of slabs. 2. For same percentage of fixity and for same span ratio case 4 and case 6, case 5 and case 9 gives same results

long span negative bending moment.

3. For same percentage of fixity and for same span ratio case 4 and case 8, case 5 and case 9 gives same results

short span negative bending moment.

4. This methodology of incorporating semi rigid connections may be adopted for any kind of thick plate, made

up of any material provided that approximate results are required to study the flexural behaviour of slabs.

REFERENCES 1. ACI Publication 318-95 (1995) “Building Code Requirement for Reinforced Concrete”, American Concrete Institute

(ACI) Detroit. 2. BS 8110-1: 1997 “Code of Practice for Design and Construction”, British Standard Institute (BSI) London. 3. IS 456: 2000 (2000) “Plain and Reinforced Concrete Code of Practice”, Bureau of Indian Standard (BIS) New Delhi.

4. M.E. Kartal, “Effects of Semi-Rigid Connection on Structural Responses”, Electronic Journal of Structural Engineering

2010. 5. B. Ahmed and S. R. Choudhury, “Simplified Deflection Method for Serviceability Deflection of edge supported slabs”, Journal

of Civil Engg., The Institution of Engg., Bangladesh, Vol. CE 27, No. 1, 1999. 6. Sharmin Reza Chowdhury, “An easy way to analyse octagonal slab”, 4th Annual Paper Meet and 1st Civil Engineering

Congress, Dhaka, Bangladesh, December 22-24, 2011 7. ANSYS 14.0, Operation Guide.



Regional Planning Approach for Sustainable Development

of Hilly Regions: A Case of Ratnagiri- Sindhudurg Region

Krishna Kumar Dhote 1,

1 Department of Architecture and Planning, MANIT, Bhopal, Professor, [email protected]

ABSTRACT

The sustainable development of any nation demands for holistic growth. Today, urbanization is seen

as synonym to development, upto some extent it is true also as cities are now realized as engines of

economic growth. But if look at the urbanization scenario it suggest that the development is lop sided i.e. the

urban agglomerations are swelling in sizes with overcrowding population, high density and often leading to

congestion and failure of civic services. These irrational development of urban centres results into eating up

of agricultural land, increased pressure on natural resources like water bodies, forests and green areas.

It has always been perpetual for planners to remove the rural urban imbalance. Strengthening economy of

rural areas will not only prevents migration and concentration of population in urban areas but will also

improve the civic infrastructure of rural areas and hinterlands. The regional planning approach for

development is seen as an approach to take care of these imbalances and as an tool to improve the economic

development of the region [1]. The nature, agriculture, forest, mineral ores and land is taken as resource. The

second major resource is human resource. The essence of Regional Planning is lies in using these resources

in a optimal manner to achieve sustainable development, development that will provide employment and

opportunities in hinterlands, boosts economy by providing adequate market for forest and agricultural

produce, social infrastructure to reduce dependency on urban areas, development of highways,

communication systems to promote tourism and industries[2]. The five dimensions of sustainability, which

are to be addressed for development, are economic, physical, environmental, social and cultural. The

development can only be termed as sustainable when these five dimensions are satisfied. The irony of the

situation is that industrialization and urbanization demands for land, which affects agricultural land and

forest, land, also the demand of water supply by industries and urban areas affects irrigation in rural areas.

Though the industries provide employment and job opportunities but at the cost of nature. Similarly the

social and cultural fabric of society, which is sustainable at local level, faces the challenges of dilution of

values. The major challenge before us in terms of climate change is result of this fast changing scenario of

utilization of natural resources and natural landscape.

The hilly regions have peculiarity of development on one hand it poses the challenges of technical

accessibility on other it has potential of natural resources and contributes to geography and climate of the

region [3]. In the present paper an attempt has been made to use regional planning approaches to harness the

potential of two districts of western ghat of Maharashtra state namely Ratnagiri and Sindhudurg region.

These two districts are separated from rest of state by the beautiful Sahayadri ranges, on other side they have

the ocean adding serenity to the region. The region, which forms the major part of Konkan, is known for its

service sector to Mumbai metropolitan area and is equally famous for the export quality mangoes,

agricultural produce and fisheries. In today's context it has been seen as counter magnet for development of

Mumbai city and also as an alternative tourist spot to Goa. The sustainability aspects of economy see a huge

potential whereas the environment and ecology envisages the threat of deforestation, pollution on beaches

and adverse effects of climate change. The present paper will present a rational approach to make utilization

of natural resources optimally to achieve holistic development proposal for the Ratnagiri-Sindhudurg region.

Key Words: Regional Planning, Sustainable Development and Hilly Regions.



1. REGIONAL PLANNING APPROACH IN INDIAN CONTEXT

In India after independence the industrialization has taken place in a significant manner, however the

agrarian base prevails in Indian economy. The diversity of culture, geography and economic activities poses

a challenge before Indian planners. Therefore, the regional planning in Indian context focuses on prevention

of rural-urban migration, development of hinterlands and rural areas to check economic disparity and

extension of physical and social infrastructure to remote locations. A careful examination of Indian situation

suggests that the classification of regions in India can be on the basis of economy, social, environmental and

administrative regions[4]. The planning regions can be delineated out of the above classifications or by

overlapping and amalgamating them. The economic region is demarcated on the basis of natural resources

and human stock to be used as resource. The social region is constituted on the basis of historicity, caste,

language and other identical cultural parameters. The administrative region is again on the basis of prevailing

administrative processes. The planning regions should be carved out of mutual actions and interactions of

physical and cultural characteristics.

In India the present case of Ratnagiri- Sindhudurg region, it lies in the Western Ghats of India.

Western Ghats is a major coastal area of Maharashtra with Sahayadri Mountain ranges n side and Arabian

coast on other. This region, which is also known as Konkan, includes Mumbai the metropolis and Goa the

international tourist destination of India. This belt is rich in natural resources. The administrative boundary

of Mumbai and Goa was excluded while delineating the region and only Ratnagiri and Sindhudurg are taken

as region for purpose of regional plan. This delineation is based on administrative setup, cultural uniformity

and presence of natural resources in the identified region. These two districts are providing service support to

adjoining metropolis but still their potential of natural resource remains unutilized. The detailed introduction

to the region delineated for study is elaborated further in subsequent sections.

2. INTRODUCTION TO RATNAGIRI – SINDHUDURG REGION

For administrative purposes, the district is divided into 3 sub-divisions Ratnagiri, Dapoli and Chiplun and 9

Tahsils and Sindhudurg is divided into 2 sub-divisions and 8 Tahsils. Earlier Sindhudurg district was part of

the Ratnagiri district. For administrative convenience and industrial and agricultural development Ratnagiri

district was divided into Ratnagiri and Sindhudurg. The Ratnagiri district lies between 16° 13’ to 18° 04’

North latitude and 73° 02’ to 73° 52’ East longitude on the Konkan strip along the West Coast of India. The

district has a North -South length of about 180 kms and an average East-West extension of 64 kms. Its

coastal length is about 167 kms. Sindhudurg district is situated between North 150.37' to 160.40' latitudes

and East 730.19' to 740.13' longitudes. It is bordered by Arabian Sea on the West, Sahyadri hill ranges and

Kolhapur district on the East, Ratnagiri district on the North. Goa State on the South and Belgaum district of

Karnataka State on the South East. Figure 1.00 shows the region selected for study.



Figure 1:00 Base map of Ratnagiri Sindhudurg Region.

2.1. Demographic Profile of the Region

According to census, the Total Population of Ratnagiri District was 1696777 in 2001 and in 2011 it is

1615069 .The Decadal growth rate comes out to be -4.82 %. Similarly the Sindhudurg District’s population

is 868825 as of 2001 and for 2011 it is 849651. There is decrease in 2011 population by 2.21% compared to

2001. And thus the growth rate is –0.02 [5]. The negative growth in turn infers the fall out to be lack of

services, infrastructure and opportunities in the region. Further if we compare the population density of these

two districts the maximum is 271 persons per hectare (pph) of Ratnagiri Taluka and minimum is 107 pph of

Vaibhawadi Taluka, the entire range of density is far below the average density of 365 pph of Maharashtra

state and 382 pph of national average. Figure 2:00 shows the population distribution in the region. The

uneven distribution again reflects non-uniform distribution of resources and employment opportunities.

Figure 2:00 Population Density map of Ratnagiri Sindhudurg Region.



2.2. Natural Wealth of the Region

Starting with the forests the total area under forest in the Ratnagiri district is about 6700 hectares, which

is about 0.82% of the district’s geographical area About 225 species of trees, 4 types of Bamboos and 15

types of grasses are recorded in the district. There are number of trees and shrubs which have medicinal

value. The Sindhudurg district is known for its dense forests. The total area under forest cover is estimated to

be around 38,000 hectares, covering 7.47% of the total geographical area. However, it is observed that the

process of deforestation is growing annually at the rate of 1.5% in the district, giving way to industrial and

residential growth.

Ratnagiri is one of the most important maritime districts in Maharashtra. Therefore, marine fishery is

naturally an important economic activity in the district. Fishing is done all along the coast, in the sea,

generally up to 65 kms from the coast. The adjoining Sindhudurg is gifted with a coastline of 121 kms with

1,600 sq. kms of conventional shelf. However, the activity of deep-sea fishing is not taken up to exploit the

marine fish catch potentials. With a view to improve the socio- economic condition of the fishermen and to

augment the fish supply, several developmental schemes are introduced by the Fisheries Department.

Of all the districts in Konkan Division, Ratnagiri district seems to be favourably placed with regard to

the availability of some of the important minerals. Manganese and Iron ores are found in the southern part of

Ratnagiri district. However the scenario of agriculture is very grim though the there are numerous streams of

water very few of them are usable for irrigation. The heavy rainfall and topographic conditions doesn’t

encourage large-scale irrigation dams or water bodies. The main food crops are rice and ragi. Interestingly 73

% of the food crops is covered by Mango and cashew nuts.

2.3. Potential for Development of Region

The natural wealth in form of forests, coastal areas, agro products, fisheries and lastly the natural beauty

makes it apt case for economic development by way of industrial and tourism growth. One major approach

for economic development would be setting up industries and tourist centres to harness the potential of the

region[6]. The proximity to Mumbai and Goa can be exploited by providing proper and adequate linkages.

The chronological evolution of Ratnagiri and Sindhudurg district indicates that these region is the birth

place of many great freedom fighters like Lokmanya Tilak and Babasaheb Ambedkar, it was ruled by

dynasties like Mauryas, Chalukyas, Portuguese and Marathas. These rulers had left their buildings as

testimony to time and have potential to be developed and heritage sites. There are locations which are of

importance with respect to religious tourism like Ganpati phule and parambhagvatas. Apart from these there

are sites with virgin sea beaches, hilly terrain and natural beauty, which are suitable for adventure sports and

eco-tourism.

As stated earlier both the districts have coastal line, which is used for fishing. A good number of

populations are involved in fishing using traditional and sustainable methods. In recent years fishing

companies had started intervening but rather than adding to productivity it has result into downfall in fish

production. This is probably because of unsustainable method adopted by them to catch fishes. This

mechanism is not only threat to local economy of fishing villages but also to the ecology of the coastal belt.

The judicious selection of technology and encouragement to traditional system will yield more production.

The Maharashtra government in its package scheme of incentives for industries 2013 has classified the

Industrial areas under various groups of A, B, C, D, D+ and no industry districts. Here A denotes the

developed industrial area and subsequently upto D that is least developed area in the hierarchical order.

Unfortunately in Ratnagiri and Sindhudurg region 13 talukas are under D+ category i.e. with no industry

districts, two under D means industrial area with least development and only two districts which includes

Ratnagiri and Chiplun in C category in Ratnagiri districts. The region has potential for resource-based

industries of agricultural, forest and mineral produce.



2.4. Development Issues and Concern of Region

The Ratnagiri Sindhudurg region lie at the heart of Konkan, the western coastal line of India seashore,

picturesque mountains and scenic natural beauty, and also famous for tropical fruit like the delicious golden

Alphonso mango, cashew, jackfruit, spice crops, coconut, areca nut and kokam. The region is bordered by

Sahayadri hills on the east and Arabian Sea on the west. It is a tract of high rainfall ranging between 3000 to

5000 mm a year and are one of the country’s water towers. The mangroves abut upon long stretches of

beaches and cover of the region is mangrove forest on the coast and tropical evergreen forest inland, with

stunted tree growth and a rich herbaceous flora on the wind swept plateaus. The fertile alluvial valleys

produce rice and coconut as the main crops; the hill slopes harbor mango and cashew nut orchards. The

estuaries and the coast support rich fisheries.

Being rich in biodiversity, ecologically sensitive and increased development pressure poses the the

following questions before us which need to be addressed for sustainable development

What are the effects of climate change on the biodiversity, forests and natural settings?

Whether the area is prone to disasters like cyclone, earthquake and floods?

Whether tourism and industrial development will have adverse effect on ecology and

environment?

Whether the development initiatives will boost local economy?

The local cultural ethos and social communities will be disturbed or enriched?

The answers and solutions to above issues raised will lead to a sustainable framework of solutions.

Regional planning approach has been envisaged as a tool to achieve rational solution. The subsequent section

examines the possibility of using regional planning for sustainable development.

3. REGIONAL PLANNING APPROACH

Urbanization in India is posing big challenges of uneven distribution of resources and opportunities

leading concentration of populations in megacities and urban areas whereas population s decreasing in small

and medium size towns. The rural poverty induced urbanization is resulting into urban poverty. Lack of

integral planning of rural and urban areas led to rural push. The traditional planning approach tries to

formulate an optimization to minimize the cost or risks. Policies are often unable to deliver the best. It

becomes imperative that economic planning is linked to spatial and regional planning, to cope up with

regional disparity and sustainable development at regional scale.

The aim of regional planning in Indian context should be to reduce spatial inequalities. The vast

geographical areas consists of significant environmental, economic and social variations poses the challenge

before planners and policy makers. In initial period after independence more emphasis was given on sectorial

than spatial and was focused on centralized urban centres and not on regions. Though the efforts were made

in first and second five year plans to raise the industrial as well as agricultural production to take care of

spatial inequalities regional planning remained focussed on few metropolitan cities. It was only in fifth five-

year plan where regional inequalities were taken care by formulation of policies at central level with respect

to resource transfer to backward regions. It can be concluded that in Indian context focus of regional

planning should be on decentralization of existing growth centres. Industrialization should be spread widely

to promote villages and small towns. Rearticulating investments should strengthen social capital and local

self-government system.

The Regional and Town Planning Act 1966 of Maharashtra empowers preparation of regional plan of the

regions in the manner in which it should indicate the land should be used for development thereon, stages of

development, the network of transportation and communication and conversion of natural resources for

development. The state government of Maharashtra has classified the state in 13 different regions Ratnagiri

Sindhudurg region is one amongst them.

4. DIMENSIONS OF SUSTAINABILTY

The fast depleting natural resources, dwindling water bodies and forests, out migration and social

imbalances and inequalities in economy draws our attention towards the five pillars of sustainability in an



integrated manner. The sustainability of the regional plan lies in economic development that respects the

environment and minimizes the negative consequences. The spatial development takes maximum advantage

of resources and maximizes environmental benefits and lastly to integrate sectorial goals with spatial for

balanced development of region. These five dimensions are physical, economical, environmental, social and

cultural. Various aspects of regional planning and sustainability are presented in Table 1.00.

Table 7 Sustainability Dimensions of Regional Plan

S.No Regional planning Aspect Dimension of

Sustainability Remarks

1 Allocation of land use as residential and industrial Physical Use of Agricultural and forest land to

be minimized

2 Allocation of land use as forest, or for mineral

exploitation Environmental

Forest land to be conserved and inning

activity should be done judiciously

3 Reservation of areas for open spaces, gardens,

recreation, zoological gardens, nature reserves,

animal sanctuaries, dairies and health resorts

Environmental Integrate with spatial planning for

adequate open spaces

4 Transport and communications, such as roads,

highways, railways, water-ways, canals and airports,

including their development

Physical and

economical

Optimization of routes for settlement

linkages, industrial and mineral

resources

5 Water supply, drainage, sewerage, sewage disposal Environmental

and Physical

Optimal use of natural resources with

minimum pollution

6 Reservation of sites for new towns, industrial estates

and any other large-scale development

Physical and

Economical

Establishment of new centres of

economic growth

7 Preservation, conservation and development of areas

of natural scenery, forest, wild life, natural resources,

and landscaping

Environmental,

Socio-cultural

and Economical

Awareness through tourism/ potential

for eco-tourism

8 Preservation of objects, features, structures or places

of historical, natural, architectural or scientific

interest and educational value

Socio-cultural

and Physical Will add to sense of belongingness

9 Prevention of erosion, provision for afforestation, or

reforestation improvement and redevelopment of

water front areas, rivers and lakes

Environmental,

Economical and

Physical

Ecological conservation, Suitability

analysis for spatial development

10 Proposals for irrigation, water supply and hydro-

electric works, flood control and prevention of river

pollution

Environmental,

Economical and

Physical

Small scale traditional methods to be

adopted

The five dimensions of sustainability are reflected in preparation of regional Plan of the region. Figure 3.00

represents schematic approach to achieve sustainability.

Figure 3:00 Regional Plan and Sustainability

Identification of

Growth Centres

Linkages

Provision of

infrastructure

Regional

Physica

l

Econo

micCc

Cultura

l

Environme

ntal

Social



5. RECCOMENDATIONS FOR SUSTAINABLE REGIONAL PLAN

The plan is envisaged not as an instrument to control the development but it should provide guideline for

outline development plan and comprehensive plans for smaller planning units at town, villages and taluka

levels. It is an attempt to consider major proposal at regional level, which will formulate policies for

economic and social development. Stress has been given on rationalization of disparities within the region,

integration of resource utilization, conservation of ecology and economic development of the region as a

whole.

5.1. Identification of Growth Centres

The spatial configuration of region shows that the urbanization is concentrated in only three pockets two

are coastal towns Ratnagiri, Sindhudurg and Chiplun is on midland between coastal area and Sahayadri

ranges. The demographic profile indicates negative growth rate of overall region, the sex ratio also suggest

that male population is lower that state and national average indicating out migration of male population in

search of employment. The physiography of the region demands for conservation of forestland and coastal

areas, the topography and climatic conditions doesn’t support large-scale irrigation projects; therefore small

parcels of agricultural land should be encouraged. The strategy of spatial distribution of population should be

to identify the potential of 17 talukas in terms of tourism, fisheries, mineral and agro-based industries. The

region is enclosed linearly with coastal area on one side and Sahayadri ranges on other therefore coastal

regulation and ecological aspects should be given due importance and the population distribution should be

encouraged on midlands.

The concentration of industries in region is again restricted to above three towns namely Ratnagiri,

Sindhudurg and Chiplun. Unfortunately all industries of the region falls into the C, D and D+ category of

Maharashtra implying low industrial development in the region. The region falls under highly eco sensitive

zone and demands only for high skilled and high technology oriented clean industries. Industries can be set

up in the region using existing transportation network of railway, roadway and water ways provided they are

equipped with effluent treatment plant, however energy and water intensive units should be discouraged.

There is scope of resource based industries (Agro, marine and forest), handicraft and cottage industries.

The region has huge potential for fishing as it stretches along the Arabian Sea. It is very sad that the

fishing production is declining in recent years. The downfall is production is because of techniques adopted

fishing companies using purse seine net as compared to gill net used by traditional fishermen[7]. The

technique adopted by companies are unethical, fishes that are of no use like juvenile fishes and pregnant

fishes which cannot be exported or consumed internally are being caught ultimately this practise is

destroying the marine ecosystem. Fishing industry has enormous potential for economic development of the

region. To encourage fishing certain policy measures are suggested first is direct marketing principle, the

agents or middlemen are taking significant amount of profit that are necessarily not the local people. Apart

from regulating sustainable traditional methods of fishing there is a strong need for creation of market

places, villages and towns connected through waterways and railway are apt cases.

Scenic beauty, places of historical and religious importance and sea beaches makes the region suitable

for tourism if connectivity and supporting infrastructure is provided. The proposed Mumbai Goa Highway

will connect number of archeologically sites, which mostly are sea forts built during different periods. The

forts, which owned by heirs, can be developed as heritage hotels under private partnership model. Presence

of tourist attraction like religious places, beaches, creek, forts, waterfalls and heritage can invite more tourist

and will enhance the economy of the region. The main challenges to promote tourism are inadequate

showcasing of tourism potential at national and international level, poor accessibility to many of the tourist

destinations and lack of wayside amenities. Figure 4.00 shows the potential growth centres with respect to

industrial and fisheries potential.



Industries Fisheries Tourism

Figure 4.00 Potential of Industry, Fishery and Tourism in Region

5.2. Identification of Environmentally Sensitive Zones

Floods Earthquake

Cyclone Landslide



Considering the physical features, the entire region can be divided into three parts coastal plains,

plateau strip and hilly areas of Sahayadri ranges. There is environmental degradation because of mining

activity and deforestation. At present 0.82% of Ratnagiri and 7.64% of Sindhudurg districts is under forest

area. The approach adopted for enhancement of forest area and biodiversity is to overlap maps of biological

richness, forest fragmentation, ecologically sensitive areas and animal corridors [8]. The areas which are

least occupied of above should only be used for development. And in order to conserve tem the share of

protected area should be increased, at present in Ratnagiri district there is no sanctuary or national park. Also

small-scale biodiversity hotspots using available species should be developed and incorporated in wild life

corridors.

Though the occurrence of disaster is rare the possibility of occurrence cannot be ignored. The

vulnerability of disaster specifically of floods, earthquake, cyclone and landsides are analysed and the area

prone to these disaster has been identified and development in these zones is limited. Figure 5.00 presents

the area vulnerable for the disasters.

5.3. Identification of Linkages

With the ports in the western coast are being developed, connectivity with Sindhudurg and Rantnagiri

districts would provide a boost to export. Lack of adequate railway network in the region force people to rely

on road transport, which is comparatively more expensive and time consuming. Due to this transport costs,

produce from the region becomes incompetent in market. The Konkan railway connects the region with

major cities of the country, National Highway 17 is developed along the coastal belt and provides good

transport facility. State bus services connect the main railways spine with settlements. The possibility of

waterway can be explored and it is difficult to construct roadways along coastline in light of coastal

regulation zone. The roads in rural areas are constructed under Pradhan mantra Gramin Sadak Yojana and

also the state government is launching Mukhya Mantri Gramin Sadak Yojana. Konkan has been a major

international trade centre with Rajapur and Harnai ports. Connectivity with major cities like Mumbai, Goa

and Pune should be improved.

Figure 5.00 Vulnerability Analysis for Natural Disaster



5. RECCOMENDATIONS FOR SUSTAINABLE REGIONAL PLAN

The Ratnagiri Sindhudurga region is endowed with scenic natural beauty, agriculture produces like

world famous tropical fruits Alphonso mangoes, cashew nuts, kokam and coconuts. The Sahayadri ranges

and coastal line provides home to rich biodiversity. Apart from natural resources it has rich culture and

human heritage. The places of religious pilgrimage, archaeological sites and tourism invite attraction from

domestic and international tourists. The challenge posed for developers and policy makers is to conserve

this environmentally sensitive zone with its complex nature of geography. The regional plan sets the strategy

of development of region, which can further be detailed out in outline development plan and comprehensive

development plan of smaller planning units may be rural settlements and towns. The development priorities

are set to check migration from rural areas, provide equal opportunity to rural hinterlands and equal

accessibility to urban and rural growth centres.

Before development the rich flora and fauna of the region, forests and water bodies need to be

conserved. This has been done carefully analysing the land surface utilization pattern. The sectors of

economy demands for land utilization leading to conflicts. The land under cultivation, marshy land, natural

cover and orchards has been reserved. The main criterion for land utilization is based on surface approach,

land capability, productivity and water management without compromising with socio-economic benefits

and ecological impact. The per capita land can be improved by converting barren land into usable land

without sacrificing the land under forest or cultivation.

The three ecosystems namely Coastal, Mountainous and Forest need to be conserved. The inter

sector conflict between mining & forestry, tourism & forestry, agriculture and urbanization needs to be

resolved by seeking right balance between compatible socioeconomic and natural developmental activity.

The growth centre identified for industries, agricultural market and tourism should be connected to optimize

transportation. The physical and social infrastructure will improve the overall quality of life. Thus the

different dimension of sustainability are often contradictory in nature but a rational approach considering the

drawbacks and merit of each, resolving the inter sectorial conflicts and a plan considering people’s

perception may lead to sustainable development of region.

6. REFERENCES

[1] Ward S. Planning and urban change. Sage; 2004 Feb 18.

[2] Dent, David, Olivier Dubois, and Barry Dalal-Clayton. Rural planning in developing countries:

supporting natural resource management and sustainable livelihoods. Routledge, 2013.

[3] Janssen, Willem, and Ali Kissi. "Planning and priority setting for regional research." Research

management guidelines 4 (1997).

[4] Sachs, Jeffrey D., Nirupam Bajpai, and Ananthi Ramiah. "Understanding regional economic growth in

India." Asian Economic Papers 1.3 (2002): 32-62.

[5] Grant, Ursula. "Urban economic growth and chronic poverty." (2006).

[6] Megu, Kangki. Development Issues in North-East Region. Mittal Publications, 2007.

[7] Korakandy, Ramakrishnan. Technological change and the development of marine fishing industry in

India: a case study of Kerala. Daya Books, 1994.



Physical land Suitability analysis for slum redevelopment: A

case of Jabalpur city

Dr. Neelam Soni1, Dr. Preeti Onkar

2 and Dr. Krishna Kumar Dhote

3

ABSTRACT

Land is scarce and is an important resource for any development. Slums are major concerns of today's

policies and programmes as they represent major challenges. Land occupied by slums is irrespective of its

suitability to hold the population. Land for housing the urban poor needs suitability assessment, a context

dependent concept defined by set of attributes of a site for identified purpose. The paper is based on

generating land suitability score for the identified slums of Jabalpur city in which most of the slums are

located on hilly areas and near water bodies. The paper explores the conceptual framework of land suitability

incorporating only the physical properties of land. The final output is in the form of ranking of slums on the

criteria of physical suitability of land using Analytical Hierarchy process.

Key Words: Slum redevelopment, Land suitability analysis, AHP, GIS.

1. INTRODUCTION

Rapid population growth requires additional lands for food production, housing, social, and physical

infrastructure, commercial and industrial use. However, like other natural resources, land is limited and man

tends to change existing land use to Land in urban area, on one hand, is a scare resource which needs to be

utilized appropriately in order to achieve balanced development and on the other hand

Land in urban area, on one hand, is a scare resource which needs to be utilized appropriately in order to

achieve balanced development and on the other hand, there is a very big need to supply land for housing the

poor. Land for housing the poor is thus becoming an insurmountable obstacle in the development facing the

growing cities while the development actions of many governments continue to focus on technical, financial

and administrative aspects of the housing problem, failing to act decisively on land issues or deliberately

avoiding or evading them wherever possible” (Angel et al, 1983).

Slums are areas of population concentrations developed in the absence of physical planning and lack

access to life essentials. Slums represent major national challenges in countries where they exist, especially

developing countries. Various intervention strategies can be adopted to upgrade and/or replace slums, but are

often faced with serious construction challenges, such as lack of access to sites and poor terrain conditions.

Moreover, during the execution of slum upgrading projects, resident families can experience significant

social and economic disruptions (Anwar & Aziz, 2014).

Slum redevelopment is not only relevant for planners and policy makers, but also for residents, property

owners, investors and citizen. Slum redevelopment is to improve urban appearances and inhabitant's

environments and enhance urban images and inhabitant's qualities of life. slum redevelopment is involved in

1 Contract Faculty, Department of Architecture and Planning, MANIT, Bhopal (MP), India

[email protected]

2 Assistant Professor, Department of Architecture and Planning ,MANIT, Bhopal (MP), India

[email protected]

3 Professor, Department of Architecture and Planning, MANIT, Bhopal (MP), India

[email protected]






not only the pure physical regeneration ,but also social regeneration associated with economic recovery,

community rebuilding and cultural.

In this respect ,land suitability analysis should be regarded as an important tool for sustained

development of slum redevelopment.

Presently running scheme of slum redevelopment (large scale) JNNURM and RAY slum free cities has

specified different models of slum redevelopment. The suitability of land remains a question for the policy

makers. The research would be an attempt to move towards providing physical solution in the form of model

incorporating all dimensions of slum redevelopment and land suitability analysis

Land suitability is a context-dependent concept defined by a set of desired attributes of an ideal site for

the intended purpose. Suitability assessment is the process of comparing desired attributes with actual

condition at a set of sites and then comparing suitability across sites. Since McHarg (1969) popularized the

application of suitability assessment in land use planning, it has become standard practice in both selecting

the best site for a particular use and choosing the use for which a site is most suitable. Instead, suitability is a

multicriteria evaluation in which experts define the most desirable attributes in terms of measurable factors,

the optimum values of those factors, and their relative importance weights (Jiang & Eastman, 2000).

Land suitability is a technique of quantifying the suitability of land for a propose development. As slums

are cause of social and economic phenomenon leading into environmental problem, attempt has been made

to redefined sustainability of land integrating social, environmental and cultural aspects with the physical

character/ properties of land. The identified parameter addressed in the framework incorporate all tangible

and intangible measures. The priorities of slum redevelopment strategies should incorporate the suitability of

land with respect to identified factor. (Dhote, Soni, & Onkar, 2013)

This study is to assess existing conditions of slums and identify interventions for improving suitability

for redevelopment of slum with desired services and infrastructure that can directly and indirectly affect

quality of life of residents. The paper explores the conceptual framework of land suitability incorporating

only the physical properties of land. The final output is in the form of ranking of slums on the criteria of

physical suitability of land using Analytical Hierarchy process.

2. LITERATURE REVIEW

The land suitability technique is used widely used to determine the fitness of the given piece of land for a

particular use. It has been used in urban planning and the GIS further reinforced with multi-criterion analysis

made this more useful. The parameters of land pertaining to slum redevelopment helps in first identifying the

problems and potential of existing slums and further gives direction for redevelopment. The optimal use of

land using land suitability analysis will inter weave the grey patches of urban slums in the city fabric. In

order to use the multi- criterion analysis of land, the parameters need to be identified and prioritize in order,

further they need to be weighed properly to achieve a rational solution (Dhote, Soni, & Onkar, 2013).

In Present scenario some method which are used are Boolean classification method, AHP modeling

method, Multi criteria analysis, weighting factor, fuzzy quantifiers; Ordered weighted averaging, Analytic

Network Process (ANP), Sensitivity analysis and Overlay analysis. The technique of land suitability analysis

has many identified methods for working the most compatible use of land. In today’s context the most

adopted method is GIS based multi criteria analysis. This method provides flexibility to incorporate the

tangible and intangible, the spatial and non-spatial data.

Remote sensing and GIS is used as technique in area of Integrated Evaluation of Urban Development

Suitability and is analyzed on the parameters like environment, water land resources and socio economic

development. It indicates that integrated evaluation of urban development could be conducted in an

operational way using remote sensing data, GIS spatial analysis technique and AHP modeling method.

(Dong, Zhuang, Xu, & Y, 2008) Multi criteria analysis is performed to evaluate development suitability of

the geo-environment for various land use categories, including High rise building, multi storey building, low

rise building waste disposal, natural conservation. Multi criteria analysis (AHP) analytical hierarchy process,

weighting factor for each urban land use. Are used with the parameter’s Topography, Ground conditions,

Groundwater and Geologic hazards. (Dai, Lee, & Zhang, 2001).

It has many other application as Environmental Impact Assessment of Land Use Planning (Jie, Jing,

Wang, & Shu-xia, 2010)in Wuhan City Based on Ecological Suitability Analysis, Site Suitability Evaluation



for Ecotourism Using GIS & AHP (Bunruamkaewa & Murayamaa, 2011).Housing Site Suitability

Assessment using GIS Based Multi criteria Approaches to. Geographical information system (GIS) and

Multi criteria analysis (MCA) with Analytical hierarchy process (AHP) are used. The system integrates two

major tools (GIS and AHP) in a manner that reach the correct solution to assist the decision maker in

determining appropriate values for the physical suitability criteria. The system was successfully tested in

determining the optimum land suitability for housing. (Al-shalabi, Mansor, & Nordin, 2006). A study on

Land suitability evaluation for development using a matter-element model demonstrated that matter-element

models provide much more information than fuzzy models.

3. STUDY AREA

The study area for score based land suitability analysis is the city of Jabalpur in Madhya Pradesh which

is class I million plus city of central India. There are 359 slums in Jabalpur, with a population of 2, 68,417 in

72,668 households. (As per Jabalpur Municipal Corporation) Jabalpur has highest slum population, that is,

about one- fifth of total slum residents of the state. The multi- functional nature of the city and its centrality

attracts a sizeable proportion of socio- economically weaker people not only from nearby villages of the state

but also from other adjacent and distant states of the country. Slums resides in steep hills of Jabalpur

4. MATERIALS AND METHODS

4.1 Selection of Slum

The study population of slums in Jabalpur comprises of 359 slums in various locations. A purposive

sample of 38 slums was selected for this study. The sample scheme was chosen to insure representation of all

categories of various land related influencing factors like typologies, land ownership, land value, location,

land use and land tenure ship. This accounts for almost 10% of the total number of slums in the city.

4.2 Preliminary Study

Preliminary study on the subject involved literature review on multiple dimensions including building

exhaustive inventory of various criteria, sub-criteria and measures for land suitability assessment,

identification of appropriate data sources and research on suitable land suitability analysis technique in

context of slums.

4.3 Primary and Secondary Data collection and Preparation

The Delphi technique is a widely used and accepted method for gathering data from respondents within

their domain of expertise. The technique is designed as a group communication process which aims to

achieve a convergence of opinion on a specific real-world issue. (Hsu & Sandford, 2007)It is an iterative

technique that generates both qualitative and quantitative data concerning collective judgment of

respondents. Here the inventory of criteria and indicators developed in (Soni et.,al. , 2013)for land

suitability score were redeveloped into a questionnaire and circulated to 15 experts in the field of town

planning including government authorities, academicians and practitioners. The experts were asked to give

ranking to various factors representative of land suitability indicators of physical parameters in initial round

of delphi.

In next iteration of delphi the experts were asked to give differential scoring to various sub criteria

represented under the pair wise comparison scheme where each sub criteria's level is represented by a matrix

of order equal to number of levels in that sub criteria. Various experts ranked values in the matrices

according to the scheme explained in Table1.0 in lines with the guidelines presented by (Saaty, 1977)(Saaty,

1980) (Saaty & Vargas, 2001)

Table 8 The preference scale for pair wise comparison in AHP

Scale Degree of preference Explanation

1 Equal importance Two activities contribute equally to the objective

3 Moderate importance of one factor

over another Experience and judgments slightly favour one level over another



5 Strong or essential importance Experience and judgments strongly favour one level over another

7 Very strong importance A level is favoured very strongly over another and dominance is

demonstrated in practice

9 Extreme importance The evidence favouring one level over another is of the highest

possible order of affirmation

2,4,6,8 Intermediate values between the

two adjacent judgments

When compromise is needed

Reciprocals Opposites Used for inverse comparison

Source: Saaty’s scale

The score assigned by various experts are assessed for consistency. Since almost 15 experts ranked the

order of the levels in expert opinion first round & numeric scores were almost similar in next round, the

principle of using median score (Hsu & Sandford, 2007)was used to derive appropriate score matrices for all

39 sub criteria of order equivalent to their levels. The overall score is computed as a weighed aggregate of

final scores computed as final level's score (computed by AHP) of sub-criteria with the weights of criteria

derived again using analytical hierarchy process. Fig.1 represents various sub-criteria levels and their scores

indicating relative importance as derived by median scores assessed by the experts. Similar pairwise matrices

were used for further calculations.

Figure 6 Level 2 AHP process of physical domain criteria’s

4.4 Indicator’s framework

The indicators finalized are categorized on the basis of physical properties of land for suitability

assessment. The following is the final list of inventory prepared for further analysis of data.

Table 9 Indicators for analysis

1.Inventory of physical domain

S.No. Parameter S.No. Parameter

1 Topography 10 Housing settled

2 Soil characteristics 11 Housing around other high risk zone

3 Geologic hazards 12 Water Supply

4 Type of construction 13 Toilets

5 Work place distance 14 Deficiency of In-house Connection

6 Availability of public transport 15 Deficiency of In-house Toilet

7 Physical Status 16 Road condition

8 Slum Location in(Land use pc) 17 Street light

9 Housing settled in geologically hazardous

zones

18 Drainage



Table 10 Factors, criteria and units used for land suitability analysis for slums

S.No. Factors Criteria Unit

1 Topography Elevation Meter

Slope Degree

2 Soil characteristics Soil characteristics Type of soil

3 Geologic hazards Natural disaster Size of scale

4 Type of construction Durability ( /stability ) Type of Material

5 Work place distance Distance to road Meter

6 Availability of public transport Distance to road Meter

7 Physical Status Physical feature(barrier) Land condition

8 Slum Location in(Land use pc) Land use Type of use

9 Housing settled in geologically hazardous

zones

Man made barrier Type of transportation

10 Housing settled Land form Physical feature

11 Housing around other high risk zone Man made/natural Physical barrier

12 Water Supply Source of water Municipal Tap

13 Toilets Type of construction Percentage

14 Deficiency of Inhouse Connection Capacity Percentage

15 Deficiency of Inhouse Toilet Capacity Percentage

16 Road condition Road surface material

17 Street light No. of light Percentage

18 Drainage Type of drainage Covered/semi coverd

4.5 Level Wise AHP

AHP is a widely used method in Multi Criteria Decision making and was introduced by (Saaty,

1977)(Saaty, 1980)(Saaty & Vargas, 2001)It is easily implemented as one of the MCDM techniques. AHP is

a decision support tool, which can be used to solve complex decision problems. It uses a multilevel

hierarchical structure of objectives, criteria, sub criteria and alternatives.(Arabinda, 2003) ((Baniya,

2008)The AHP has three basic steps. It begins by decomposing the overall goal (Suitability) into a number of

criteria and sub-criteria. The goal itself represents the top level of the hierarchy. Major criteria comprise

level two, sub-criteria make up level three, and so on (Duc, 2006)

AHP techniques as listed in flowchart was first applied to matrices of sub-criteria. Various metric

representations of scores of sub-criteria levels were computed and scores were standardized using

methodology suggested in (The Analytic Hierarchy Process (AHP Lesson 1) This is done by Multiplying

together the entries in each row of the matrix and then taking the nth root of that product. The consistent

eigen vectors are checked for standardization. To ensure the credibility of the relative significance, AHP

provides measures to determine inconsistency of judgments mathematically. Based on the properties of

reciprocal matrices, the consistency ratio (CR) was calculated and was measured against CR threshold

suggested by (Saaty, 1980) which is 0.10. More details of the CR calculation were given in (Ma et al, 2005)

and (Hossain et al, 2007) CR's were calculated for all matrices formed by various levels of all 39

parameters(sub criteria) and were found to be consistent with average CR value as 0.08 which is acceptable.

This gives us quantitative scores of physical characteristics existing in the slums of study area. Further,

similar process was adopted to compute weights for criteria.

5.0 RESULTS AND DISCUSSIONS The scores of individual criteria are computed for each 38 slums in the study. Table 4, represents the

labels of slums under study. Table 4 represent level wise scores of various criteria and sub criteria and scores

computed for various criteria heads

Table 4. Level wise scores –Physical criteria



Physical Criteria

Level 1 Level 2 Level 3

Criteria w1 Criteria w

1 Sub-Criteria S

ij

Physical 0.42 Topography 0.051402783 2.86 0.541527

5.71 0.381631

8.53 0.076842

Soil characteristics 0.045548501 Gravely sandy clay 0.336057

Sandy clay with variations 0.147646

Sandy Coarse 0.468359

Sandy silt Silty clay Silt &

sandy clay

0.047937

Geologic hazards

Type of construction 0.038786781 Kachcha house 0.20

Pakka house 0.80

Work place distance 0.093742432 10-11km 0.046033

1-2km 0.36767

3-4km 0.243663

4-5km 0.16766

7-8km 0.105558

8-9km 0.069415

Availability of

public transport

0.0721509 Far to main road 0.13755

Main road 0.513241

Near to main road 0.275101

very far to main raod 0.074108

Physical Status 0.123491843 Beside Nala 0.08544

Beside Mountain 0.085413

Beside Talab 0.085518

Mountain Area 0.085424

Near Big Transpotation 0.085498

Near Railway Line 0.056936

Near River 0.085534

Other Dangerous

&Objectionic

0.316416

Other Non Dangerous& Non

Objectionic

0.11382

Slum Location

in(Land use pc)

0.171327156 Agricultural Land 0.11088

Commercial 0.037456

Industrial 0.196435

Institutional 0.15592

Organizasional 0.128159

Other Pahadi Area 0.091138

Residential 0.280012

Housing settled in

geologically

hazardous zones

0.09528132 Flood Prone Area 0.20

Nill 0.80

Housing settled 0.079441888 Talab 0.285714

Nala 0.571429

Pahadi 0.142857

Housing around

other high risk zone

0.065724058 Nill 0.80

Railway Line 0.20

Water Supply 0.027802605

Yes (Boring) 0.571429

No 0.142857

Yes 0.285714

Toilets 0.016913935 37.00 0.80



0 0.20

Deficiency of

Inhouse Connection

0.020375812 0-20 0.065902

20-40 0.108652

40-60 0.166648

60-80 0.250182

80-100 0.408615

Deficiency of

Inhouse Toilet

0.015271129 0-20 0.065899

20-40 0.108654

40-60 0.166653

60-80 0.250169

80-100 0.408625

Road condition 0.011360667 Kaccha 0.20

Pacca 0.80

Street light 0.010144199 No 0.20

Yes 0.80

Drainage 0.013397105 No 0.20

Yes 0.80

5.1 Land suitability score

The land suitability score is based on criteria physical factors which are derived using Delphi technique.

A level -wise analytical hierarchy process is applied to derive weights of various criteria and various

indicators representing them. The scores of individual criteria are computed for 38 slums in the study.

Table 11 Land suitability score

Slum Name Physical

Vishwavidhyalay Pahadi 10 No. 0.30

Chowdhary Mohalla1 0.33

Kanchan Basti 0.35

Wakfha ki Bhumi 0.28

Badhoura gaon 0.27

Pahadi Polipathar 0.37

Vishwavidhyalay pahadi (Press ke Piche) 0.36

Lodhi Mohalla, 0.30

Patel Mohalla 0.33

Peer Baksh Line 0.33

Garha marg Omti Nale Ke Kinare 0.33

Ghoda Aspatal 0.35

Sanjay Nagar Colony 0.30

Railway Line Ke Pass Bhulan Basti 0.29

Dugai Mohalla 0.34

Bhita 0.38

Baraat Road 0.41

Chuee Khadan madiya 0.31

Bilhari (Mandla Road) 0.39

Benisingh ki Talaiya Mominpur 0.32

Shindi Mohalla 0.41

Chamroti Kachiyana 0.38

Chandmari Talaiya Azad Nagar 0.37



New Gokalpur 0.40

Odiya Mohalla 0.40

Sabji Mandi 0.42

Bakra Kabela 0.38

Bhadpuda Basti 0.38

Bada Patthar Kol Mohalla 0.39

Bhan talaiya School ke Piche 0.38

Mansurabad 0.42

Dr. Batalia Ke Samne 0.45

Naya Mohalla 0.48

Chamroti Maida Pass 0.43

Lalit Colony ke Dakshin 0.45

Durga Nagar Colony Basti 0.42

RamHaran Ka Bagicha 0.42

Chowdhary Mohalla 0.45

The classification of land suitability in three categories is based on the research base of site suitability

evaluation for eco-tourism (Bunruamkaewa & Murayamaa, 2011) and classification system developed by S.

Kalogirou (Kalogirou, 2002)

Table 12 Suitability range

Range Physical

Least suitable 0.27-0.34

Marginally suitable 0.35-0.42

Moderately suitable 0.43-0.48

5.2 Land suitability ranking

The ranking is developed in descending order as the slum with highest score has been ranked at highest

position and slum with lowest score is ranked in lower order. For example chowdhary mohalla slum is in

moderately suitability range which is highest of all ranges and so its ranking is highest within the 38 slums.

There are some slums on same ranks and so the highest ranking is 18 in the physical suitability domain

Table 13 Ranking of slums

Slum Code Slum Name Physical

50\2 Vishwavidhyalay Pahadi 10 No. 4

55\6 Chowdhary Mohalla1 7

67\3 Kanchan Basti Ret Naka 9

10\4 Wakfha ki Bhumi 2

11\4 Badhoura gaon 1

68\7 Pahadi Polipathar 11

50\3 Vishwavidhyalay pahadi (Press ke Piche) 10

59\1 Lodhi Mohalla, 4

53\1 Patel Mohalla 7

3\1 Peer Baksh Line 7

1\6 Garha marg Omti 7

6\6 Ghoda Aspatal 9

11\1 Sanjay Nagar 4

1\1 Railway Line Ke 3

12\2 Dugai Mohalla 8



69\6 Bhita 12

3\4 Baraat Road 15

5\1 Chuee Khadan madiya 5

70\1 Bilhari (Mandla Road) 13

40\10 Benisingh ki 6

18\3 Shindi Mohalla 15

9\2 Chamroti Kachiyana 12

33\5 Chandmari Talaiya Azad Nagar 11

45\4 New Gokalpur 14

4\4 Odiya Mohalla 14

15\2 Sabji Mandi 16

19\11 Bakra Kabela 12

45\3 Bhadpuda Basti 12

47\3 Bada Patthar Kol Mohalla 13

18\1 Bhan talaiya School ke Piche 12

39\5 Mansurabad 16

3\2 Dr. Batalia Ke Samne 18

4\3 Naya Mohalla 19

10\5 Chamroti Maida 17

6\5 Lalit Colony ke Dakshin 18

30\2 Durga Nagar Colony Basti 16

5\2 RamHaran Ka Bagicha 16

25\4 Chowdhary Mohalla 18

Figure 7 Physical Ranking of selected slums

In physical ranking the slums that are in lower ranks are in flood prone area and are located on

agricultural land. They have kachha houses with no infrastructure provisions. While slums in higher rank are

in residential areas and irrespective of kachha houses they have proper infrastructure facilities.

5.3 Land suitability assessment for physical suitability

As mentioned in conceptual framework this domain had 32 parameters under study. Amongst which

18 parameters were found to be significant for measuring the suitability of land for slum redevelopment.

This domain comes out to be prominent as it is directly affecting the inhabitant’s basic needs.



6.0 CONCLUSIONS

With respect to the techniques implemented in this study, the integration of AHP in GIS techniques

has been proven beneficial for supporting decision-making. In addition, AHP analysis provides reflection of

real situation of study area. This analysis was effectively used to calculate the details of the factors and class

weights for slum redevelopment. Therefore, the integration of the GIS with AHP combines decision support

methodology which in turn facilitates the creation of land use suitability map for slum redevelopment. This

study can be used as a basis for evaluating the suitability of other areas for slum redevelopment.

The suitability scores suggest suitable measure that needs to be taken for slum redevelopment. This

will not only suggest necessary score to be improved for any slum but will enable the decision makers for

providing tenureship to slum dwellers for short term, midterm and long term depending on the suitability

range. The ideal suitability range can be achieved or at least improved to a particular suitability score. The

thematic model can be further developed in to application based model for suitability analysis incorporating

multiple criteria’s and sub criteria’s.

REFERENCES

1. Al-shalabi, M. A., Mansor, S. B., & Nordin. (2006). GIS Based Multicriteria Approaches to Housing Site

Suitability Assessment. , Germany: XXIII FIG Congress Munich.

2. Angel et al, S. a. (1983). Slum reconstruction; land sharing as an alternative to eviction in Bangkok”, in

Angel,S,et.al (editors), Land For Housing the Poor, Select Books. Thipparat.

3. Anwar, O. E., & Aziz, T. A. (2014). Integrated Urban-Construction Planning Framework for Slum Upgrading

Projects. Journal of Construction Engineering and Management .

4. Arabinda, L. (2003). Integrating GIS and multi-criteria decision making techniques for land resource planning.

Enschede: M.S. Thesis, International Institute for Geo-Information Science and Earth Observation,.

5. Baniya, N. (2008). Land suitability evaluation using GIS for vegetable crops in Kathmandu Valley, Nepal.

Retrieved December 15, 2013, from http://edoc.hu-berlin.de/dissertationen/baniya-nabarath-2008-10-

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Geoinformatics for Spatial Infrastructure Development in Earth and Allied Sciences .

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land suitability classification for tilapia farming in Bangladesh. Aquaculture International, 15: . , 425-443.

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International series in operations research and management sciences. Kluwer Academic Publisher.



“EFFECTS OF CATIONIC LIGANDS ON NITRATE

REMOVAL FROM WATER BY BISMUTH BASED MEDIA”

P B Prasad1 & M K Singh

2

1. Research Scholar ISM Dhanbad, working in BHEL as Sr. Engineer, mobile no 8959311145,

[email protected]

2. Hon. Secretary IEI(Anpara local centre), working in BHEL as Sr. Engineer(PSNR), mobile no

9721454999, [email protected]

ABSTRACT

Nitrate contamination in surface and groundwater is an increasingly problem for all over the world.

Although nitrate is found in most of the natural waters at moderate concentrations, elevated levels in ground

water mainly result from human and animal wastes, and excessive use of chemical fertilizers industrial

waste. More than 95% of the rural population and about 30 to 40% of urban population depend on ground

water source for their domestic requirement. Current available technologies are often inadequate to meet

economic and ecological demands, but in addition the commercial technologies often require large

centralized treatment units. In the above research work it has been established that the presence of cationic

ligands in hydrous metal oxide complexes generally improves the anionic sorptive properties as well as

granular size of the product. The present study was directed to investigate the effects of cationic ligands such

as CaCl2, AlCl3 and FeCl3 on nitrate removal potentials of hydrous bismuth oxide (HBO) powder, after of

experimental works we find conclusion that Calcium and Ferric chloride salts improve the nitrate removal

potentials of both HBO2 and HBO3 powders and also The performance of both Calcium and Ferric salts are

found comparable. Whereas HBO2 with calcium salt remains predominantly yellow in colour that with

Ferric salt becomes brick red in colour. Hence use of calcium salt as cationic ligand appears more preferable

also there is need of future work for analysis of particle size of the above ligands apart from above,

Magnesium being a common divalent cation also needs to be included for its effects on nitrate removal by

HBO powders.

Key words: HBO(Hydrous Bismuth Oxide), Nitrate contamination.

INTRODUCTION

1.1 General:

Nitrate contamination in surface and groundwater is an increasingly important problem for all over the

world. Although nitrate is found in most of the natural waters at moderate concentrations, elevated levels in

ground water mainly result from human and animal wastes, and excessive use of chemical fertilizers. The

other most common sources of nitrate are uncontrolled on land discharges of municipal and industrial

wastewaters, run off septic tanks, processed food, dairy and meat products decomposition of decaying

organic matter buried into ground. These fertilizers and wastes are sources of nitrogen-containing

compounds which are converted to nitrates in the soil. Nitrates are extremely soluble in water and can move

easily through soil into the drinking water supply (Singh and Shrimali, 2001). Excess of nitrate can cause

several environmental problems. The effect of nitrate itself is described as primary toxicity. Its high intake

causes abdominal pains, diarrhea, vomiting, hypertension, increased infant mortality, central nervous system

birth defects, diabetes, spontaneous abortions, respiratory tract infections, and changes to the immune

system. Secondary toxicity of nitrate is microbially reduced to the reactive nitrite ion by intestinal bacteria.

Nitrate has been implicated in methemoglobinemia, especially to infants under six month of age

Methemoglobin (MetHb) is formed when nitrite (for our purposes, formed from the endogenous bacterial

conversion of nitrate from drinking water) oxidizes the ferrous iron in hemoglobin (Hb) to the ferric form.

MetHb cannot bind oxygen, and the condition of methemoglobinemia is characterized by cyanosis, stupor,




and cerebral anoxia. Symptoms include an unusual bluish gray or brownish gray skin color, irritability, and

excessive crying in children with moderate MetHb levels and drowsiness and lethargy at higher levels.

Hemoglobin (Fe2+) NO2_ Methemoglobin

(Can combine with oxygen) (Cannot combine with oxygen)

In tertiary toxicity, the reaction between nitrite and secondary or tertiary amine in acidic medium can

result in the formation of N-nitroso compounds, some of which are known to be carcinogenic, tetratogenic,

and mutagenic.

H3C--- NH + HNO2 H3C-N – N = O + H2O

CH3 CH3

Dimethyl Amine Dimethyl nitrosamine (Carcinogenic)

A diet, adequate in vitamin C, partially, protects against the adverse effects of nitrate–nitrite.

Methaemoglobinaemia in infants can only be mitigated by blood transfusion (Schoeman and Steyn, 2004).

To protect consumers from the adverse effects associated with the high nitrate intake, nitrate

consumption should be limited and standards have been established (Sahli et al., 2002). According to World

Health Organization (WHO), drinking water must contains no more than 50 mg/L of nitrate and EPA

established a maximum contaminant level of 45 mg/L. European Community recommends levels of 25 mg-

NO-3 /L. Several methods for nitrate removal from drinking water resources have been applied. The methods

available for the removal of nitrate are Ion exchange, Biological Denitrification, Catalytic Reduction ,

Reverse Osmosis and Electrodialysis.

1.2 Sources and Causes of Nitrate Contamination:

Because nitrate is unreactive and water-soluble, it will remain in a well or aquifer unless it is flushed

out by water containing lower nitrate levels. Agriculture is considered as the main source of nitrate

contamination in groundwater. Excessive use of chemicals and fertilizers increases the risk of groundwater

contamination. Nitrates and nitrites also form during chemical production and they are used as food

conservers. This causes groundwater and surface water nitrogen concentration, and nitrogen in food to

increase greatly. It is a common nitrogenous compound due to natural processes of the nitrogen cycle.

Anthropogenic sources have greatly increased the nitrate concentration, particularly in groundwater. The

largest anthropogenic sources are septic tanks, application of nitrogen-rich fertilizers to turf grass, and

agricultural processes. These fertilizers and wastes are sources of nitrogen containing compounds which are

converted into nitrates in soil. Nitrates are highly soluble in water and can easily percolate through soil into

the drinking water supply. The other most common source of nitrate are uncontrolled on land discharges of

municipal wastes and industrial waste waters, run off septic tanks, processed foods, dairy, and meet products,

decomposition of decaying organic matter buried into ground. Some of the causes of nitrate contamination

are like Drought where the same quantity of nitrate is present, but in less water can dramatically increase the

nitrate concentration of a water source. Water that had been Safe might, under drought conditions, exceed

the EPA safe drinking water limit conversely, sudden increases in groundwater following flooding or

excessive rain can also cause nitrate levels to rise in wells by flushing nitrate into a new area from a

contaminated site. Large areas of the US are under drought conditions this summer. If your business is

located in an agricultural region and drought conditions are present, new nitrate problems will start to show

up.

1.3 Standards of nitrate in ground water: The current maximum allowable concentration of nitrate in drinking water ranges from 2.5 mg NO3

- /L

in Norway to 23.0 mg NO3- /L in Netherlands (Rittman and Huck, 1989). World Health Organization

(WHO) has recommended 50 mg NO3- /L. The United States, Canada and India suggest a maximum

permissible limit of 10 mg N/L (equivalent to 45 mg NO3-/L). Because there are no conclusive

epidemiological studies which link nitrate to cancer in humans, carcinogenicity was not taken into account in

the establishment of the Maximum Contaminant Level (MCL) for nitrate by USEPA.

1.4 Effects of ground water nitrate contamination:



The adverse effect can be in the form of deterioration of water quality and potential hazard to human and

animal health. Studies have correlated ingestion of high nitrate and stomach cancer. Other adverse effects

can be noted as, abdominal pains, diarrhea, vomiting, hypertension, central nervous system, birth defects,

diabetes, birth defects, spontaneous abortions studied by (Hill et al., 2004).

1.5. Effects of high nitrate concentration on health:

I. Blue Baby Syndrome:

Cases of blue-baby syndrome usually occur in rural areas which rely on wells as their primary source

of drinking water. Often these wells become contaminated when they are dug or bored and are located close

to cultivated fields, feedlots, manure lagoons or septic tanks (Comly, 1987). The most contaminated wells

are usually those that were dug rather than drilled and have poor or damaged casings. Until recent awareness

of the dangers of nitrate contaminated groundwater prompted testing for nitrate concentrations, along with

other contaminants, wells with dangerously high nitrate concentrations usually went unnoticed until health

problems were brought to attention. A few isolated cases of methemoglobinemia, primarily in the rural

United States, have served as the catalyst for what has grown into a broad awareness and concern for nitrate

contamination.

II. Stomach and Gastrointestinal Cancer:

Scientists claim that nitrate represents a potential risk because of nitrosation reactions which, with

appropriate substrates present, form N-nitroso compounds which are strongly carcinogenic in animals. In

other areas of the world such as Columbia, Chile, Japan, Denmark, Hungary, and Italy, similar studies have

suggested a correlation, although there still exists no concrete evidence to support this theory. At present, no

other toxic effects have been observed under conditions of high nitrate levels. Even at exposure to levels of

111mg/L there were no adverse conditions in infants except for methemoglobinemia (Comly, 1987).. Other

claims that intake of nitrate contaminated groundwater is linked to birth defects, and hypertension and high

blood pressure in adults are also unsubstantiated.

1.6 Treatment process for nitrate removal:

In many areas of the world the simultaneous (i) lowering of water tables, (ii) increasing use of

fertilizers and pesticides and (iii) contamination by chemical and non-chemical products has significantly

reduced the fraction of fresh water which can be used for human purposes. As a consequence, it is estimated

that the market for water remediation technologies, e.g. technologies to treat contaminated water which can

bring it to drinking water quality for human use, will double in the next 5–10 years. There are thus social,

ecological and economic driving forces which stimulate the development of new water remediation

technologies.

Current available technologies are often inadequate to meet economic and ecological demands, but in

addition the commercial technologies often require large centralized treatment units. In many cases, wells

serve small local communities which cannot be connected to centralized water treatment units. It is thus

necessary to develop technologies which are compact, transportable and easily manageable. Available

technical data, experience and economies indicate that ion exchange process is most suitable method for

ground water supply for its simplicity effectiveness, recovery and relatively low cost.



6.1 SUMMARY OF RESULTS AND CONCLUSIONS

From a critical overview of available literature, it has been established that the presence of cationic

ligands in hydrous metal oxide complexes generally improves the anionic sorptive properties as well as

granular size of the product. The present study was directed to investigate the effects of cationic ligands such

as CaCL2, AlCl3 and FeCl3 on nitrate removal potentials of hydrous bismuth oxide (HBO) powders. From

the results of experimental works done and analysis of the data presented in the thesis, following

observations are made:

1. Calcium and Ferric chloride salts improve the nitrate removal potentials of both HBO2 and HBO3

powders.

2. Aluminum salt gives irregular changes in nitrate removal properties of HBOs. It also decreases the

pH of resultant water possibly due to acidic characters of its polymeric salts. Hence use of Al salt for

such purposes is not suggesting.

3. The performance of both Calcium and Ferric salts are found comparable. Whereas HBO2 with

calcium salt remains predominantly yellow in colour, that with Ferric salt becomes brick red in

colour. Hence use of calcium salt as cationic ligand appears more preferable.

4. The pH of treated water remain in the admissible range for drinking purpose and the chloride

concentration in water appears proportional to the nitrate removal.

5. For both HBO2 and HBO3 and either with CaCl2 or FeCl3, the maximum nitrate removal by the

powders have been observed at a cationic proportions of 0.050 M. Hence this proportional mix

appears the most preferable for such purposes.

6.2 Scope of Future works

From the above thesis work we reach to the following scope of future works:

1. From the results it is observed that the HBO powders prepared in this set of experiment were

possibly overdried. HBO2 powders with controlled drying conditions should be prepared and

investigated for their nitrate capacities.

2. Magnesium being a common divalent cation also needs to be included for its effects on nitrate

removal by HBO powders.

3. Particle size of different powders needs to be evaluated for their practical applications as filtering

media in drinking water treatment.

REFERENCES

[1]. Ayyasami, P. M., and Shanthib, K.(2007). “Two satge Removal of Nitrate from ground water using biological

and chemical dinitrification.” J of Bioscience and Bio engineering. 104- 2.p. 129-134.

[2]. Ball, M., and Harries (1988). “Ion Exchange Resin Assesement”. In Michael. S.(Ed.), Ion Exchange for industry.

Ellis Horwood Publishers, Chishester, West Sussex England.

[3]. Centi, G., and Siglinda, P. (2002). “Department of Industry Chemistry and Engineering of Materials, University of

Messina, and INSTM”. Consortium for Science and Ttechnology of Materials, Salita Sperone 31, 98166 Messina

Italy.

[4]. Shrimali, M., and Singh, K. P., 2001. New methods of nitrate removal from water. J. Environmental pollution 112

(2001) 351-359.

[05]. Singh, P. K., and Ghosh, D. K., (2000). “Nitrate removal from Water by Bismuth Based Media.” Water Recycling

and resource management in the Developing World.p. 456-459.

[06]. Singh, P.K., (1999). Nitrate Removal from water by bismuth based media.’ Ph.D. Thesis, department of Civil

Engineering. I.I.T. Kanpur.



[07]. Spalding, R.F., and Exner, M. E., (1993). Occurance of nitrate in ground water a review Journal of

Environmental Quality. 22. 392-402.

[08]. USAEP (1993). Consumer factsheet on Nitrate/Nitrites, http:/www.epa.gov/ogdwd/dwh/c-ioc/nitrates.html.

[09]. Vaaramaa, K., and Lehto, J., (2003). “Removal of metals and anions from water by ion exchange process”.

Journal of Desalination. 155. 157-170.

[10]. WHO, 1993. Guidelines for Drinking Water Quality. 1. Recommendations, 2nd Edition. World Health

Organisation, Geneva.



Promising Prefab Technology for Mass Housing in Indian

Scenario

Ajay Chourasia1, Yogesh P Kajale

2, Shubham Singhal

3 and Jalaj Parashar

4

1Principal Scientist, CSIR-Central Building Research Institute, Roorkee, E-mail: [email protected]

2Vice-President, BG Shirke Construction Technology Pvt. Ltd., Pune, E-mail: [email protected]

3Research Assistant, CSIR-Central Building Research Institute, Roorkee, E-mail: [email protected]

4Senior Technical Officer, CSIR-Central Building Research Institute, Roorkee, E-mail: [email protected]

ABSTRACT

With rapid growth in urbanization, sustainable development and utilization of resources is of paramount

importance, which can be addressed by development of proven prefab system for mass housing

construction. The developed system shall have edge in regards to safety, speed, serviceability. The paper

attempts to highlight the features of one of the prefab system viz. 3-S prefab building system. The

sustainability aspect of building construction vis-à-vis the prefab system of industrialize housing

construction is elaborated. The system describes use of prefab elements to a maximum possible extent with

connections facilitated through certain level of cast-in-situ concrete at project sites. The building system has

been evolved and perfected to cater to the seismic requirements as well as typical conditions prevailing in

India.

Keywords: sustainable, mass housing, prefabrication, 3-S system, precast, reverse cyclic load, performance

evaluation, autoclaved lightweight cellular concrete.

1. INTRODUCTION

Largest consumer of natural resources such as water, sand, crushed rock, gravel, minerals, timber etc. is

the construction industry. The demand for housing units, energy, clean water & air, safe & rapid transport

etc. is increasing tremendously with the growing population, urbanization and industrialization. On the other

hand, available natural resources are limited in quantum and also becoming scares day by day. Construction

Industry is primarily dependant on certain manufacturing industries such as cement, steel and aluminum;

which are amongst the most energy intensive apart from major consumer of scares natural resources.

Adoption of environmentally friendly and sustainable technology in the construction is of paramount

importance.

The rocket pace of urbanization, population leads to growing infrastructure needs. Such growing needs

will required to be balanced against equally important human and species need of preserving life-sustaining

environment on entire earth; which is being threatened by uncontrolled use of natural resources leading to its

depletion and increasing pollution. Necessity to adopt Sustainable Building Technology is therefore of vital

importance to overcome the threat to our standard of living and more importantly to the entire fabric of life

support system on which the planet earth is dependent.

As the population grows, enormous demand is put on natural resources and on the supply of construction

materials to build new infrastructure needed to support human’s basic necessities i.e. food – shelter –

clothing. Recently, China has reported the highest rate of construction activity of any country in the world

and it is forecasted that the Pearl River Delta region (comprising Hong Kong, Shenzhen, Zhuhan and Macau)

alone could become the world’s largest megacity. In India, Mumbai, Delhi, Chennai and Kolkata are also on

the verge of such disproportionate surge of development needs in near future. Such development will require



an enormous amount of infrastructure and therefore will create tremendous pressure on resources and the

environment. Engineers, as highly respected designers of the infrastructure, are in a unique position to

influence such developments to be SUSTAINABLE, thereby resulting into ‘GREEN FUTURE’.

2. SUSTAINABLE CONSTRUCTION

Sustainable construction can be defined as the one which meets the needs of the present without

compromising the ability of future generations to meet their own needs. The sustainable technology of

construction therefore requires maintaining the harmony of the earth’s eco-system.

The building i.e. ‘The Shelter’ protects the mankind from nature’s extremes such as cold, heat, rain and

snow. These building structures affect our environment too, since it consumes enormous amounts of energy,

water, material and creates large amounts of waste. The concept of sustainable building construction is to

incorporate construction technology that result in environment protection, water conservation, energy

efficiency, usages of recycled products and renewable energy. Such technology ensures that waste is

minimized at every stage during construction and operation of the building, resulting in low costs. Apart

from being environmentally responsive and profitable, the buildings constructed by adopting ‘Sustainable

Technology’ look for creation of a healthy place to live and work in.

The prerequisites for sustainable construction can therefore be as under.

Judicial use of construction materials there by requiring lesser materials i.e. products that conserve

natural resources

Use of products that avoid toxic or other emissions

Reduction in wastage of materials during construction of buildings & utilizing wastes to make

construction materials

Reducing emissions during the production of construction materials

Using more durable materials in buildings thereby requiring lesser maintenance cost

Use of energy efficient building materials and products that save energy or water i.e. the materials

requiring low energy for their production as well as will consume lesser energy during life cycle of

building due to its’ use

Use of products that contribute to a safe, healthy built environment

Use of materials which can be recycled

Use of construction system minimizing air, water and noise pollution during construction

Prefab building techniques can certainly be able to fulfill many of the above and hence could be one of

the most preferred choices for building construction. One of the available prefab building construction

systems in India is therefore studied here under w.r.t. above parameters.

3. PREFAB TECHNOLOGY – A SUSTAINABLE CONSTRUCTION TECHNIQUE

The ‘3-S’ prefab system has been used substantially in various mass housing schemes of housing

boards and it is reported that about 200 thousand housing units have been constructed in buildings of Ground

plus three storey to Ground plus twenty four storey by using this prefab technology for different clients over

past four decades.

The prefab system being subjected to time testing, therefore is evaluated here in this paper in the

context of Sustainability since it is found to be one of the system fulfilling majorities of the pre-requisites of

energy efficient building.



Figure 1: Isometric view of 3-S prefab building system

3-S prefab solution (Fig.1) for the housing sector includes following elements.

Precast RCC dense cement concrete slabs or Autoclaved light-weight energy efficient cellular

reinforced cement concrete slabs for floor and roof.

Autoclaved light-weight energy efficient cellular cement concrete building blocks.

Precast reinforced dense cement concrete structural components e.g. columns, beams, toilet slabs,

stairs, etc.

Galvanised powder coated press metal frames and shutters for doors / windows.

The above elements and the prefab system have been developed and perfected over past four decades

taking into consideration the local conditions and by resorting to mechanisation in the construction process

to deliver the quality goods within the shortest time frame.

4. HIGHLIGHTS OF THE 3-S PREFAB CONSTRUCTION SYSTEM

An engineered 3-S system has been developed and perfected to achieve Speed, Strength, Safety and life

cycle economy. The said prefab system comprises of all structural building components which are

manufactured in factories / on-site casting yard under objective quality control. 3-S system differs from

many other available prefab building systems particularly in respect of the connections since ‘3-S’ system

utilizes cast in-situ wet connections extensively for jointing of various prefab structural elements. The 3-S

system involves following activities for construction of building:

Cast in-situ sub-structure including foundations, stem columns, plinth beams, and plinth masonry,

Procurement of prefab elements such as columns, beams, slabs, lintels, chajjas, stairs, and masonry

blocks etc.



PRECAST R.C.C.STAIRCASE

DOOR FRAME SECTION

STEEL DOOR AND WINDOW FRAMES & SHUTTERSSHIRKE POLYNORM NON-WOOD

BUILDING.BLOCKS

SIPOREX

ELEMENTS OF 3-S PREFAB SYSTEM

PRECAST R.C

.C.L

INTEL

PRECAST R.C

.C.L

INTEL

PRECAST 3-S B

EAM

PRECAST R.C

.C.C

HAJJA

PR

EC

AS

T R

CC

3-S

CO

LU

MN

SIPOREX FLOOR / R

OOF SLABS

Erection of precast components, jointing of these components using cast in-situ self compacting

concrete with appropriate reinforcement,

Laying of reinforced cast in-situ screed over slab panels, construction of walling flooring, plastering,

water-proofing, etc.

The ‘3-S’ prefab system utilizes Precast Dense Concrete Hollow Column shell of modular design size in

combination with precast dense concrete Rectangular/'T' shape/'L' shape beams and light-weight reinforced

autoclaved cellular concrete slabs for floors and roofs (Figure 2 and Figure 4). The hollow cores of columns

are concreted with appropriate grade of in-situ self compacting concrete once they are erected in position.

All the connections and jointing of various prefab elements are accomplished through in-situ concreting

along with secured embedded reinforcement of appropriate size, length and configuration.

Figure 2: Elements of 3-S prefab building system

The reinforcement provision is determined analytically considering rigid joint behavior and the structural

detailing is made as per manufacturer’s practice; as described here under.

Foundation to Column and Column to Column joint: Dowel bars of appropriate size are provided

with required development length in the foundation; over which the stem is cast and hollow column shell is

then erected, followed by in-situ concreting of hollow core with appropriate mix of self compacting concrete.

Column to column joints are also made in the similar manner with provision of dowel bars / continuity bars

as in the case of foundation to column.

Column to Beam joint: Appropriate notches are provided in the shell of hollow columns wherein the

precast concrete beams are placed. Mechanical anchorages are provided for the bottom reinforcement of



beams as per the requirement of structural design. Design reinforcements are placed at the top surface of

beam passing through the protruding rings of precast beam and anchored into core of hollow columns. The

top portion of beam is then concreted along with column core concreting.

Beam to Beam joint: This is accomplished through precast miter joint of appropriate size and

reinforcement getting embedded into in situ concrete.

Beam to Slab joint: Light-weight reinforced autoclaved cellular concrete (Siporex) slabs are placed in

between the beams over their flanges with adequate bearing. The Siporex slabs units are fitted sidewise with

tongue and groove joints. Reinforcement mesh is placed over the slabs; which is suitably anchored into

peripheral beams. Deck concrete of 40mm thickness is then poured on top of slabs.

Self-compacting concrete is used for column cores and beam tops concreting operations. Details of 3-S

system as adopted by manufacturer are illustrated in Fig.3 and Fig.4.

Figure 3: L-section of 3-S column-beam

The continuity and rigidity of various joints in ‘3-S’ prefab system is achieved as per the manufacturers’

norms by joining the individual members through in-situ concrete with additional reinforcement as required

in the structural design.



Figure 4: Isometric view of 3-S column-beam assembly

This prefab building system has the following distinct advantages, which rank it as one of the technology

for ‘Sustainable Precast Construction’.

Time-tested technology

It is reported that various evaluations, reviews, static and cyclic tests have been carried out by different

experts at various universities and government bodies like IIT Bombay[1]

, Stanford University[2]

, Tor Steel

Research Foundation in India[3]

, and CIDCO[4]

etc. to ascertain the design parameters and to validate the ‘3-

S’ prefab system. It is observed from the findings of these reports that obtained results are in good agreement

with the required performance criteria / provisions.

Recently, Central Building Research Institute, Roorkee (CBRI) had carried out reverse cyclic lateral load

test on full scale ‘3-S’ prefab frame structure model (Ref Fig.5) at their building dynamics laboratory to

ascertain the behavior of the joints and connections under seismic loading.

FOUNDATION

STEM

1

2

TO PROVIDE NOTCHES FOR FIXING `3-S'BEAM

NOTCHES IN COLUMN'S OUTER SHELL TO

PROVIDE A SEAT TO THE `3-S' BEAM

PRECAST SHELL OF `3-S' COLUMN

STARTER BARS FROM STEM FOR FIXING

AT BOTTOM END OF `3-S' COLUMN

'3-S' COLUMN (GROUTED AFTER ERECTION)

PLINTH BEAM

4

3

6

5

SUPPORT SLAB UNITS

T-FLANGE OF THE `3-S' BEAM TO

RIB OF `3-S' BEAM WITH NOTCH AT END

FOR JOINT WITH COLUMN

BEAM MAIN REINFORCEMENT (PROTRUDING

VERTICAL CONTINUITY STEEL FOR JUCTION

OF UPPER & LOWER COLUMN

FORKED FLANGES AT TOP OF THE COLUMN

FROM RIB) ANCHORED UP/DN INTO COLUMN

HOLLOW CORE OF THE PRECAST

9

8

7

TOP ANCHOR BARS13

12

11

10

SHEAR LOOPS FROM RIB PROTRUDING

INTO IN-SITU PORTION

SLAB UNITS15

14



Figure 5: Photograph and finite element model of 3-S full scale structure

From the observations of the recent full scale reverse cyclic load test[5]

it is seen that structure behaved

in the elastic mode and withstood more than 6 times cyclic lateral load corresponding to Maximum Credible

Earthquake value for Seismic Zone IV. Even at this loading; no collapse mechanism was observed and the

structure responded in elastic range which indicates that the ‘3-S’ prefab structure having light-weight

autoclaved cellular reinforced concrete slabs has a large ductility and can withstand even more lateral load.

Although cracks are observed in the structure; the same disappeared at no load conditions in the cyclic

behavior of loading. The diaphragm comprising of panels of autoclaved reinforced cellular light-weight

concrete having 40mm thick nominally reinforced in-situ deck concrete was effective in reverse cyclic lateral

load transfer mechanism and no opening of panel joints were observed. All joints and connections observed

intact even under high magnitude cyclic lateral applied loading. The manufacturers’ detailing for the ‘3-S’

prefab structural elements and connections therefore conformed to the performance requirements. The

experimental results on ‘Full-Scale Building Structure’ established the desired performance and behavior of

‘3-S’ prefab building system under all design load conditions including Seismic Zone IV for High Rise

Buildings.

5. NEED FOR SWITCHOVER TO SUSTAINABLE ALTERNATIVES

Steel, cement, glass, aluminium, plastics, bricks, etc. are energy-intensive materials, commonly used for

building construction. Generally these materials are transported over great distances. Extensive use of these

materials can drain the energy resources and adversely affect the environment. It is therefore essential to

adopt energy efficient innovative materials and prefab technology to meet the ever-growing demand for

buildings. There is an immediate need for optimum utilization of available energy resources and raw

materials to produce simple, energy efficient, environment friendly and sustainable building alternatives and

techniques to satisfy the increasing demand for buildings. Some of the guiding principles in adopting the

sustainable alternative building technologies can be summarized as follows:

Energy Efficient: thermal mass benefits

High thermal mass of precast concrete enables it to absorb, store and later radiate heat.

Using precast concrete in passive solar designs allows natural heating in winter and cooling in

summer, thereby reducing the need to rely on artificial heating and cooling.



Improved internal building amenity

Use of precast concrete can even out internal diurnal building temperatures.

3-S Prefab technology having light-weight cellular concrete elements for slabs and walling can

improve indoor air quality, providing comfortable temperature inside the home.

Durable, long life, reuseable, low maintenance structures

With a long life expectancy owing to dense concrete structural elements, 3-S precast structures are

durable.

3-S Precast structures can be extended and refitted internally. Structures do not need to be

demolished and can simply be renovated internally conserving resources, reducing waste and

landfill.

3-S Precast is easy to keep clean, requiring minimal maintenance.

3-S Precast is tough and can withstand wear and tear, requiring minimal repairs.

3-S Precast concrete gains strength as it ages, won’t shrink, distort or move and will not deteriorate

with exposure to climatic change.

3-S Precast concrete construction is seismic resistant, offering protection against earthquake hazards.

Locally supplied

Materials used in 3-S precast construction are all supplied locally. This reduces haulage and fuel

costs and also diverts resources from landfill.

3-S Precast elements are all locally manufactured and supplied to sites meaning reducing haulage

and fuel costs.

Local highly skilled erection crews erect 3-S precast concrete elements safely on site.

Uses less concrete, cement and steel

Less concrete and steel are required for 3-S precast concrete because of its higher quality and

lightweight.

Less concrete is used in 3-S precast flooring systems due to use of lightweight cellular concrete

slabs.

Precast allows reduced levels of cement in the concrete mix due to higher quality manufacturing

processes and also due to substantial reduction in dead loads.

Foundation work i.e. excavation and concreting is reduced to a substantial extent due to reduction in

dead loads of the structure.

Minimises waste during manufacture and on site

Manufactured in reusable moulds.

Most waste during manufacture is recyclable.

Exact elements are delivered to site.

Less site air pollution, noise and debris.

Reuses waste resources and recycled materials

Waste materials (such as slag and fly ash) which would otherwise be used in landfill are

incorporated into the precast mix design.

Recycled aggregate / manufactured sand is incorporated in to the precast mix design thereby

avoiding use of natural sand.

Recyclable precast elements

Precast concrete elements can be crushed and reused as aggregate for road bases or construction fill,

providing economic and environmental savings.

Faster construction



3-S Precast construction allows other trades to begin work more quickly, speeding the construction

time and saving costs.

Fast construction on site means fewer disturbances for surrounding properties.

Precast elements can be delivered just in time for erection, reducing unnecessary handling and

equipment use.

Acoustic performance

The high thermal mass of precast concrete assists with sound insulation to reduce noise and absorb

noise impact.

Fire resistant

3-S Precast concrete is non-combustible, does not melt and therefore does not require additional fire-

proofing applications.

3-S Precast concrete does not emit toxic fumes under fire and can limit smoke spreading in

buildings.

The 3-S Precast building structures possesses high resistant to fire due to use of cellular concrete

materials in walls and floors.

Environmental benefits

3-S Precast concrete element is an inert substance which does not emit or give off gases or

compounds.

Precast does not attract mould or mildew.

Precast concrete absorbs CO2.

Being termite proof means the unlikelihood of requiring chemical spray to reduce termites and

vermin which is safer for the environment.

OHS benefits

Less trades on site means safer sites with less equipment, workers and materials.

Reduced congestion - construction sites are cleaner and tidier, with minimal waste on site.

3-S Precast floors can provide a safe immediate working platform for the erection crew as well as for

other trades.

Environmental sustainability of SIPOREX and 3-S Prefab:

Reduces Air Pollution at Construction sites because of site activity is minimal to erection and

jointing

Use of fully "Cured" and "Matured" components considerably reduces water consumption

High thermal insulation results in achieving energy efficiency

Raw materials as well as energy requirement in manufacturing is considerably less

Production utilizes fly-ash to a great extent

Manufacturing process is non-toxic and environmental friendly which does give off any harmful

emissions during production

No waste generated in the production process due to reuse of waste material

Water used for curing and making steam is re-circulated to minimize the water wastage

Reduces wastages considerably owing to better quality / process controls and repetitive task

Low workability mixes can be designed with lesser w/c ratio as well as lesser fine aggregate contents

Exact concrete consumption can be controlled

Very minimal requirement of water for construction

Non-generation of construction debris

Conservation of wooden material - Elimination of use of timber / wooden scaffolding by use of

pressed steel door windows & steel shuttering & scaffolding



Conservation of Natural resources- Optimum use of scarce natural resources like sand and

aggregates

Conservation of cement and reinforcement: Substantial saving in quantity of cement &

reinforcement steel, coarse & fine aggregates due to reduction in dead weight of structure

Environment friendly, Energy efficient GREEN BUILDING Construction

Safe & durable structure

6. CONCLUSION

The ‘Prefab technology for sustainable building’ movement is advancing at a rapid pace in other parts of

the world and is yet to take accelerating start in India. Considering the tremendous technological and

ecological benefits that such prefab buildings can result in, several corporate and government agencies are

required to adopt precast construction system.

By adopting prefab building technology, using light weight building materials such as AAC to the extent

possible, cement replacement materials such as fly ash in concrete, designing for durability as well as

undertaking life cycle analysis of construction projects, it is possible to direct the construction industry, and

particularly the concrete building industry down a more sustainable path. As highly regarded professionals,

engineers are in a position to be in the driver’s seat of this process. And by doing so, engineers have an

opportunity to influence the course of human history beyond the realm of technology.

ACKNOWLEDGMENT

The authors gratefully acknowledge Director, CSIR-CBRI, for permitting to publish the paper.

REFERENCES

[1] Buragohain D.N., Limaye R.G., Ranganathan R. (1992), “Evaluation of Design and Testing of 3-S

Joints”.

[2] Shah C.H (1991), “Structural Design Calculations for HIG-III (S+7) Buildings for Powai-Mass Housing

for MHADA (BH&ADB), BOMBAY” Department of Civil Engineering, Stanford University,

Stanford.

[3] Vishwanatha C.S. (1995), “Full scale load test on assembly of R.C. Precast units of M/s B.G.Shirke

Construction Technology Pvt. Ltd., being used at National Games Housing Project, Koramangala,

Bangalore” Tor Steel Research Foundation in India, Bangalore.

[4] Ramesh C.K. (1986), “Performance Evaluation Test on G+3 Building subjected to the simulated loading

representing the actual real life loading” Indian Institute of Technology, Bombay.

[5] Chourasia Ajay, Singh S.K., Parashar Jalaj (2010), “Testing and Evaluation of 3-S Prefabricated System

to Establish Behavior of Various Joints under all Design Loads including seismic (Zone IV) on Full

Scale Building” Central Building Research Institute, Roorkee.



Seismic Performance of Prefab RC Shear Walls: A Review

Ajay Chourasia1, Shubham Singhal

2 and Jalaj Parashar

3

1Principal Scientist, CSIR-Central Building Research Institute, Roorkee, E-mail: [email protected]


3Senior Technical Officer, CSIR-Central Building Research Institute, Roorkee, E-mail:

[email protected]

ABSTRACT

Rapid urbanization and population growth around the globe has resulted in enormous pressure on

construction industry, which has evolved a need of speedy and efficient construction technology with safety

from different hazards. Precast technology offers speedy and quality construction with the reduced

construction cost on mass scale production. However, in spite of significant R&D, technological

advancement, precast shear walls did not demonstrate satisfactory performance when subjected to lateral

forces, particularly due to poor behavior and response of connections. Several researchers have evaluated

different types of connections such as loop bar connection, shear keys, O-connectors, seam connection, joint

connecting beam etc. between precast wall panels to study their seismic behavior. This paper presents a

state-of-art literature review on existing technologies and development in the field of precast RC shear wall

systems.

Keywords: Prefab shear wall, joint connections, seismic behavior, cyclic lateral load.

1. INTRODUCTION

Precast construction technology possess a high quality control due to industrialized construction where

the structural components are manufactured in a factory and then transported to the construction site for

assemblage, thus vanishing the possibilities of commonly occurring errors at the construction site. Precast

construction incorporates several advantages over cast-in-situ, such as high quality, reduced construction

cost, speedy and more sustainable construction, which has led precast construction technology to gain

popularity among engineers. Utilization of factory tools and machines along with the elimination of adverse

weather impacts on the construction in precast technology leads to quality control in the most prominent

way. Precast structures have proved to be highly satisfactory in sustaining gravity loads; however their

effectiveness for resisting lateral loads is still not clear. Thus, it is of utmost importance to investigate

seismic resisting features in precast shear wall. Provision of shear walls in the building is the most common

and reliable seismic resisting feature which behave as vertical cantilever beams and adequately transfer

lateral loads from the superstructure to the foundation.

Previous earthquakes have clearly demonstrated poor performance of joint connections, especially in the

regions of high seismic zone. 2012 Emilia earthquake has caused damage to the industrial precast concrete

structures, resulting in high economic losses. Damage was attributed to inadequate and weak connections

between precast panels (Magliulo et al. 2013). Similarly, L’Aquila earthquake resulted in collapse of walls of

precast buildings, which was a direct consequence of inadequate anchorage between the precast wall panels

(Grimaz and Maiolo, 2009). Many other past earthquakes such as Vrancea earthquake (Tzenov et al. 1978),

Friuli earthquake in Italy (EERI 1979), Montenegro earthquake (Fajfar et al. 1981), Spitak earthquake in

Armenia (EERI 1989), Northridge earthquake in Los Angles (Bonacina et al. 1994), Kocaeli earthquake

(EERI 2000; Saatcioglu et al. 2001) have exhibited damage in precast concrete structures. The main reason

of damage primarily lies in the failure of connections and insufficient ductility of precast systems, leading to

early crushing of concrete and low deformation capacity of the system. However, there are instances of

satisfactory behavior of precast walls as well, such as in the case of 7.2 magnitude Kobe earthquake, in

which most of the precast structures performed well (Muguruma et al. 1995). Thus, precast structures are

capable of sustaining high seismic loads, provided their connections are properly designed and constructed.



This has necessitated the requirement of appropriate analysis and experimental tests on precast RC shear

wall to evidently identify their lateral load resisting behavior.

2. CONNECTION SYSTEMS IN PREFAB JOINTS

Different types of connections such as grouted joints, mechanical connectors, steel connections, shear

keys etc. may be provided to transfer the forces between the precast RC shear walls (ACI 318-1995).

Hardware materials used for connections include reinforcing bars, plain wire, coil inserts, deformed bar

anchors, bolted and threaded connectors, welded headed studs, post-tensioning steel etc. These should be

adequately anchored in the concrete to achieve specified strength and ductility (MN L-l 23-88, PCI, 1988).

Some of the connections that are popular in prefab industry are discussed below.

2.1. Loop Bar Connection

Loop bars are one of the most commonly used connection types, in which loop bars protruding from the

wall panels are connected by lap splicing in the intermediate joint as shown in the Fig. 1. A transverse bar is

inserted between the loop bars which forms a locking mechanism and ensures integrity of the connection.

Loop bars from one wall panel transfer force through middle vertical bar which pulls out the loop bars in the

second panel. The gap produced is filled with concrete or grout to impart rigidity to the joint. Effectiveness

of this type of connection depends on the spacing between loop bars, amount of transverse reinforcement and

the embedded length of loop bar into the wall panel concrete (Henrik and Linh, 2013).

Figure 1: Loop bar connection

2.2. Vertical Interlocking Joint

Interlocking joints as shown in the Fig. 2 are designed to effectively transfer the load from one panel to

another. In this system, connection between two mutually perpendicular walls is provided by inserting a

vertical bar at the interface of two walls. However, overall seismic resistance capability of this system has

not been verified by the experimental tests (MN L-l 23-88, PCI, 1988).



Figure 2: Vertical joint in precast walls

2.3. O connectors

A mild steel oval shaped O-connector was used to connect the vertical joint between wall-column or wall-wall

(Sritharan et al. 2015) as shown in Fig. 3. Experimental investigation on this system has revealed that it has the

capability to dissipate energy by undergoing flexural yielding in the loading plane.

Figure 3: O-connector

2.4. Reinforcing Bars

This connection comprise of a straight protruded continuity reinforcement bar from the top panel is

welded to a 75 x 75 x 10 mm steel angle placed in an exposed pocket of the lower panel as shown in the Fig.

4(a). The gap in the joint is filled with the dry pack and compacted to achieve the desired strength (Khaled et

al. 1995).

Alternatively, continuity reinforcement bar from the top panel is placed inside the splice sleeve

embedded in the bottom panel as shown in Fig. 4(b). The splice sleeve is pressure grouted with high strength

and non-shrink mortar. (Khaled et al. 1995).



(a) (b)

Figure 4: (a) Reinforcing bar welded into steel angle; and (b) Reinforcing bar with splice sleeve

2.5. Shear Keys

To enhance the seismic resistance of the connection, shear keys along the joint of wall panels are

practiced as shown in Fig. 5. The surface of wall panels may be kept triangular, semi-circle, rectangular or

trapezoidal instead of plain surface. It has been observed that the shape of shear key results in significant

enhancement in shear resistance of wall panels. The experimental study showed that the precast wall with

dry pack shear key increased the shear capacity by 60% as compared to same configuration plain surface

wall (Rizkalla et al. 1989). However, provision of continuity bars or mechanical connectors is recommended

along with dry pack in shear keys, as dry pack alone may not demonstrate satisfactory performance under

lateral loads.

Figure 5: Shear key along the joint

2.6. Joint Connecting Beam

Xilin et al. (2016) proposed a connection in which upper and lower wall panels are connected by a

horizontal joint connecting beam. Steel bars are stretched out of the wall panels and bent to form rectangular

closed steel rings. Bond between the walls is strengthened by chiseling the interface. When the upper and

lower wall panels are overlapped, stirrups are inserted into the spacing between steel rings as shown in Fig.

6. Longitudinal bars are then inserted and fixed through the stirrups and rings followed by concrete casting

of adequate strength.



Figure 6: Horizontal Joint connecting beam for RC wall panels

3. EXPERIMENTAL STUDIES ON PRECAST WALLS

Significant efforts have been made by the researchers to evaluate the performance of precast reinforced

concrete shear walls. Some of the studies are:

Twigden et al. (2017) conducted experimental investigation to study the response of post-tensioned walls

subjected to cyclic load test on two single rocking walls (SRW-A and SRW-B) and two precast walls with

end columns (PreWEC-A and PreWEC-B) with supplemental damping which is provided by the O-

connectors in the form of energy dissipation as shown in Fig. 7. Wall dimensions, confinement details and

initial post-tensioning were kept constant in all the walls. High strength bar of 15 mm diameter was provided

in SRW-A as PT tendon whereas in other three walls were provided with 15.2 mm diameter strands.

Concrete filled square hollow sections with width equal to that of the thickness of the wall were used as end

columns. The number of O-connectors used in PreWEC-A was 4 and in case of PreWEC-B was 6 and other

parameters were kept constant. The two SRW (SRW-A and SRW-B) with varying axial force ratio exhibited

imperfect bilinear elastic response till 2% lateral drift. PreWEC had greater hysteresis area when compared

to SRW. Also, the hysteresis area was higher for PreWEC-B than PreWEC-A due to more number of O-

connectors. Flexural yielding dominated the behavior of O-connectors and resulted in the initiation of

fracture at 3% lateral drift in case of both the PreWEC.

Figure 7: Wall cross section details (Twigden et al. 2017)



Jianbao et al. (2017) conducted an experimental investigation on seismic performance of prefabricated

RC shear walls connected with vertical seam and compared with similar monolithic RC shear wall, having

similar geometry i.e., 1.8 m high and 1.5 m wide with 250 mm thickness. Vertical seam between the wall

panels was provided with HRB400 ɸ10 @ 150 mm strengthening reinforcement. Displacement controlled

horizontal load was applied at the top of wall. The results showed effective transfer of load at seam

connection and well comparable with monolithic wall in terms of lateral load resistance and stiffness

degradation. Although, higher displacement was observed in seam connected wall, which resulted in better

energy dissipation capacity.

Xilin et al. (2016) conducted an experimental study to evaluate the performance of joint beam

connection. Two 200 mm thick precast walls of size 2.6 m high and 1 m wide were connected through a

horizontal beam with varying depth of connecting beam. The systems were compared to the geometrically

identical cast-in-situ wall. Fig. 8 shows shear mode of failure at the base of precast wall, while the

connecting beam was intact. Load bearing and deformation capacity of precast walls was slightly inferior

when compared to the cast-in-situ walls. However, no deviation was observed in stiffness degradation and

ductility of tested walls. It was concluded that the extent of damage can be controlled by increasing the depth

of connecting beam.

Figure 8: Damage in precast wall connected with joint beam (Xilin et al. 2016)

Sritharan et al. (2015) carried out an experimental study on precast wall connected with end columns

through oval shaped mild steel energy dissipating 204 mm long and 31.8 mm thick O-connectors, shown in

Fig. 9. The wall and end columns were anchored into the foundation through unbounded post-tensioning

strands. A 6.1 high precast wall with 1.83 m width and 150 mm was connected to 200 x 150 mm columns at

both the ends through 5 pairs of O-connectors. Fig. 10 shows the cross-section and reinforcement detailing of

the wall connected to columns by the means of O-connectors. Reverse cyclic load was applied at the top

through hydraulic actuator and was continued till the drift of 3.5% was achieved. The performance of the

system was compared to a similar monolithic cast-in-situ wall and it was perceived that the precast system

performed extremely well by demonstrating 38% higher elastic stiffness and 12-17% higher lateral load

resistance. Failure mode was observed to be ductile and the system effectively contributed in energy

dissipation.



Figure 9: O-connector

Figure 10: Typical cross-section of wall connected to columns with O-connectors (Sritharan et al. 2015)

An experimental study on vertical loop bar connection was carried out in which performance of two

precast wall panels connected with vertical loop bars was evaluated under shear load (Rossely et al. 2014).

Reinforcement bars with 8 mm diameter were utilized to form loops which were connected to a 10 mm

diameter transverse bar in the joint. Both the walls were 0.6 x 0.6 m in length and width with 125 mm

thickness and M30 concrete. Connection details are shown in Fig. 11. Precast walls exhibited a maximum

shear stress of 1.83 N/mm2 at a displacement of 18 mm. Most of the concrete spalling and crushing was

observed at the joint and the walls demonstrated brittle failure mode. It was recommended to increase the

transverse bar diameter, embedded length of loop bars and strength of concrete in the joint to enhance the

ductility of the system.



(a) (b)

Figure 11: (a) Front view of walls connected with loop bars; and (b) Cross-section of the connection

(Rossely et al. 2014)

Analytical study on the behavior of loop bars was carried out by Vaghei et al. (2013) in which precast

wall to wall connection was simulated as a 3D finite element model in ABAQUS which is shown in the. Fig.

12. The wall panel was assumed to be 1200 mm in height, 600 mm in width and 125 mm in thickness, while

loop bar length was assumed to be 469.5 mm in length as shown in the Fig. 12. The concrete and

reinforcement were modeled as C3D8R and T3D2 elements respectively and the finite element model was

subjected to dynamic explicit non-linear analysis. Displacement controlled loading was increased from 0 to

10 mm at an interval of 2 seconds upto 20 seconds. Response of loop bar connection in walls was studied

with respect to deformation, maximum principal stresses and plastic strain in concrete and steel.

Reinforcement in the left panel exhibited maximum principal stress of 284 N/mm2, while concrete

demonstrated maximum principal stress of 30 N/mm2 and maximum deformation of 6.48 mm. It was

perceived that the interface damage initiated in the upper portion of the panel as upper hooks has higher

stress concentration as compared to lower hooks. However, crack propagation initiated at the bottom of the

wall, starting from left panel and propagating towards right panel. Exact behavior of loop bar connection

could not be understood by this study and requires further experimental investigations to clarify the response

of loop bar connections.

Figure 12: (a) Geometry of panel and hook; and (b) Reinforcement detailing in the panel (Vaghei et al. 2013)

Ashok et al. (2013) explored the concept of dowel action and shear friction for horizontal connection that

transfers shear forces between precast wall panels. An experimental study was conducted in which two 0.5 x

0.4 x 0.1 m precast RC wall panels with M35 grade concrete were connected using 12 mm dowel bars at 150

mm spacing and the gap between the panels was filled with the grouting material. Load was applied through

a hydraulic jack of 50 tonn capacity. The system exhibited an ultimate load of 180 kN with 25 mm

displacement. Minor cracks were observed near the connection, attributed to the bond failure between the grout and wall panel. However, behavior of this type of connection has not been compared with

the cast-in-situ wall. Thus, performance of this type of connection is still ambiguous and requires further

research.



Figure 13: Connection using dowel bar

In an experimental study performed by Brian et al. (2013), seismic performance of precast RC wall

connected to the foundation by the means of Grade 400 mild steel bars and high strength unbonded post-

tensioning strands was evaluated and compared to the similar monolithic cast-in-situ wall. In order to reduce

the strain in strands and to keep the ducts away from the critical base panel ends, the post tensioning

reinforcement was placed near the centre line of the wall. Reverse cyclic lateral displacement controlled load

along with centrally placed downward axial load was applied on 0.4 reduced scale models with height to

length aspect ratio of 2.25. The whole experimental set-up is shown in the Fig. 14. The precast wall achieved

load carrying capacity and drift of 551 kN and 2.3% respectively as compared to 534 kN and 1.15% for cast-

in-situ wall. The cast-in-situ wall developed higher residual uplift at the base joint which lead to strength

degradation and excessive horizontal slip. On the other hand, precast system with unbounded post tensioning

strands exhibited better energy dissipation, superior ductility and higher horizontal displacement.

Figure 14: Experimental set-up for precast wall with PT strands

Another experimental study on the behavior of post tensioning tendons was performed by Holden et al.

(2003), in which a precast reinforced concrete wall was compared to a geometrically identical cast-in-situ

concrete wall. Precast concrete wall was connected to the foundation with unbounded post-tensioned carbon

fiber tendons and additional low strength tapered longitudinal reinforcement is provided to increase energy

dissipation. Both the walls possessed aspect ratio of 2.7 and 125 mm thickness which were constructed using

concrete of strength 40 MPa. The models were subjected to quasi-static reverse cyclic lateral loading. Test

results demonstrated that the precast wall achieved 3% drift as compared to cast-in-situ wall which achieved

2.5% drift. Latter dissipated more energy than the precast wall; however permanent structural damage was

seen in the plastic hinge region at the base of the cast-in-situ wall. Moreover, cast-in-situ wall exhibited

significant residual deformations and cracks upon unloading, which was not observed in the case of post-

tensioned precast wall, indicating satisfactory behavior of the system.

Perez et al. (2002) studied the lateral load behavior of precast reinforced concrete walls with unbonded

post-tensioning tendons. Wall panels along the height were connected through unbonded post-tensioning

tendons and the base panel was provided with the confinement reinforcement to sustain higher compressive

strains that may generate due to the lateral load. Six storey, 5/12 scale-down walls with 9.91 m height and



2.54 m width were subjected to static monotonic and cyclic lateral loading. The base shear capacity of 687.7

kN and 671 kN was obtained through monotonic and cyclic loading respectively. Upon cyclic loading, the

precast wall demonstrated non-linear elastic behavior along with the minor energy dissipation in each cycle

of lateral loading. Excellent self centering behavior of the wall was observed upto yielding of post-tensioning

steel, which concludes satisfactory performance of post-tensioned walls during lateral loading. However, as a

result of axial flexural compression, fracture of confining reinforcement was observed as shown in Fig. 15.

Figure 15: Observed failure mode in RC wall with PT tendons (Perez et al. 2002)

Rizkalla et al. (1989) performed an experimental investigation to study the behavior of shear keys

used in horizontal connections for post-tensioned shear walls. Wall panels with and without shear keys were

tested under static shear loading to study the behavior and effect of shear key in the wall panel. The whole

set-up of wall panels and connection is shown in Fig. 16. Presence of shear keys enhanced the shear capacity

as wall panel with shear keys exhibited 60% more maximum shear capacity in comparison to the wall panel

with plane surface. Whereas, ultimate shear capacity of shear connection was 25% higher as that of plain

surface connection. Higher shear capacity is attributed to the interlocking action of shear keys with dry pack.

Figure 16: Shear keys along the joint

4. CONCLUSION

Precast construction is still not very common in India, attributed to the lack of knowledge on the joint

connections, their design acceptability, mass-scale production and competent agencies etc. Prefab

construction when adopted on tall buildings in high seismic regions, then it is essential to provide RC shear

walls. However, the behavior of precast RC shear wall and its behavior under lateral loads and connection of

shear wall-column have not been explored properly. IS: 11447-1985 and IS: 15916:2010 codes do provide

design and construction guidelines for prefabricated concrete structures but lack design and details for their

connections. The experimental studies indicate that the base of the precast wall and the region adjacent to the

joint are most prone to the stress development. These regions demonstrate damage initiated in the form of



cracks, concrete spalling and crushing which later results in extensive slip and permanent deformation at

higher loads. Thus, connections are the most crucial part of precast structures that are required to be designed

accurately in order to effectively transfer the lateral forces occurring during the event of earthquake.

Several types of connections such as loop bar connection, joint connecting beam, reinforcing bars,

O-connectors, shear keys, vertical interlocking joint etc. have been investigated to connect precast walls with

columns, beams and foundation. It can be deduced that the loop bar connection and O-connectors are the

most effective connection types for precast walls as it exhibited significant load carrying capacity along with

lateral drift. However, more experimental investigation is necessitated to ascertain the behavior and response

of vertical interlocking joint, joint connecting beam and reinforcing bars. It was also found that the unbonded

post-tensioning strands were effective in energy dissipation and imparted remarkable ductility in the system.

The debonding of post-tensioning bars results in elimination of high bond stresses and associated tensile

cracks in the concrete near the bars. The use of unbonded post tensioning reinforcement in well-designed

precast concrete shear wall results in a system which can undergo significant nonlinear lateral drift without

jeopardizing their ability to self-centre and hence reducing the permanent lateral drifts. Thus, it can be

concluded that the precast wall panels connected with the loop bar connection or O-connectors along with

the provision of unbonded post-tensioning tendons are the most desirable precast structural system, leading

to excellent structural performance when subjected to lateral loads. Loop bar connection is highly accepted

in precast industries around the globe; however, adoption of O-connectors is almost negligible as they have

their own limitations that may arise during actual implementation, explicitly due to welding, placement of O-

connector at the joint and lack of design specifications.

The literature review unfolds that much attention was paid to study in-plane behavior of individual

precast RC shear walls without significant research on their connection with columns, beams and slabs.

There is very little research on the out-of-plane behavior of precast walls and precast frames, thus behavior

of currently available connection types is still unclear when implemented at global level. Thus, full-scale

prefabricated building models should be investigated under lateral loads for intense study and more rigorous

research on the behavior and structural response of different connection types is required. Further,

formulation of adequate design guidelines for connections are essentially required in order to built up

confidence among engineers to accept and emulate precast technology with cast-in-situ construction.

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tensioned precast wall systems." engineering structures 31, no. 12 (2009): 2966-2975.

[2] Bora, Can, Michael G. Oliva, Suzanne Dow Nakaki, and Roger Becker. "Development of a precast

concrete shear-wall system requiring special code acceptance." PCI journal 52, no. 1 (2007): 122.

[3] Henry, R. S., S. Aaleti, S. Sritharan, and J. M. Ingham. "Design of a shear connector for a new self-

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[4] Heuvel, J. S. Multiple Shear Key Connections for Precast Shear Wall Panels.

[5] Holden, Tony, Jose Restrepo, and John B. Mander. "Seismic performance of precast reinforced and

prestressed concrete walls." Journal of Structural Engineering 129, no. 3 (2003): 286-296.

[6] IS 11447-1985, Code of Practice for Construction with Large Panel Prefabricates. Bureau of Indian

Standards, New Delhi.

[7] IS 15916-2010, Building Design and Erection using Prefabricated Concrete – Code of Practice. [8]

Bureau of Indian Standards, New Delhi.

[8] ISBN 978-0-642-32784-0, National Code of Practice for Precast, Tilt-Up and Concrete Elements in

Building Construction, The Australian Safety and Compensation Council, 2008.

[9] Joergensen, Henrik B., and Linh C. Hoang. "Tests and limit analysis of loop connections between

precast concrete elements loaded in tension." Engineering Structures 52 (2013): 558-569.

[10] Junbao, Hao. "Structural behaviour of precast component joints with loop connection." PhD diss.,

2004.

[11] Kurama, Yahya C., and Qiang Shen. "Post-tensioned hybrid coupled walls under lateral loads."

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285-290.



Confined Masonry Construction for Mass Housing

Ajay Chourasia

1 , Ankush Bhalke

2 and Shubham Singhal

3 1Principal Scientist, CSIR-Central Building Research Institute, Roorkee, E-mail: [email protected] 2M.Tech. Student, Vellore Institute of Technology, Vellore, E-mail: [email protected]


ABSTRACT

Masonry constructions are the pervasive building stock in India. Nonetheless, in some of the cases,

such constructions are designed for gravity loads only without any engineering measures for earthquake

resistance. As a result, masonry buildings suffer widespread damage even at moderate ground shaking. This

has necessitated for alternative building technologies with improved seismic performance. Confined masonry

(CM) construction shows better promise as a technology that performs well in earthquake, if built properly..

However, there are a very few experimental efforts to seismically evaluate the technology in the country and

lack of Indian standards, which restrains to adopt the technology. The intent of this paper is to analyse

experimental data and performance of CM buildings in major earthquakes, world over, and highlight the

comparison of performance of different types of masonry buildings viz. unreinforced masonry (URM),

reinforced masonry (RM) and confined masonry (CM), tested on a full-scale model under lateral quasi-static

loading, in Indian context. The masonry models are of 3.01x3.01m size in plan with a height of 3.0m using

solid burnt clay bricks of size 220x110x70mm in 1:6 cement-sand mortars with 220 mm wall thickness. To

examine economic aspects of CM building, ensemble of typical housing in India, were designed as RC,

URM, RM and CM, for the uniform design parameters. The construction costs were computed for different

structural elements and comparison of each typology was performed with reference to the construction cost

of RC building. The results shows that CM, RM and URM buildings allows for average cost reduction of

structure by 30%, 33% and 36% respectively, as compared to the RC framed structure.

Keywords: Confined masonry, Stone Concrete, Autoclave Aerated Concrete, Mass Housing.

1. INTRODUCTION

Masonry still find its wide use in today’s buildings, in low-to-medium rise construction, than any other

material. The success of brick masonry, in particular, is mainly due to its sustainability, durability, fire

resistance, acoustic and thermal insulation characteristics and relative simplicity of realisation. However,

Unreinforced masonry (URM) buildings, have proven vulnerable in seismic events, with significant

damages in buildings and numbers of fatalities, world-wide, including India. To increase the seismic

resistance of masonry, different methods for reinforcing masonry have been attempted over the years, and

led to the concept of confined masonry (CM) and reinforced masonry (RM) systems.

The issue of seismic performance and safety of existing masonry buildings is characterized by various

uncertainities. This paper presents focuses into the subject of confined masonry, performance of CM

buildings in major earthquakes, analysing and comparing experimental data on masonry buildings, in Indian

scenario and probing into economical aspects. It is expected that this paper stimulates towards confined

masonry as a structural system, in India, as well.

Increasing demand for flexibility, comfort and energy saving in industrial and residential buildings has

led both manufactures and designers to adopt new construcctive systems. Extensive work has been done on

brick masonry; however, there is a need of carrying out research work using stone and AAC blocks in order

to minimise the use of natural resources.



2. MASONRY CONSTRUCTION SCENARIO IN INDIA

Masonry construction is commonly adopted in India, both in urban and rural areas. Masonry

construction holds benefit of locally available material, economical, simple construction and resistance to

fire. According to the Census of India in 2001 and 2011 (Housing data), the distribution of houses based on

predominant materials of wall shows that, in India, there were 249 and 304 million houses in the year 2001

and 2011 respectively, comprising around 85% masonry houses. Also there is decline in proportion of

mud/unburnt bricks, wood, GI/metal sheet houses in 2011 as compared to 2001, with appreciable increased

use of burnt clay units in masonry. Due to socio-economic constraints some of the buildings are with unburnt

solid clay bricks or mud walls of 450-600 mm thickness upto two storey as load bearing walls. Previous

earthquakes have highlighted the inherent weaknesses of this type of construction and offer vivid

demonstration of its vulnerability. Moreover, masonry construction in India is limited to the use of burnt clay

bricks, while stone and AAC block masonry construction is almost neglegible.

The wide band of variability in construction material, its mechanical properties, and workmanship

exists across the country for masonry construction, which pose challange to characterize the behavior of

seismic of such buildings in a quantifiable manner. For example bricks in Gangetic belt have elastic moduli

in the range of 1500 to 4000 MPa (compressive strength varies between 10 to 19 MPa) while bricks in

southern part of India have moduli in the range of 400 to 1000 MPa (compressive strength between 3 to 9

MPa). This must be compared with the strength of brick in the US and UK, which has average compressive

strength of bricks as 100 and 75 MPa respectively.

The excessive use of cement based mortar (cement:sand, cement:stone dust:sand) have led gradual

exclusion of lime mortar in recent construction. The mortar composition for masonry varies based on wall

thickness, construction practice etc. Generally, cement-sand mortar of 1:6 proportion by volume is adopted

for 220 mm thick masonry walls while richer mix of 1:4 is used for 115 mm thick non-load bearing

(partition) walls. The mortar thickness in masonry ranges between 10-15 mm in masonry works. The roof of

such construction are either of wooden truss with GI sheets or clay tile or RCC slab, while floors are either

of RCC slab, or wooden logs (as beam) with mud/RC floors, simply resting over walls. The majority of

masonry construction is built by rules of thumb and traditions of construction technology that are handed

down from one generation to the next. This has resulted in increasing vulnerable building stock in the

country, and opening a large window for a promising masonry construction technology, confined masonry,

which performs well in seismic events, if built property.

3. CONFINED MASONRY

Confined masonry consists of unreinforced masonry wall panels surrounded by lightly reinforced horizontal

and vertical “confining” RC members. In some cases, the masonry units are staggered or “toothed” at tie

column locations to create better interlock between the masonry and RC member. In confined masonry

buildings, masonry walls are erected first and concrete in column is casted later. The sequence of

construction of confined masonry building is shown in Photo 1. The RC slab is adequately connected with

tie-column and to confine masonry between lintel and roof level. It is preferred to provide tie beam at lintel

level. The structural detail of typical CM building, as per EC8 requirements is shown in Fig.1 along with its

view in Photo 2.



Figure 1: Structural details of Confined Masonry building

Photo 2: View of confined masonry building

3.1. Performance of Confined Masonry in Past Earthquakes

In a worldwide review on performance of confined masonry buildings in past major earthquakes showed

that it performed satisfactorily within the framework of seismic design philosophy (Photo 3). A few of the

major earthquakes around the world: Colombia Earthquake (25 January 1999- Mw=6.2), Mexico Earthquake

(a) (b) (c) (d) (e ) (f) Photo 1: Sequence of construction of confined masonry building

(a) Construction of masonry wall with provision of reinforcement in tie column (b) providing shuttering on

two faces of tie column (c) casting of tie column followed by subsequent masonry (d) provision of keys in

concrete and masonry for better bonding of concrete with masonry (e) subsequent shutting of tie column (f)

completed confined masonry model



(30 September 1999- Mw=7.5); El Salvador Earthquake (13 January 2001- Mw=7.6); Mexico Earthquake (21

January 2001- Mw=7.6); Atico, Peru Earthquake (23 June 2001- Mw=8.4); Pisco, Iran Earthquake (26

December 2003- Mw=6.6); Great Sumatra Earthquake (26 December, 2004- Mw=9.3); Peru Earthquake (15

August 2007- Mw=7.9); Colima, Chile Earthquake (20 February, 2010- Mw=8.8) and Iquique, Chile

Earthquake (1 April 2014 – Mw=8.2) have demonstrated the performance of CM buildings with good and

poor construction practices. Damage data reveals that the typical damage patterns are: shear failure of

walls; shear and bending failure at ends of tie-column; separation of tie column from walls; inadequate wall

densities in two orthogonal directions, and development of first storey mechanisms (Photo 4). In some of

the cases, damage occurs at upper storeys of the building, with associated out-of-plane damage, mostly due

to absence of integral box behaviour of the storey.

The predominant reasons of failure in CM buildings are attributed to: missing / largely spaced tie

columns; inadequate anchorage of reinforcement of tie beam and column; largely spaced lateral ties in

column; large aspect ratio of masonry panel; asymmetric distribution of walls in plan; inadequate wall

density, poor workmanship, poor quality of materials used, and gross construction errors. None of the case of

foundation failure of CM buildings has been reported. Nevertheless, confined masonry construction, if

constructed properly, has generally shown a good seismic performance and no significant damage occurred

during past earthquakes.

Photo 3: Good performance of Confined Masonry construction in Earthquakes – (a) Six-storey confined masonry

building in Ica, 2007 Peru Earthquake (EERI, 2007); (b) No Damage to confined masonry buildings, while collapse of

other masonry buildings in El Salvador, 2001 San Salvador Earthquake (EERI 2001); (c) Six-storey confined masonry

building remained undamaged in 2007 Pisco (Peru) Earthquake (Blondet, 2007)

Photo 4: Damage to confined masonry buildings – (a) In Llolleo, 1985 Chile Earthquake (Moroni,

Gomez, and Astroza, 2003); (b) In El Salvador, 2001 San Salvador Earthquake (Yoshimura and Kuroki

2001); (c) In Mexico, 2003 Colima earthquake (EERI, 2003); (d) In Mexico, 1999 Tehuacan Earthquake

(a) (b) (c)

(a) (b) (c)

(d) (e) (f)



(EERI, 1999); (e) Collapse of Confined masonry with soft stories, relevant irregularities and bad detailing in

2007 Pisco (Peru) Earthquake (EERI, 2007); (f) In Banda Aceh, Indonesia, Tsunami-induced out-of-plane

failure of masonry walls at the ground floor level after 2004 Great Sumatra earthquake (Boen, 2005).

3.2. Experimental tests on Confined Masonry

The behaviour of confined masonry walls under lateral cyclic loading have been widely evaluated by

several researchers such as Tomazevic, M. et al. (1988, 1997, 2000, 2004, 2007, 2009); Wijaya, W., et al.

(2011); Gouveia, J. et al. (2007); Yoshimura, K. et al. (1996, 2000, 2003, 2004); Zabala F. et al. (2004);

Yanez F. et al. (2004); Marinilli, A. (2004); Kumazawa, F. et al. (2000); Aguilar, G. et al (1996); Meli, R.

(1973); and Umek A. (1971). The examples of full-scale test on shake-table are by Tomazevic, M. et al.

(1996), Kazemi, M.T. et al (2010), while quasi-static test procedure was adopted by Agarwal, S.K. et al.

(2007) for URM and RM models and Chourasia, A. et al. (2013, 2014) for CM model.

The review of experimental results and performance of CM buildings in past earthquake shows a

complex global behaviour. The diverse behaviour of the reported results is mainly due to diagonal shear

failure, however,in some cases flexure failure at initial stage within elastic limit has been noticed which may

be attributed to low vertical loads. More interestingly it is observed that, in higher number of storey in CM

buildings, deformation and damages are concentrated at first storey showing shear failure (Tomazevic,

M.,2007). It is also noted that failure mechanism is strongly dependent on horizontal reinforcement ratio,

leading to uniform distibution of cracks in masonry. In general, the brittle behaviour of hollow clay bricks

/concrete block has been observed as compared to solid clay brick units. However, different CM buildings

are of varying material and geometrical configuration, local tradition, and are not fully representative of

Indian architecture. For this reason, a comprehensive masonry test programme was undertaken in Indian

context at CSIR-CBRI, aiming to evaluate the seismic behaviour of indigenously built masonry buildings

during earthquake. A full-scale test on one room size masonry model 3.01x3.01m in plan and 3.0m high, has

been conducted under quasi-static cyclic lateral displacements of different types of construction practices

prevalent. The three types of masonry buildings tested are in unreinforced masonry (URM), reinforced

masonry (RM) and confined masonry (CM). The experimental load-deflection envelope of different masonry

models i.e. URM, RM and CM, is shown in Fig. 2. The overall observation shows major improvements in

seismic performance of CM building over URM and RM. Some of the features are: increase in strength and

ductility; enhancement in connections between walls; improvement in stability, integrity and containment of

masonry walls; and higher energy dissipation capacity.

Figure 2: Comparison of average lateral load-deformation envelope for different masonry systems



As confined masonry building construction uses the same materials and techniques, to that of URM with

higher level of safety, there is ample opportunity to adopt this technology in India, as a feasible housing

alternative, however its economics need to be analysed in detail, as compared to other structural systems.

4. ECONOMIC ASPECT

Majority of the building stocks in India range upto 4 storeys, comprising different building typologies

viz. i.e. Reinforced Concrete framed structure with masonry infill (RC), URM, and RM. Adequate seismic

resistance along with minimisation of construction cost of building is one of the challenges to be addressed

by the structural engineer. The experimental results demonstrate the higher seismic resistance of confined

masonry (CM) buildings, as compared to URM and RM. Hence, to balance the strength, safety and

economy, CM may be adopted as appropriate solution. However, to clarify the economy in construction,

rigorous cost analysis is warranted.

To carry out economic study of different building typologies in Indian buildings, 20 complex building

plans ranging upto 4 storeys were considered. Fig. 3 shows a typical plan of a building consisting of living

rooms, kitchen, stair-case and balcony etc., which is commonly adopted building layout in India, with a

storey height ranging between 3 to 3.50m. These buildings were designed as RC, URM, RM and CM for

common design paramters i.e. seismic zone – IV (PGA = 0.24g), live load (2 kN/sqm), and founded on soil

having safe bearing capacity of 100 kN/sqm at 1.50m from natural ground level. Similarly, uniform material

properties viz. grade of concrete (M20), grade of reinforcement (Fe415), masonry (compressive strength -

3.5 MPa, in cement:sand (1:6) mortar with 19.2 kN/cum as masonry density) were considered in the design.

Confined masonry buildings were designed with three different features. Firstly, the buildings were

comprised of only tie-column and bond-beams (CM1), secondly, in CM2 building with additional feature of

RC element around openings. The CM3 building consists of RC elements around opening for full

height/width of the panel and 1 number of 8 mm dia, horizontal reinforcement in mortar joint of masonry at

every fourth course. A typical details illustrating the various options of confined masonry considered for

deriving economic aspects are given in Fig. 4.

The RC buildings were designed in accordance with the relevant Indian standards viz. IS-456:2000, IS-

875:1987, IS-1893:2002, and IS-13920:1993. Similarly, URM, RM and CM buildings were designed as per

IS-4326:2013, IS-1905:1987, IS-456:2000, IS-875:1987, and IS-1893:2002. In addition, EC6 (2005) was

also referred in the design of CM buildings. The detailed estimation of quantity of each building sample was

carried out for different items and their costs are calculated based on prevailing market rates in India and

CPWD-Delhi Schedule of Rates (DSR) (2014).

Fig. 5 shows the average overall construction cost along with cost of major items for different building

typologies. To have more clarity in cost comparison, the values are expressed in terms of percentage of total

cost of RC building, as a reference. As can be seen that URM costs 64.4% to that of RC building costruction

cost. Similarly, RM, CM1, CM2 and CM3 costs an average of 67.6%, 69.33%, 70.76% and 71.68%

respectively. The figure indicates that average cost of construction of foundation is almost similar in case of

URM, RM and CM while it is slightly higher for RC buildings. However, higher cost component are

involved towards reinforcement and concrete for RC buildings.

Based on the above analysis, it can be summarised that CM, RM and URM buildings allow for average

cost reduction of structure by 30%, 33% and 36% respectively, with reference to RC framed buildings.

However, CM offeres significant amount of saving as compared to construction cost of RC building,

contrary CM assures higher level of safety when compared with URM/RM buildings.

AAC is a lightweight product which offers excellent sound and thermal insulating properties. Due to its

lower density it reduces dead weight on structure and can be used as non-structural applications, especially

cladding and infill panels which helps in saving cost.

Where as, the Stone Concrete blocks consists of mortar and rubble stones. Rubble stones are easily

available and this leads to economy.



Figure 3: Typical plan of a building for economic analysis



(A) :OPTION - I

Figure 4(a) Typical details of various options incorporated in confined masonry



(B):OPTION - II

Figure 4(b) Typical details of various options incorporated in confined masonry



(C): OPTION - III

Figure 4 (c) Typical details of various options incorporated in confined masonry



Figure 5: Average construction cost of masonry buildings with reference to RC framed structure

Conference on Infrastructure Sustainability in Hilly Regions (CISHR), 21st

-23rd

Dec. 2017


5. CONCLUSION

The goal of the present paper is to develop a framework that provides the essential information to

construct confined masonry buildings with good seismic resistance, considering the scenario of masonry

buildings in India. To meet the objectives, extensive reported experimental data and damages of CM

buildings in major earthquakes, are analysed to express its seismic behaviour. Also, the test results of full-

scale single storey masonry buildings viz. URM, RM, CM, in Indian context, under lateral cyclic quasi-static

loading have been taken into account. To demonstrate economic aspects of CM building, ensemble of

typical housing in India, were designed as RC, URM, RM and CM, for the uniform design parameters. The

conclusions drawn are:

1. In the present masonry building scenario and its vulnerability in India, confined masonry shows much

promise as a technology that performs well in earthquakes, if built properly.

2. Confined masonry construction, if constructed properly, generally showed a good seismic performance

and no significant damage during major earthquakes, worldover.

3. The failure mechanism of CM buidling under seismic actions is mainly due to diagonal shear failure.

Flexural failure at initial stage within elastic limit occurs due to low vertical loads.

4. In 3 to 5 storey CM buildings, deformation and damages are concentrated at first storey showing shear

failure, hence calls for adequate checks for shear.

5. CM buildings exhibited higher strength and ductility as compared to URM and RM buildings. The

performance of Indian CM buidlings over URM and RM in terms of strength showed about 3.42 and

2.63 times improvement respectively.

6. CM, RM and URM buildings allows for average cost reduction of structure by 30%, 33% and 36%

respectively, to that of RCC buildings. However, CM offers reasonable saving when compared with the

construction cost of RCC building and offers higher level of safety when compared with URM/RM

buildings.

7. Compare to conventional concrete, AAC is typically a lower density (which inturn reduces the seismic

inertial forces acting on structure) ranging from one-sixth to one-third and by a lower compressive

strength which is almost reducing the same ratio. Therefore it is recommended in future that bricks can

be replaced by AAC blocks units.

8. As rubble stones are easily available and to minimize the natural resource this stone concrete block

becomes economical in saving cost.

Extensive work has been done on brick masonry; however, there is a need of carrying out research work

using stone and AAC blocks in order to minimise the use of natural resources. Manufacturing of bricks

affects green environment by the emission of CO2. Therefore, focus shall be on seeking eco-friendly

solutions for greener environment. Material cost, energy consumption and carbon emission parameters

help in highlighting suitable options for sustainable construction. AAC blocks and stone concrete blocks

are eco-friendly materials that give a prospective solution to building materials.

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block walls reinforced with steel fibers. Construction and Building Materials, 152, pp.394-405.



Investigation on Headed Bars as Anchorage Device in Beam-

Column Joint- A Literature Review

Ajay Chourasia1, Shipali Gupta

2 and Shubham Singhal

3

1Principal Scientist, Central Building Research Institute, Roorkee, Email: [email protected]

2M.Tech Student,Vellore Institute Technology University,Vellore, Email:[email protected]

3Research Assistant, Central Building Research Institute, Roorkee, Email:[email protected]

ABSTRACT

Major failures in the structure occur in the joint rather than elsewhere in the structural elements, for

which the primary cause is bond strength of the joint. Conventional method for strengthening beam-column

joint is the provision of development length in reinforcing bars. However, this method has many

disadvantages such as steel congestion, slip of bar, honey-combing, corrosion etc. Extensive research has

been carried out on headed bars as an alternative solution to hooked bars in terms of its shape, size, bearing

ratio, embedment depth, bond capacity and slip. This paper presents a state of art literature review on the

parameters affecting the performance of headed bars.

Keywords: Beam- column joint, embedment depth, Headed bars, Pull-out capacity, single headed bars,

multiple headed bars.

1. INTRODUCTION

Beam -column joint is the point at the column where beam section is inserted. The most critical region in

the structure during design and construction stages is beam column joint. Past two decades many studies are

going on mechanical anchorage device in beam column joint. Currently used anchorage device is hooked

bars which creates many difficulties during design and construction stage. Due to the heavy steel provision

in joint hinders proper concreting leads to honey-combing, increases slip, and decreases bond capacity for

inadequate provision of development length due to restricted dimension of the member.

Headed bar, a deformed reinforcing bar with a head at its end. Simple installation, reduce congestion,

minimise slip, time saving and effectively increases the anchorage strength are the advantages of headed bars

over hooked bars. Use of headed bars simplified the design and detailing of reinforcing bars. Many

experimental researches has been performed to investigate the effectiveness of head and bar size, head shape,

head attaching technique and embedment depth of headed bars into the concrete, location of headed bars,

clear spacing of bar and clear cover of bar and its comparisons to standard hooked bars. The headed bar was

first introduced in ACI 318-08, using limited test data with strict strength and design data (ACI Committee

318, 2008). Increasing demand of headed bars made it necessary to perform deep investigation on it for

proper design provision. The performance of headed bars in pull-out test depends on various variables like

loading types, embedment depths, edge distances, plate types, single or multiple number of bars, distances

between bars, reinforcement strengths, concrete strengths and existing reinforcement. This study focuses on

the different parameter which affects the bond behaviour of headed bars and to find the gap to be

investigated later for recommendation in construction practices.

2. BOND AND FORCE TRANSFER MECHANISM

Bond can be defined as the interaction between the reinforcing bars and concrete require transferring

tensile load from steel to concrete. The bond between two materials is very essential to resist the external

load. The factors responsible for the bond strength between the deformed reinforcing bars and concrete are

chemical adhesion, friction and deformation over the surface of reinforcement. The minimum length of

reinforcing bar to develop its yield stress on application of external load is referred as development length.

Development length consists of bond length and bearing length, both together resist the tensile load.

When there is insufficient space for provision of bonded length of bar, the bar is terminated in the form of

hook or head which add bearing force to bond force.



(a) (b) (c)

Figure 1: Typical diagram of bond transfer mechanism (a) hooked bars (b) headed bars (Devries, 2015) (c) Bond force

transfer mechanism (ACI 408R-3)

For the proper bond the deformed bar is more efficient as compared to smooth reinforcing bar attached

to head as the bond of smooth bar is zero and deformation increases the anchorage strength .(ACI 318-08).

(Wallace et al., 1990) performed a cyclic load test on beam column joint made with double headed

bars for high seismic zone and found that the anchorage capacity of headed bar was far better than standard

hooked bars and hence recommended the minimum anchorage length of 12db and net head bearing ratio as

4.The increasing demand of use of headed bar, proper provision of development length of headed bar in

tension is presented by ACI 318-08:

ldt= 0 .19 ψe fy db / √ fc ≥ larger of 8 db or 152mm.

Where fy is the specified strength of the headed bars, fc is the specified strength of the concrete db is

the diameter of the deformed bar and ψe =1.2 for epoxy coated reinforcement and 1 for other cases.

Diameter of the head bar is not considered in the equation of the development length is the major drawback.

The minimum development length should not be less than 150mm (ACI 352R-02/ASTM

A970).Development length is very important when large numbers of highly stressed bars are embedded into

the concrete tends to create weakened plane and cause longitudinal cracks to propagate along the bars (Kim

et al. 2011). The load bearing ratio was assumed constant for both bond length and bearing length, further

studies found that the load bearing ratio influence by the bonded length. If the bonded length is more the

majority of the load resisted by bonded length else as the slip of bar occur the remaining load will be carried

out by bearing length (DeVries, 2015).

The longitudinal headed deformed bar provided in beam or slab, terminates at the far face of confined

are of supporting member to prevent the column reinforcement interference and helps to anchor compressive

force likely to form a connection with the improvement in the performance of the joint. The transverse

reinforcement provision along with the headed bar is recommended to help in limiting the splitting crack at

the vicinity of the head.

3. PARAMETERS AFFECTING PERFORMANCE OF HEADED BARS

3.1 . Bearing Ratio

As per ACI 352R-02, the net head bearing ratio should be at least 9 (ASTM A970) where as in ACI 318-

08 the net head bearing ratio should be at least 4. Head size recommended by ASTM A970 allows the

transfer of forces along the bar into the multi-axial stresses in the concrete without crushing the concrete and

bending the head. The anchorage capacity of bars depends linearly on net bearing ratio of headed bars. As

the head area is directly proportional to pull-out capacity of headed bars (Bakir et al. 2002).

In order to resolve the problem of steel congestion in the joint, the use of large diameter head is

restricted and the researchers showed that increase in the area of head enhance the anchorage capacity with

negligible slip of head (Choi et al. 2006). In high seismic zone, the design and detailing of beam column

joint creates the problem of steel congestion due to the heavy reinforcement provision which can be

resolved using small head with bearing ratio less than 4, results in high seismic performance than

standard hooked bars (Kang et al.2010).



3.2 . Head Geometry

Researches are in progress on the shape and size of head plate square, circular, elliptical, and

rectangular.

(a) (b)

Figure 2: (a) Circular, Square, Rectangular and Elliptical Headed Bar (Park et al., 2003) (b) Headed Bar Embedded

Into the Concrete (Choi et al. 2006)

Test results on various shape of heads demonstrate that circular shape head plate is more efficient in

enhancing the pull-out capacity (Park et al., 2003).

3.3 .Single and Multiple Headed Bars

The effect of detailing of single headed bars with large diameter of deformed bars have large number

of disadvantages as compare to detailing with single headed bars with small diameter of deformed bars

provided in the tension side of the flexural member. Single headed large diameter bar causes splitting of

cover along the bar with the generation of high compressive stress at the vicinity of the heads and leads to

diagonal cracking shown in figure (Mihaylov , 2013)

(a) (b)

Figure 3: (a) Arrangement of Single Headed Large Bars In Beam (b) Diagonal Cracking Due To Compressive Stress at

the Heads (Mihaylov, 2013)

When multiple headed bars were pulled out at the same time, cracks appeared on the concrete face

starting from the head location and progressed upward creating an angle between 35º and 45º. When single

headed bar was pulled out, cracks did not appear on the face of the concrete a small cone was projected out

surrounded the headed bar and no bar fracture was also observed (Choi et al., 2006)

3.4 . Head Attaching Technique

The head attaching technique of head to bar influence the Pull-out capacity. Progressive researches

have been done on head attaching techniques such as friction welding, general welding, and threaded

technique. Test results showed that threaded technique is more effective compare to other technique in

relation to proper bond and interaction between the bar and anchor or head.



Figure 4:

Friction Welding Technique (b) General Welding

Technique Series 1 (c) General Welding Technique Series 2 (d) Threaded Technique

3.5 .Embedment Depth

The author carried out the investigation on the factors influencing the ultimate load carrying capacity

of concrete members with headed bars. Results anticipate that as embedment depth increases, the failure

surface increases with the enhancement of resistance to external loads. The embedment depth decides the

size of failure cone (Bakir et al. 2002). The embedment depth of headed bar is conveniently classified by

researcher as shallow embedment depth and deep embedment depth. When the ratio of embedment depth to

the clear cover of bar is less than 5 it is shallow embedment depth else deep embedment depth (Devries

1999, Thomson et al. 2005, 2006). When the headed bars are provided as main reinforcement both in beam

and column, embedment depth of 10 db is insufficient to develop bar yield stress placed closely (Choi et

al.,2006).

3.6 . Location of Headed Bars

The different location of headed bars in concrete member is corner, edge and centre. The ultimate load

carrying capacity of concrete member depends on the location of headed bars, which is highest at the centre

due to presence of large failure area and least at the corners as shown in the figure (Bakir et al. 2002)

Figure 5: Graph on Pullout Capacity vs Location of Headed Bars

3.7 . Edge Distance

The distance from the closest edge of the concrete to the centre of the bar is referred to as edge

distance (ACI318-08). Choi et al., 2002 studied that the headed bar at the edge of the concrete specimens

leads to side blow out failure.

3.8 . Clear Cover

(

a) (

b)

(

c)

(

d)



As per ACI 318, sec 12.6.1 the clear cover for the bar should not be less than 2db. On provision of side

cover 3db in an extreme beam column joint headed bars subjected to high seismic loads not showed any side

blow out failure (Wallace et al. 1990).

3.9 Clear Spacing

As per ACI 318, sec 12.6.1 the clear spacing between the bars should not be less than 4db. A test data

indicates that when the headed bars anchored in a beam ends are closely spaced at 2 db, no proper failure

was observed and is acceptable for longitudinal headed bars (Kim et al. 2011). When the heads are closely

spaced, crushing of concrete may occur due to overlapping of pull-out cone formation subjected to external

tensile load. The Pull-out capacity of individual headed bar reduces placed closely spaced in groups.

4. MODES OF FAILURE OF HEADED BARS

Researches showed various modes of failure on headed bars subject to Pull-out test. The test results are

compared with existing codes of concrete anchors. When the concrete tensile strength is higher than the

specified yield strength of bar or anchor ductile or yield failure of bar takes place. (ACI 349-97). Under

tensile load, concrete is subjected to constant tensile stresses at a cone angle of 45º from the bearing surface

of the head, projected cone area towards the free surface of the concrete is concrete tensile break out cone

failure shown in figure (a) (ACI 349-97). The CCD method considered some parameter for the failure of

headed bar under tensile loading as the tensile stresses on the concrete from the headed bar transfer making

an inclination angle of 35º between failure surface and concrete surface as shown in the figure(b). Failure

occurs as concrete splitting failure, concrete tensile break out cone failure, concrete tensile break out

pyramid failure, and pull through failure and bar yield failure.

(a) (b) (c)

Figure 6: (a) Concrete Break out Cone Failure (b) Concrete Break out Pyramid Failure

(c) Bar Yield Failure

5. CONCLUSION

Researchers investigated on various parameters and found that the pull out capacity of headed bars

influence with the head geometry, head attaching technique, spacing between the heads, clear cover of the

head. It can be concluded that circular shape is much effective as compared to other head shapes. Increasing

head bearing ratio increases the anchorage capacity of the headed bars but should be less than 4 as it reduces

the problem of steel congestion in beam-column joint. To alleviate the problem of large cone failure clear

spacing between the headed bars should be more than 4 times the bar diameter. Provision of headed bars

near the edge of the concrete leads to side blow out failure due to improper bond between the concrete and

headed bars. Single headed bar with small diameter deformed bar have more advantages over multiple

headed bars in terms of concrete cone failure. Individual headed bars leads to small cone failure with no bar

fracture failure whereas in multiple headed bars large concrete cone failure occur considering all the headed

bars at a time. Headed bar as main reinforcement in beam and column helps in reducing the compressive

stress by enhancing the bond between the bars and concrete, while as transverse reinforcement, it limit the

concrete crack splitting failure. Hence, headed bar can be used both as main reinforcement and transverse

reinforcement in beam and column.



REFERENCES

[1].DeVries, R.A., Jirsa, J.O. and Bashandy, T., 1999. Anchorage capacity in concrete of headed reinforcement with

shallow embedments. Structural Journal, 96(5), pp.728-737.

[2].Chaim, D.U., Hongzl, S.G. and Lee, C.Y., 2002. Test of headed reinforcement in pullout. KCI Concrete

Journal, 14(3), p.1102410

[3].Park, H.K., Yoon, Y.S. and Kim, Y.H., 2003. The effect of head plate details on the pull-out behaviour of headed

bars. Magazine of Concrete Research, 55(6), pp.485-496.

[4].ACI 408 Committee, 2003. Bond and Development of Straight Reinforcing Bars in Tension (ACI 408R-

03). American Concrete Institute, Detroit, Michigan, US, p.49.

[5].Thompson, M.K., Ziehl, M.J., Jirsa, J.O. and Breen, J.E., 2005. CCT Nodes Anchored by Headed Bars-Part 1:

Behavior of Nodes. ACI Structural Journal, 102(6), p.808

[6] Thompson, M.K., Jirsa, J.O. and Breen, J.E., 2006. CCT nodes anchored by headed bars-Part 2: Capacity of

nodes. ACI Structural Journal, 103(1), p.65

[7].Choi, D.U., 2006. Test of headed reinforcement in pullout II: deep embedment. International Journal of Concrete

Structures and Materials, 18(3E), pp.151-159.

[8].ACI Committee, American Concrete Institute and International Organization for Standardization, 2008. Building

code requirements for structural concrete (ACI 318-08) and commentary. American Concrete Institute.

[9].Kang, T.H., Ha, S.S. and Choi, D.U., 2010. Bar pullout tests and seismic tests of small-headed bars in beam-column

joints. ACI Structural Journal, 107(1), p.32

[10].Lam, K.M., Kim, W.S., Van Zandt, M. and Kang, T.H., 2011. An Experimental Study of Reinforced Concrete

Beams with Closely-Spaced Headed Bars. International Journal of Concrete Structures and Materials, 5(2), pp.77-

85.

[11].Van Mier, J.G.M., Ruiz, G., Andrade, C. and Yu, R.C., ANCHORAGE STRENGTHS OF LAP SPLICES

ANCHORED BY HIGH-STRENGTH HEADED BARS

[12].Mihaylov, B.I., Bentz, E.C. and Collins, M.P., 2013. Behavior of Deep Beams with Large Headed Bars. ACI

Structural Journal, 110(6), p.1013.

[13].DeVries, R.A., Load Distribution between Bond and End-Bearing for Hooked and Headed Bars in Concrete.

In AEI 2015 (pp. 269-278). [14].Gond, S. and Kulkarni, S.M., 2015. BOND STRENGTH BEHAVIOR OF HEADED REINFORCEMENT BAR

WITH VARYING EMBEDMENT LENGTH. [15].Shao, Y., 2016. Anchorage of Headed Reinforcing Bars in Exterior Beam-Column Joints (Doctoral dissertation,

University of Kansas).

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