journal of applied engineering sciences

JOURNAL OF APPLIED ENGINEERING SCIENCES

Volume 2 (15), Issue 1/2012

University of Oradea Publishing House

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EDITORIAL BOARD

EDITOR-IN-CHIEF: Corneliu BOB ([email protected]), Politehnica University of Timi�oara, Romania EXECUTIVE EDITORS: Sanda Monica FILIP ([email protected]), University of Oradea, Romania

Dan GOMBO� ([email protected]), University of Oradea, Romania Aurelian-Stelian BUDA ([email protected]), University of Oradea, Romania Marcela-Florina PRADA ([email protected]), University of Oradea, Romania

MEMBERS

József ÁDÁM – Budapest University of Technology and Economics, Department of Geodesy and Surveying, Hungary

Marian BORZAN – Technical University of Cluj-Napoca, Romania Alexandru C�T�RIG – Technical University of Cluj-Napoca, Romania

Daniel DAN – “Politehnica” University of Timisoara, Romania Petre DRAGOMIR – Technical University of Constructions Bucharest, Romania

Gabriela DROJ – University of Oradea, Romania Mihai ILIESCU – Technical University of Cluj-Napoca, Romania

Ludovic KOPENETZ – Technical University of Cluj-Napoca, Romania Maricel PALAMARIU – “1 Decembrie 1918” University of Alba Iulia, Romania

Johan NEUNER – Technical University of Constructions Bucharest, Romania

TECHNICAL EDITOR: Gabriela-Argentina POPOVICIU ([email protected]), University of Oradea, Romania

Aims and Scope: Journal of Applied Engineering Sciences (JAES) is a scientifical journal devoted to presentation and discussion of information on the ultimate issues in the civil, installations, geodesic, electrical and energetical engineering fields. The journal addresses news and various problems which such fields confronts both national and international level. JAES is designed for scientists, researchers (including doctoral students), engineers and managers, regardless of their discipline, who are involved in scientific, technical or other issues related in the journal domains. Emphasis is placed on integrated approaches. These approaches require both technical and non-technical factors. Even the dissemination and application of innovative information is very important, the implementation of existing literature in the JAES related topics and the adress’s contributions also requires a clear understanding from as many other scientific areas as possible.

UNIVERSITY OF ORADEA, FACULTY OF CONSTRUCTIONS AND ARCHITECTURE 4, Barbu �tef�nescu Delavrancea Street, 410058 – ORADEA – ROMÂNIA

www.uoradea.ro * http://arhicon.uoradea.ro * http://www.arhiconoradea.ro/JAES/HOME.htm * Phone/Fax: 004-0259-408447 Cover design by George-Lucian Ionescu

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Preface

The first number of the magazine JOURNAL OF APPLIED ENGINEERING

SCIENCES was published in 1997 under the name of Annals of University of Oradea – CONSTRUCTIONS AND HYDROEDILITARY INSTALLATIONS fascicle. Until 2010 the magazine has been issued annually, thus, it has reached its XIIIth consecutive edition. Since 2003 the magazine has been publishing scientific works presented within the National Conference – international event – MODERN TECHNOLOGIES FOR THE 3rd MILLENIUM, where valuable specialists have met, both from the major university centres in the country and world-renowned professors from universities from abroad. The national and international prestige of the magazine has been constantly increasing.

All scientific papers accepted to publishing are thoroughly analyzed by a scientific committee formed by Romanian and foreign university professors, internationally recognized in their area of expertise.

In 2009 the magazine has undergone an assessment by C.N.C.S.I.S. being rated in category B, and in June 15th 2010 it has been rated B+ by the same C.N.C.S.I.S., being accepted in two International Databases (IDB) – Ulrich’s and Copernicus. Since 2011, the Journal is registered on SCIPIO, the Romanian Publishing Platform website.

Since 2010 the magazine has been issued twice a year, in July (containing scientific papers of Romanian and foreign specialists) and one in November, usually containing scientific papers that had been presented at the National Conference – international event – MODERN TECHNOLOGIES FOR THE THIRD MILLENIUM, the VIIIth consecutive edition being held in 2010. Starting with 2011, the magazine is quarterly published under its new name, JOURNAL OF APPLIED ENGINEERING SCIENCES (JAES), and is in competition for ISI classification.

As a result of the various subjects treated so far in our magazine’s pages, from the Civil engineering and installations, Geodesic engineering, Electrical and energetical engineering in constructions fields, JAES are included in the large Civil Engineering area. It means that the magazine, through the category domain established above, includes resources on the planning, design, construction, maintenance of fixed structures and ground facilities for industry, occupancy, transportation, use and control of water, even harbor facilities. At the same time, resources may cover the sub-fields of structural engineering, geotechnics, earthquake and geodesic engineering, ocean engineering, water resources and supply, marine engineering, transportation engineering, and municipal engineering.

Editorial Staff

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Responsibility for content of the published material exclusively belongs to the authors, the auditing team’s role being to get fit and verify accuracy of those included in these works.

this Review is accredited by C.N.C.S.I.S., code 877, rate B+

(http://cncsis.ro/userfiles/file/CENAPOSS/Bplus_aprilie_2011%281%29.pdf) and it’s registered in the International Database Index Copernicus and Ulrichs

http://journals.indexcopernicus.com/masterlist.php?name=Master&litera=a&start=330&skok=30 http://www.ulrichsweb.com/ulrichsweb/ulrichsweb_news/uu/newTitles.asp?uuMonthlyFile=uu201

012/new_titles.txt&Letter=B&navPage=9& and also in the Scientific Publishing & Information Online SCIPIO Platform

http://www.scipio.ro/web/journal-of-applied-engineering-sciences

ISSN / ISSN-L 2247 – 3769 / e-ISSN2284 – 7197 University of Oradea Publishing House May 2012

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CONTENTS

Cîrstolovean Ioan Lucian Measurements on solar equipments to produce hot water in different climate situation

and determination. The equipments’ efficacity ..................................................................

7

Constantinescu Horia, M�gureanu Cornelia Shear design procedure for high strength concrete beams without stirrups ……………...

13

Dârmon Ruxandra Sustainable fire safety design for building frontages …………………………………...

19Domni�a Florin, Ho�upan Anca, Popovici Tudor Total air pressure loss calculation in ventilation duct systems using the equal friction

method .................................................................................................................................

25

Dub�u C�lin Gavril Calculation of kinematic characteristics throughout the range of conditions to maximize

the power extracted from wind in the case of small size turbine ........................................

31

Jumate Elena, Manea Daniela Lucia Application of X Ray Diffraction (XRD) and Scanning Electron Microscopy (SEM)

methods to the Portland cement hydration processes …………………………………...

35

Kopenetz Ludovic Gheorghe, C�t�rig Alexandru, Li�man Drago� Florin Problems concerning in situ bevahiour of complex structures …………………………..

43

Kopenetz Ludovic Gheorghe, Li�man Drago� Florin Realtime behavioural monitoring of cable transport structures …...................................

49Li�man Drago� Florin, Kopenetz Ludovic Gheorghe Advanced in situ monitoring techniques for the behaviour of heritage structures ……..

55

Lupan Lidia-Maria, Moga Ioan Linear thermal bridges at vertical elements of building structure ….................................

59

Molnar Iulia Correlations between geotechnical parameters of Transilvanian cohesionless soils based on triaxial laboratory tests results ………………………………………………………...

65

Mo�oarc� Marius, Stoian Valeriu Seismic energy dissipation in structural reinforced concrete walls with staggered

openings …………………………………………………………………………………..

71Peredi �tefan, Chira Alexandru Modal analysis with response spectrum for an unreinforced masonry structure using

Seismic Romanian Codes …………………………………………………………………

79

Pintea Augustin Shear capacity for prestressed-prefabricated hollow core concrete slabs, without shear

reinforcement …..................................................................................................................

83

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R�dulescu Adrian T.G., R�dulescu Gheorghe M.T. Management of structural monitoring using optical fibers ………………………………

89

R�dulescu Adrian T.G., R�dulescu Gheorghe M.T. The monitoring of Beska Danube Bridge, Novi Sad, Serbia, in a permanent quasi-static

regime …..............................................................................................................................

95

R�dulescu Virgil Mihai G., R�dulescu Corina MGIS (Mining Geographical Information System) a new concept for the

informatization management on mining companies. Some considerations on the MGIS field ……………………………………………………………………………………….

103

R�dulescu Virgil Mihai G., R�dulescu Corina MGIS (Mining Geographical Information System) a new concept for the

informatization management on mining companies. Introducing mining data in GIS and its transformation in MGIS .................................................................................................

109

Trifa Florin-Sabin Considerations on the calculus of carbon fibre reinforced polymer (CFRP) strengthened beams ……………………………………………………………………………………..

113

Trifa Florin-Sabin The influence of shear on the inelastic displacement of eccentric compressed reinforced concrete members ………………………………………………………………………..

123

Trofin Florin Climate change influence on hydrotechnical structures, existing and future .....................

133

Authors Index ……………………………………………………………………………..…

143Guide for authors ……………………………………………………………………………. 145

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MEASUREMENTS ON SOLAR EQUIPMENTS TO PRODUCE HOT WATER IN DIFFERENT CLIMATE SITUATION AND DETERMINATION.

THE EQUIPMENTS’ EFFICACITY

CÎRSTOLOVEAN Ioan Lucian, University Transilvania Brasov, Faculty of Buildings Engineering, e-mail: [email protected]

A B S T R A C T The paper represents a theoretical and experimental approach on solar equipment use to produce hot water. The capability is one of the fundamental criterion referring to the achievement of quality and performance of the system.wich produce thermal energy Before the introduction of the statistical control of thermal energy production, a compatibility analysis is done which is supposed to establish if the set tolerances by specifications are compatible with the installations’ capacity of satisfying these demands. We need to evaluate the performance of a thermal energy production system This is given by certain characteristics, especially important being those which represent exigencies of comfort, for example of the thermal one, and economic exigencies in exploitation, predominant being the fuel consumption and implicitly its cost.

Keywords: reliability, durability, solar equipment, capability, thermal energy

Received: January 2012 Accepted: January 2012 Revised: March 2012 Available online: May 2012

INTRODUCTION The heat load is at the basis of the dimensioning of the solar panels of the application we

want to integrate. We will determinate the monthly and annual values for heat load. The annual heat load of the energy necessary must be compared with the energy delivered by the sun. This rate rises in summer time and decreases in winter time. The annual evolution of solar energy is considerably influenced by the position of the panels field.

MATERIALS AND METHODS 1. Designing the solar system in the laboratory

In the laboratory of the faculty was designed a solar system for producing hot water for 3 persons. The geographical dates of the building where the solar panels were placed are:

• Latitude 45,66 grd; • Longitude 25,61 grd.

The necessary volume of hot water is determined with the relation: Vac= a * Nu/1000 [m3/zi] (1) where: a – specific necessary of hot water at 60 ˚C/ person for a day; Nu – number of persons from the building.

According to the standard EN 15316-3-1 (Heating Systems in buildings. Method of calculation of requirements and systems efficiencies) the values for a are informative. For a new building with reduced loss of hot water, it can be considered a= 60 l/ day* person. For the case under analysis, for 3 persons, the volume of hot water [9] at 60˚C is:

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V= 3*60/1000 (2)

V= 0,18 [mc/zi] (3) The volume V2 of the hot water tank at 450 C is:

V2= V �T1/�T2 [mc] (4) V2= 0,27 [mc] (5)

The volume of the tank that was chosen: 300 l The daily energy for preparing hot water is given by the relation:

Q= V*�*c*�T [J] (6) Q= 0,18*992,2*4182*45 [J] (7) Q= 33,6* 106 [J] (8)

The daily energy expressed in kWh/day is:

Q=9,33 [kWh/zi] (9) Annual energy:

Q= 3405 kWh/an (10)

The surface of the collecting solar panels [9] is determined with the relation:

(11) where: Acol – solar panels surface; Korient – the positioning factor;

SD – solar coverage degree SN – usage degree; Qcons – energy consumed for preparing hot water; Qrad – solar radiation / m2.

(12)

RESULTS AND DISCUSSIONS

This experimental study about the heating process of the water with solar panels was made in the faculty’s laboratory. The measurements were made and recorded in csv files with the help of the vDIALOG program. The program offers the possibility to visualize the temperatures in the areas where temperature sensors were placed and of the functioning pace of the pump. From the multitude of dates recorded, for the capability analysis, we analyzed the temperature of the hot water in the boiler by the SB1, SB2 sensors, the climate temperature and the water temperature from the solar panel measured by the KOL 1 sensor. The sensors’ position is presented in fig 1. The

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solar radiation was measured with Voltcraft equipment. For this paper the measurements were determined between the 1st of March 2011 and 30th of April 2011. From the data recorded by the computer, 9 days were chosen which were significant for the process of heating home water. The major criteria for choosing them were the diversity of the weather values and the intensity of the solar radiation. In this paper we present the graph recorded by the program and the graph made by the interpretation of the files recorded by the computer on the 5th of March 2011. 1. The capability of the process of heating home water by means of solar energy

The capability of the thermal energy process [2] (to prepare hot water) represents one of the fundamental criterion referring to the meeting of quality and performance of the system. Before the introduction of the statistical control of the thermal energy production, a compatibility analysis is done which is supposed to establish if the set tolerances by specifications are compatible with the installations’ capacity of satisfying these demands. The performance of a solar system’s equipments needs to be evaluated. This is given by certain characteristics.

Fig.1. The schema of the system of producing home hot water with solar panels

The statistic control [1], [2] of the process is meant to supervise the proceedings of the

process, essentially, its capacity and if the normal proceeding of the process contains deviations from the specifications in order to interfere in the process to have it corrected. A process is “under control” if it satisfies the demands imposed by the statistics of the control which are:

a) regulated process; b) precise process;

The production and distribution process of the energy is considered dynamically stable if the position and diffusion indices maintain in time the grouping centre and the diffusion field, according to the specifications.

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This paper presents a capability analysis of the process of heating the water from the solar boiler at a temperature between 30-60˚C. The schema of the system that helped with the measurements is presented below. With the results of measurements on the system the statistic calculus [2] was done. The steps were:

- from the number of the temperature values obtained by measuring SB1 there were chosen only those which are between 30 and 60˚ C. For the analyzed days, the number of these temperatures is N;

- we calculated the absolute and relative frequency with which identical values were recorded of the SB1 temperatures;

- locating parameters were determined; - dispersion parameters were determined;

The results obtained are presented in Table 1.

Table 1. The results of the statistic calculation

When analyzing the results we concluded: the arithmetical mean, median and the mode have

near values which means that statistic distribution of the measurement values is normal and actually symmetrical. The higher value of the dispersion demonstrates that we have a large area for temperature values. It is normal to be like this because the variation of solar radiation produced high variation of the hot water in tank. In the interval (xm-�; xm+�) (32,41; 41,25) there are 168 values out of 287, namely 58% from all values .

Minimal temperature: 1,5 °C- 9 o’clock Maximal temperature: 4.6 °C- 15 o’clock

Fig 2. The variation of the hot water temperature in the solar tank for March 5th 2011

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Sunday, 5 March 2011

Temperature min: 1,5 °C-ora 9 Temperature max: 4.6 °C-ora 15

Fig 3. The graphic made by vDIALOG programme for March 5th 2011

Sunday, 5 March 2011

Temperature min: 1,5 °C-ora 9 Temperature max: 4.6 °C-ora 15

Fig 4. The position of the sun in relation to the solar panels for March 5th 2011

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CONCLUSIONS The solar system for producing home hot water in the faculty’s laboratory is similar to the

solar system design for a residential house with a number of three or four inhabitants. For this reason the results from the experimental measurements made in the laboratory can be used to establish the energetic efficiency of buildings. The capability study concluded that the system has the capability of producing home hot water necessary for a residential building. The temperature of the hot water from the solar boiler reached different temperatures depending on the solar radiation, but it should be noticed the fact that the solar system equipment produced hot water with a temperature higher than 55˚ C. In this case, the contribution of thermal energy produced by a heating source with conventional fuel would be zero which would lead to a decrease of the yearly energy necessary and would thus increase the energetic performance to the building. REFERENCES 1. MOCANU, R., UNGUREANU, N. (1982); Strength of materials, Material tests; Cap. Statistic design to

experimental results, Ed Tehnic�, Bucharest. 2. CÎRSTOLOVEAN, L. (2009), Contribu�ii privind realizarea performan�ei prin calitate în concep�ia �i

realizarea instala�iilor pentru construc�ii (Contributions concerning the fulfilment of the performance by quality in the development and achievement of the installations for building). PhD Thesis,Universitatea Tehnic� “Ghe. Asachi“ Ia�i, Ph.D. Advisor Nicolae Ungureanu, Ph.D.Prof.eng.

3. *** http://www.esrl.noaa.gov/gmd/grad/solcalc/, accesed at March-April 2011. 4. *** http://www.fchart.com, accesed at March-April 2011. 5. *** http://squ1.org/wiki/Solar_radiation, accesed at March-April 2011. 6. *** http://-www.termo.utcluj.ro, accesed at March-April 2011. 7. *** http://en.wikipedia.org/wiki/File:Solar_Spectrum.png, accesed at March-April 2011. 8. *** Oventrop Software. 9. *** Vaillant-Solar systems design.

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SHEAR DESIGN PROCEDURE FOR HIGH STRENGTH CONCRETE BEAMS WITHOUT STIRRUPS

CONSTANTINESCU Horia*, M�GUREANU Cornelia,

Technical University of Cluj-Napoca, *e-mail: [email protected] (corresponding adress)

A B S T R A C T This paper presents a comparisson between experimentaly determined shear capacity values of high strength concrete beams without transverse reinforcement and values for shear capacity obtained with provisions of the Model Code 1990, Eurocode 2 and Model Code 2010. The concrete grade used for casting the beams is C80 and the longitudinal reinforcement is of type Bst500S steel.

Keywords: high strength, concrete, beams, design shear capacity

Received: March 2012 Accepted: March 2012 Revised: April 2012 Available online: May 2012

INTRODUCTION Discovery of silica fume and the development of additives and super plasticizers have spurred

great development in the field of reinforced concrete structures, and the development of high strength concrete. Along with increased compressive strength, these concretes also offer a number of other advantages.

Some of these advantages are: - rapid strength increase at early ages, leading to a reduction of the time needed before the

concrete forms can be removed; - reduced deformations caused by creep and shrinkage; - excellent durability in aggressive environments; - increased abrasion resistance; - reduction in the quantity of materials needed due to the reduction of cross sections and

increase in spans; - reduction in tension loss in the case of pre-stressed elements [1], [2], [3], [4]. The different characteristics of high strength concrete requires research into all aspects of its

behaviour. Investigation into the shear behaviour of high strength concrete elements is necessary because

shear leads to sudden failures even in normal strength concrete elements, and more so in elements cast using high strength concrete, in which the resistance of the concrete matrix is close to that of the aggregates used leading to propagation of cracks through aggregate rather than around it [5].

The experimental results which will be presented regard the failure loads of 3 reinforced high strength concrete beams without shear reinforcement tested with 3 different shear span to depth ratios (a/d = 1.00; 1.50 and 2.00).

These results will be compared to design shear resistance values obtained using provisions of the CEB-FIP Model Code 1990 (MC’90) [6], SR EN 1992-1-1 (EC2) [7] and CEB-FIP Model Code 2010 (MC’10) [8] [9] in order to see if the formulas contained within the three design codes give accurate aproximations of the experimental failure loads of the specimens.

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MATERIALS AND METHODS 1. Description of test beams

The concrete used for casting experimental specimens was produced using local aggregates 0-4[mm] – river sand, 4-8[mm] – crushed quarry stone, 8-16[mm] crushed quarry stone. Portland Cem I 52.5 cement was used and fly ash was added. A policarboxilate, water reducing additive was used. The water/binder ratio was 0.26.

Strengths of the produced concrete were established experimentally on test specimens. Tests were conducted in accordance with RILEM testing procedures [10].

The obtained concrete had a mean value of compressive strength fc = 100MPa, determined at the age of 28 days on concrete cubes 150x150x150 mm stored up to that age in water at 20±2ºC. According to [5], [6], [7] and [8], the concrete grade obtained is C80, with a characteristic compressive strength of 80MPa.

High strength concrete beams tested in shear, had a width b = 120mm and a height h = 240mm and with a total length of and were reinforced using deformed rebar with a characteristic yield strength, fyk = 500MPa. The diameter of the longitudinal bars was 20mm. The detailing of the test specimens is shown in (fig.1).

Fig.1. Reinforcement details for beams without transverse reinforcement tested in shear with different a/d ratios

a/d = 2.00

a/d = 1.50

a/d = 1.00

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2. Testing procedure of beams The beams were tested as simply supported loaded by a concentrated load at mid span. Two

tests were performed on each of the 3 beams, one at each end for a total of 6 tests. The load was gradually increased up to failure of the beam. Each load increment represented

approximately 1/10 of the estimated ultimate load. At each stage the force applied was kept constant for approximately 10 minutes in order to allow the stabilisation of strains and cracks, to record values from deformeters and to measure crack widths, positions and heights. The test setup and measuring equipment used is shown in (fig. 2).

Fig.2. Test setup for beams without transverse reinforcement

tested in shear with different a/d ratios

RESULTS AND DISCUSSIONS All of the tested beams failed in shear, the failure load values are given in (table 1), which

also contains the failure shear force modified by the reduction factor � = a/2d prescribed by Eurocode 2 for loads applied at the top of the beam and at a distance 0.5d � a � 2d from the face of the support.

Table 1. Failure loads

Test no. a/d ratio Failure load Pu [kN] � = a/2d Vu = Pu/2*�

[kN] Average Vu

[kN] 1 1.00 632 0.50 158 2 1.00 551 0.50 138 3 1.00 551 0.50 138

145

4 1.50 397 0.75 149 5 1.50 410 0.75 154

152

6 2.00 308 1.00 154 154

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Each of the three codes studied give formulas for determining the bearing capacity in shear of beams without shear reinforcement, these formulas are given below.

MC’90 [6] gives two formulas for elements without shear reinforcement the first one is for beams, equation (1), and it actualy calculates the load at which shear cracking will occur. The second, given in the subchapter which deals with slabs for members with parallel cords and no shear reinforcement, equation (2).

( ) bdfad

V ckcr3/1

3/1

1003

15.0 ρξ��

��

�⋅= (1)

( ) bdfV ckRd

3/11 10012.0 ρξ⋅= (2)

where: d/2001 +=ξ with d in mm; � is the ratio of bonded flexural tensile reinforcement, ( bdAs / ), anchored at the suport; b is the web breadth; a is the distance from major load to support;

( ) 3/13 ad is an empirical expression allowing for the influence of the transverse compresion from the loads and support reaction.

EC2 gives equation (3).

( ) bdfkCV ckcRdcRd3/1

,, 100ρ⋅⋅= (3)

where: dk /2001+= with d in mm; ccRdC γ/18.0, = , for 5.1=cγ , 12.0, =cRdC ;

� is the ratio of bonded flexural tensile reinforcement, ( bdAs / ), anchored at the suport; b is the minimu web breadth in the tension side; And MC’10 gives equation (4).

bdf

kVc

ckcRd γυ ⋅=, (4)

where: υk has different values depending on the level of approximation regarded: level I approximation, appropriate for the conception and design of a new structure:

15.03.11000

200 ≤⋅+

=z

kυ for 0=wρ

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level II approximation, is only applicable to members with a minimum amount of stirrup reinforcement:

0=υk level III approximation, applicable to beams as well as slabs and any amount of shear reinforcement:

( ) ( )zkk

dgx 7.010001300

150014.0

+⋅

+=

ευ for 0=wρ

15.116

48 ≥+

=g

dg dk but for concrete strengths in excess of 70MPa, the aggregate size ( gd )

is recommended to be taken as zero, so 15.1=υk .

ss

EdEdx AE

VzM⋅⋅+

=2

ε represents the longitudinal strain at mid-depth of the member.

Using equations (1) through (4) design values for the shear resistance of the tested beams

were determined and are given in (table 2).

Table 2. Average failure loads compared to calculated shear resistance

Test no.

a/d ratio

Average Exp. [kN]

MC’90 equation (1)

[kN] (Experimental

Calculated)

MC’90 equation (2)

[kN] (Experimental

Calculated)

EC2 [kN]

(Experimental Calculated)

MC’10 level I [kN]

(Experimental Calculated)

MC’10 level III

[kN] (Experiment

al Calculated)

1 2 3

1.00

145 62

(145/62=2.3) 34

(154/34=4.3) 32

(145/32=4.5) 19

(145/19=7.6)

13 (145/13=11.1

) 4

5 1.50 152 54 (152/ 54=2.8)

34 (152/34=4.5)

32 (152/32=4.75)

19 (152/19=8)

14 (152/14=10.1

)

6 2.00 154 49 (154/49=3.14)

34 (154/34=4.5)

32 (154/32=4.8)

19 (154/19=8.1)

15 (154/15=10)

It is of note that equation (1) given in MC’90 which gives the closest approximation of

experimental results and is given in the model code in the chapter reffering to beams in shear is not used in Eurocode 2, which is based on MC’90, but is replaced by equation (2) written as equation (3) which in MC’90 is given for slabs.

In MC’10, the contribution of longitudinal steel to shear resistance disappears completely greatly decreasing the design shear capacity of members without shear reinforcement.

Even though shear failure, especialy in high strength concrete beams, is sudden and violent, having such high safety margins in design values takes away from high strength concrete elements

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one of their main advantages namely the reduction in the quantity of materials needed due to the reduction of cross sections and increase in spans.

This difference between experimental and calculated values of shear resistance in beams must be investigated further in order to ensure that future design codes give more accurate aproximations of experimental failure loads and ensure both safety and financial efficiency of high strength concrete members.

ACKNOWLEDGMENTS

Paper prepared for the Project "Doctoral studies in engineering science in order to develop a society based on knowledge - SIDOC", Contract POSDRU/88/1.5/S/60078, with the aid of a type A research grant, code CNCSIS 1036 “Betoane de înalt� rezisten� �i performan� realizate cu fibre de oel de carbon �i pulbere de cauciuc. Comportarea în zone seismice, medii agresive, solicit�ri dinamice �i de uzur�. Ecologia mediului.” (High strength and performance concrete cast with carbon fibers and rubber powder. Behaviour in seismic arreas, aggressive enviroments, under dynamic loads and abrasion. Environmental ecology) – research funded by CNCSIS 2004-2006. Director: Prof. Dr. Ing. Cornelia M�gureanu.

REFERENCES 1. M�GUREANU Cornelia (2003), Betoane de Înalt� Rezisten�� i Performan�� (High Strength and

Performance Concrete), UT Pres, Cluj Napoca, ISBN: 973-662-013-1. 2. NISTIR (1996), Shear Design of High-Strength Concrete Beams: A Review of the State-of-the-Art,

NISTIR 5870. 3. Heghes B. (2009), Ductility of high strength - high performance concrete, PhD. Thesis, Technical

University of Cluj Napoca, UT Press. 4. NEGRUTIU Camelia (2010), Durability of high strength and high performance concrete, PhD. Thesis,

Technical University of Cluj Napoca, UT Press. 5. Task Group 8.2 (2008), Constitutive modelling of high strength/high performance concrete – State-of-art

report, International Federation for Structural Concrete (fib), Sprint-Digital-Druck, ISBN 978-2-88394-082-6, ISSN 1562-3610, Stuttgart, Germany.

6. Comite Euro-International Du Beton (1993), Ceb-Fip Model Code 1990, Thomas Telford, ISBN 0-7277-1696-4, Great Britain.

7. Asociatia de Standardizare din România: SR EN 1992/2004, Eurocode 2: Design of concrete structures – Part 1-1: General rules and rules for buildings, ASRO, Romania.

8. International Federation for Structural Concrete (2010), Model Code 2010 First Complete Draft, DCC Siegmar Kastl e.K., Germany.

9. Bentz E. (2010), MC2010: Shear strength of beams and implications of the new approaches, Shear and Punching shear in RC and FRC elements, Proceedings of a workshop held on 15-16 October 2010, Salo, Italy, ISSN 1562-3610, ISBN 978-2-88394-097-0.

10. RILEM (1975), Technical Recommendations for the Testing and Use of Construction Materials, E & FN SPOON, ISBN 0419 18810X.

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SUSTAINABLE FIRE SAFETY DESIGN FOR BUILDING FRONTAGES

DÂRMON Ruxandra, Technical University of Cluj-Napoca, e-mail: [email protected]

A B S T R A C T Within the last decade, in Romania, hundreds of buildings, mainly blocks of flats, were insulated with expanded polystyrene. Being a combustible material, expanded polystyrene (EPS) can be considered a hazardous material related to fire safety concern when is used as a cladding system at the exterior face of a wall. The national General Regulations concerning the buildings fire safety recommend that the combustible cladding systems for frontages to be interrupted by non combustible strips, to prevent the upward fire spread to the next floors. The norm doesn’t give any requirements or specification about the materials to be used, the dimensions or the positions for the non combustible strips, therefore this provision is often disregarded or improperly set up. In the article is proposed a calculation method based on numerical simulations for the estimation of the minimum height for the horizontal stripes of mineral wool, build-up over the window openings.

Keywords: expanded polystyrene, mineral wool, fire behaviour, numerical simulation, performance based design


INTRODUCTION

In the context of the harmonized European technical regulations for building design, in Romania were enacted the Eurocodes in order to comply with the new requirements and performance based criteria. The regulation process is developing slowly, for some domains like fire safety engineering, hindered by the lack of specialists.

Due to the huge amount of energy loss in the constructions sector, it was initiated a national program for the thermal rehabilitation of civilian buildings. The storeyed blocks of dwelings took priority over the other structures and within over ten years the expanded polystyrene (EPS) was the prevalent solution for thermal insulation. Although it has good thermal insulation properties, low weight and economicity, EPS, also has some drawbacks concerning the fire behaviour and ignitability, being a combustible material. In accord with the General Norms for Fire Safety, Article 52 [1], in case of using combustible cladding systems for building frontages, these would have to be interrupted by non combustible strips. The mineral wool is the most appropriate material for this purpose, since it is a non combustible material and a good thermal insulator.

Based on the fire reaction tests carried out on cladding systems made with polystyrene (EPS) and with mineral wool, the aim of this research was to determine by numerical simulations, the optimal, satisfactory height of the non combustible strips.

MATERIALS AND METHODS

During this work research were carried out two series of fire reaction tests: The Ignitability, according SR EN ISO 11925-2 [2] and Single Burning Item-SBI, according SR EN 13823 [3] on four cladding systems produced by Saint-Gobain Weber Romania s.r.l., having the following commercial designation and characteristics:

• Therm Weber Family – M1- expanded polystyrene with special adhesive and acrylic plaster; • Therm Weber Family – M2- expanded polystyrene and acrylic plaster; • Weber Therm Prestige – M3- mineral wool and silicate plaster [4]; • Therm Weber Family – M4- expanded polystyrene with special adhesive and silicate plaster.

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The test results from FBI test were modeled with the software Fire Dinamic Symulator, version 5 (FDS 5), which is a free open source computational fluid dynamics model for low speed thermally driven flow. Smokeview is a vizualization program used to display the output of FDS [5].

The study case analised in this article for upward flame spread on the facades of residential buildings it is a block of flats, type T 770 having P+4E.

1. Polystyrene versus mineral wool based cladding systems

The ignitability and the burning properties of the materials used for the exterior wall cladding system have an important role in the fire behaviour of a building frontages. The shape of the fire plume outside a ventilated fire compartment and the upward flame spread is influenced by the total heat flux emerging from the window openings, the thermal properties of the cladding system as thermal inertia and igition temperature and the evironmetal conditions like external temperature, wind speed rate and pressure [6].

Expanded polystyrene (EPS) is a suitable material for the thermal insulation of the exterior walls, having the thermal conductivity around 0,036 [W/mK]. It is easy to carry off and it does not increase much the dead load of a frontage due to its low density ranging between 16 and 640 [kg/m3]. The cladding systems based on expanded polystyrene are inexpensively compared with other materials like mineral wool, slag wadding or fiber glass, therefore the thermal insulation with EPS gained the top of popularity in Romania. Although it has multiple advantages, one of the major drawbacks of expanded polystyrene is its fire behaviour. Under the fire action polystyrene is highly flammable and while burning it produces carbon dioxide and toxic products as polycyclic aromatic hydrocarbons [6] and chlorine. The use of polystyrene for exterior cladding was restricted in many countries like United Kindom, Germany or Canada, being considered a fire hazard. In many cases the coarse execution or the use of low class materials lead to the failing of the plaster which protects the insulating material, as it can be seen in the Figure 1, below.

Fig.1. Flawed plaster of a cladding system with expanded polystyrene

(the picture was taken by the author in Cluj-Napoca, Unirii Street)

In order to decrease the fire hazard for the exterior cladding systems the design norms reccommend the use of non combustible materials. Mineral wool has the highest fire reaction class, namely, A1 and good thermal properties, which make it suitable in case of thermal insulation and fire protection systems for buildings.

2. Fire reaction tests

In order to analyze and compare the fire behavior of thermal systems, two series of fire reaction tests were carried out on three cladding systems based on expanded polystyrene, respectively M1, M2 and M4 and one cladding system based on mineral wool, foregoing denoted

a

b

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M3. All test samples had the same thickness of insulation, both polystyrene and mineral wool, of 10 cm. 2.1. Ignitability. The first series, Ignitability, according SR EN ISO 11925-2 [2] estimates whether the product sample inflames under the exposure of small flame, for 30 seconds. As it can be seen in the Figure 2, the pictures took after the test indicate a good fire behavior for both kind of cladding systems. The acrylic plaster, either silicate plaster didn’t set on fire.

Fig.2. Test samples after the Ignitability fire reaction test

a.The system M3- mineral wool based, b.The system M4 – polystyrene based (source: pictures from the Fire Reaction Test Results from National Fire Test Laboratory, Bucharest)

2.2. Single Burning Item (SBI). The Single Burning Item (SBI), according SR EN 13823 [3] is a fire reaction test which simulates a natural scale object burning in a corner of 3m x 3m x 3m room. The reaction of the samples to a burner located inside the room is monitored instrumentally and visually. In Figure 3 it is shown the SBI test for a system based on expanded polystyrene.

Fig.3. Single Burning Item (SBI) fire reaction test

a.The system M4 - during the SBI test, b.The system M4 – after SBI test (source: pictures from the Fire Reaction Test Results from National Fire Test Laboratory, Bucharest)

3. Numerical simulations using FDS 5 The numerical simulation integrates the test result data from Single Burning Item fire reaction

test approaching a combination between the fire model based on nominal temperature-time curves and computer models based on fluid dynamics (CFD) in order to study of the fire phenomena occuring during the fully developed stage of an enclosure fire. This method is appropriate for current design cases, when no reliable data are available about the chemical composition and the real distribution of the fire load inside a fire compartment. For the current study research, it was modeled a bedroom from a flat, situated at the first etage of a four storey block, type T 770. The

a b

a

b

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thermal load was estimated as the quantity of heat release rate while burning the furniture and fabric materials, based on other studies available in the fire safety literature.[9], [10] Thus, the maximum heat relesase rate (HRR) for the fully developed fire was calculated according the Annex E of Eurocode 1991-1-2 [8], using the following formula:

fif AHRRQ ⋅= (1)

where: fHRR = 250 [KW/m2] for dwellings, according Table E.5 [8], and fiA represent the floor

area of the fire compartment, which is 11,22 m2 for this study case, having the size dimensions shown in Figure 4, below.

Fig.4. The fire compartment floor as it was modeled in FDS

a.The fire compartment where the ignition started, b.The upper floor of the fire compartment

The calculation domain for the numerical simulation was a cuboid, having 5,40 m x 4,00 m x 5,40 m, divided by FDS in 116 640 unit volumes. The initial state condition were the ambient temperature and pressure and the total simulation time was set at 900 seconds. For each time step it was measured the external temperature and the heat loss in 12 points on the frontage, with 12 termocouples and it was calculated the profile of the temperature variation in front of the spandrel between the first and second floor of the building. The frontage was insulated with the system M4, based on expanded polystyrene, using the fire reaction tests data. It was analised the plan temperature field, crossing the center of the window opening.

RESULTS AND DISCUSSIONS

Even though the four insulation systems were classified in the same fire reaction class, B s1 d0, the fire behaviour of the system with mineral wool is much better. The smoke production rate and the fire spread rate are centralized in the graphs below, for the analyzed insulation systems and it can be seen that the mineral wool based system, namely M3 had the slowlier rates for both parameters in discussion.

a b

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Fig.5. Centralized test results from FBI test for the Weber thermal systems

a. Centralized smoke production rates, b. Centralized fire spread rates

Mineral wool is a very suitable material for the thermal insulation of the building frontages, increasing the fire performance. It can be applied for any kind of structure: masonry, concrete or wood and it does not affect the architecture of the frontage. Another noticeable feature of mineral wool insulation, compared with polystyrene is that the former can be also used in case of curved frontages, while the latter is only possible to use on straight surfaces. Even though it has many advantages, mineral wool is not very popular in Romania due to its cost. To increase the sustainability of the building frontages and in order to comply with the General Norms for Fire Safety, it should be adopted a mixed solution for thermal insulation, with expanded polystyrene and mineral wool, as it is shown in Figure 6.b.

Fig.6. Block of flats, type T770

a. before thermal insulation, b. after it was insulated with polystyrene and mineral wool strips

The efficient height of the non combustible strips can be determined for different building types depending of the destinations, the dimensions of the fire compartment and opening factors, considering several fire scenario. In this article it is analyzed for the wide spread block, type T770 having P+4E. Providing non combustible strips over the window openings increases the general fire behaviour of a building and prevents the upward fire spread on the exterior cladding.

CONCLUSIONS

The numerical simulations performed with Fire Dynamic Simulator, version 5, for residential buildings, revealed that the fire plume emerging from a window opening of 1,80 m width and 2,10 m height, is about 2,5...2,8 m long and it does not attach to the spandrel above.

a

b

a b

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In order to improve the passive fire protection of a frontage insulated with expanded polystyrene based cladding system and to comply with General Norms for Fire Safety, it is proposed a solution of a mixed cladding, by providing non combustible strips made with mineral wool on the spandrel between two adjacent floors of the building. The strips must be continous on the frontage and their height has to be at least 80 cm, to be effective. ACKNOWLEDGMENTS

The fire reaction tests have been financed by Weber Saint-Gobain Romania s.r.l. through a research contract with Technical University of Cluj-Napoca and the test results data were made available for this article.

REFERENCES 1. *** “Ordinul nr. 163 din 2007, pentru aprobarea Normelor generale împotriva incendiului”, M.O.

nr.216, 29/03/2007 (2007)- (Decree nr. 163 from 2007 concerning the approval of the General Norms for Fire Safety, published in M.O nr.216 from 29/03/2007).

2. *** SR EN ISO 11925 (2002), Încerc�ri de reac�ie la foc. Aprinzibilitatea produselor pentru construc�ii care vin în contact direct cu flac�ra – Partea a 2 a : Încercare cu o singur� flac�r�, ICS 13.220.50, ASRO. – (EN ISO 11925 (2002), Reaction to fire tests. Ignitability of building products subjected to direct impingement of flame - Part 2: Single-flame source test).

3. ***SR EN 13823 (2004), Încerc�ri de reacie la foc. Produse pentru construcii, cu excepia îmbr�c�mintei de pardoseal�, expuse unui singur obiect arzând, ICS 13.220.50, ASRO – (EN 13823, Reaction to fire tests for building products. Building products excluding floorings exposed to the thermal attack by a single burning item).

4. ***www.casesigradini.ro, http://www.casesigradini.ro/revista/a4/3006/2/Constructii/Sisteme-termoizola nte-pentru-fatadele-viitorului, accessed on 21/03/2012.

5. ***http://www.fire.nist.gov/fds/, accessed on 21/03/2012. 6. DÂRMON, R. (2010), The Fire Spread Outside of a Building, Acta Technica Napocensis – Series Civil

Engineering and Architecture, Vol. 53, pp.152-155, ISSN 1221-5848, Cluj-Napoca, Romania. 7. HAWLEY-FEDDER, R.A., PARSONS, M.L. and KARASEK, F.W. (1984), Products Obtained During

Combustion of Polymers Under Simulated Incinerator Conditions, II Polystyrene, Journal of Chromatography, No. 315, pp. 201–210, Elsevier Science Publishers B.V, Amsterdam, The Netherlands, DOI: 10.1016/S0021-9673(01)90737-X.

8. ***SR EN 1991-1-2 (2004), Eurocod 1: Ac�iuni asupra structurilor – Partea 1: Ac�iuni generale – Ac�iuni asupra structurilor expuse la foc, ICS 13.220.50, ASRO – (SR EN 1991, Eurocode 1: Actions on structures - Part 1-2: General Actions – Actions on structures exposed to fire, Annex E: Fire Load Densities).

9. BWALYA, A.C., LOUGHEED, G.D., KASHEF, A., SABER, H. (2008), Survey Results of Combustible Contents and Floor Areas in Canadian Multi-Family Dwellings, National Research Council Canada, Fire Research Program, Research Report, IRC-RR-253, Ottawa, Ontario, Canada; http://www.nrc-cnrc.gc.ca/obj/irc/doc/pubs/rr/rr253/rr253.pdf, accessed at 23/03/2012.�

10. BWALYA, A.C., GIBBS, E., LOUGHEED, G.D., KASHEF, A., SABER, H.H. (2009), Combustion of non-open-flame resistant Canadian mattresses in a room environment, Fire and Materials 2009: 12th International Conference, pp. 1-12, San Francisco, U.S.A.; http://www.nrc-cnrc.gc.ca/obj/irc/doc/pubs/ nrcc50552/nrcc50552.pdf�, accessed at 23/03/2012.

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25

TOTAL AIR PRESSURE LOSS CALCULATION IN VENTILATION DUCT SYSTEMS USING THE EQUAL FRICTION METHOD

DOMNI�A Florin*, HO�UPAN Anca, POPOVICI Tudor

Technical University of Cluj-Napoca, *e-mail: [email protected] (correponding author)

A B S T R A C T This method may be applied in two situations: when the total air pressure drop for the system is given and when some values for the so-called economical velocity are imposed along the successive ducts. In the first case, the method consists in determining the necessary diameters for some ducts and for all the system. Thus, in this case, the total available pressure is divided to the total length of the main duct, giving as result the friction loss per meter of duct, also known as friction loss factor or specific linear pressure drop [Pa/m], which includes both major and minor losses.

Keywords: air pressure loss, ventilation duct system, airflow rate, equivalent resistance, nomogram.


INTRODUCTION

The presented method gives a simplified approach for determining both the major and minor head losses for air distribution networks. Through the method, these losses are evaluated simultaneously, which allows, on one hand to obtain low energetic consumption, and on the other hand, a fast balancing of the branches in order to attain the designed airflow rates.

At a global level, the entire air transporting network offers the same specific pressure drop, measured in Pa/m and which includes both major and minor losses.

When air is moving along the air pipe network, it can be accepted that the conditions for considering air as an incompressible gas are fulfilled; therefore the perfect gas laws are applicable.

For making the computing easier, a certain equivalence coefficient is introduced by multiplying the exponent 2 of air flow volume, obtaining a result which is proportional to the total head loss. Therefore, this coefficient is a hydraulic similitude factor [1].

By the means of this coefficient, no more iterative calculations will be used for the pressure balancing throughout the system. For applying the method, three diagrams were proposed, combining together the necessary data for dimensioning a ventilation network.

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MATERIALS AND METHODS 1. Determination of total air pressure loss H

The applicability range of the method includes situations which can be reduced to two distinct cases:

-the value of total air pressure loss for the network is given or is imposed; -some particular values of so-called economical velocity are requested for the consecutive

ducts along the network[2]. Whether in the situations corresponding to the first case, the framework of the method

determines the necessary diameter for each duct, the second case situations are solved by determining the diameters and the air pressure loss for each sector and in the entire network.

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Thus, in the first case, the available pressure H (which must overcome the major and minor losses) is divided to the length of the main branch (considered usually as being the longest branch and having the biggest local pressure drops) [1], [3].

�

=�

HR ; [Pa/m] (1)

where: � �means the total duct length of the main branch. On a certain branch, the “n” duct, having ln length, will generate a total pressure loss Hn, as

follows: ;RH nn �⋅= [Pa] (2) The air pressure losses in junctions result from balancing the pressure at the knots. Knowing the air pressure drop for each individual duct, the length of the ducts, the necessary

airflow volume and the sum of local pressure drop coefficients, the design diameter of the duct may be determined [1], [4].

The total air pressure H for the main branch is:

;2v

dH

2⋅ρ⋅��

��

� ξ+⋅λ= ��

[Pa]; (3)

where: λ is the friction coefficient; � – the duct length; d – the equivalent diameter; ξ – the local pressure drop coefficient; ρ – air density; v – average air velocity.

Knowing that the duct has circular section and transports the airflow rate D, from the

continuity equation comes:

;d

D4v 2⋅ρ⋅π

= [m/s]; (4)

Replacing this expression of the velocity in (3) we find:

;DAdd

H 245 ⋅⋅��

��

� ξ+⋅λ= ��

[Pa]; (5)

where:

;16

A 2 ρ⋅π= [m3/kg]; (6)

is a quantity which can be considered as constant for a certain network. The µ coefficient is introduced:

;dd 45�ξ

+⋅λ=µ �[m-4]; (7)

therefore relationship (5) becomes:

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27

;DAH 2⋅µ⋅= [Pa]. (8) To a certain value of the µ coefficient may correspond different combinations of the

quantities � , d and � ξ , whereas for D=ct. �i µ = ct., the air resistance of the network is constant. So, air ducts having different lengths, diameters and local pressure drop coefficients but providing the same value for the µ coefficient, are called similar.

In other words, the µ coefficient is a hydraulic criterion (or dimensionless group) of similarity [5], [6].

By the means of this µ coefficient, it is no longer need to make tedious repetitive calculations in order to balance the pressure in junctions, when dimensioning a ventilation network [5], [6].

2. The equal friction method. Total air pressure loss calculation using the equivalent

resistance When the head loss corresponding to a certain airflow rate is known, the following equations

may be written in the case of a junction where two branches 1 and 2 meet:

;DDD

HH

21

21

�

+==

(9)

where: D is the resulting airflow, from the summation of the branches airflows [1], [7]. Therefore: ;DD 2

22211 ⋅µ=⋅µ (10)

or:

.DD

1

2

2

1

µµ= (11)

The relationship (11) may be written as follows:

;D

DD

1

21

2

21

µµ+µ

=+ (12)

At the considered junction we may write: ;ADA)DD( 2

22221p ⋅⋅µ=+µ (13)

where: µp is the equivalent coefficient of the two ducts connected in parallel. The relationship (13) becomes:

.DD

D

21

22p +

µ=µ (14)

Based on (12) �i (14), it follows:

.111

21p µ+

µ=

µ (15)

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By generalization, for n ducts connected in parallel, the equivalent coefficient µp is:

.11 n

1i ip�

= µ=

µ (16)

The relationship (16) presents an analogy with the connection in parallel of the electrical

resistors. When two consecutive ducts 1 and 2 transporting the same airflow rate are discussed (series

connection), the following equations may be written:

;HHH

DDD

21s

s21

�

+===

(17)

where: Hs is the total head loss for the two ducts and Ds is the airflow rate which passes through the ducts.

Thus: ;DADADA 2211ss ⋅⋅µ+⋅⋅µ=⋅µ ⋅ (18)

becomes: ;21s µ+µ=µ (19) where: µs is the equivalent coefficient of the two ducts connected in series.

By generalization, for n ducts connected in series, the equivalent coefficient µs is:

.n

1iis �

=

µ=µ (20)

The relationship (20) presents an analogy with the connection in series of the electrical

resistors. The total head loss of the entire air distribution network is the main value used for the

calculation of the fan necessary pressure. When the values D, � and �ξ are known for each duct, the equivalent diameters of the

ducts are determined by the means of nomograms [1]. In the relationship (7), for the similarity coefficient, we substitute the friction coefficient λ by

its value given by Prandtl, von Karman and Nikuradse:

;d

lg214,11 ε−=λ

(21)

and the result is:

.dd

dlg2lg214,145

�ξ++ε−=µ (22)

For ventilation ducts made of steel plate, the absolute roughness is 1,0=ε mm, so we obtain:

.dd

dlg14,945

2 �ξ+−=µ (23)

where: d is the equivalent diameter of the section, [m].

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29

For circular pipes, the equivalent diameter is the same with the geometrical diameter, but for rectangular ducts having the sides ratio a/b ≤ 10, the equivalent diameter must be calculated with the following formula:

;)ba(

ba3,1d 8

2

55

+⋅= [m] (24)

The relationship (23) is presented below under a nomogram form (Figure 1), in which three

independent variables are combined: µ, d and �ξ . The mass airflow rate D is added, too. The

variables involved in this graphical representation have the following ranges: - airflow: 100…50.000 kg/h; - head loss: 6…300 mmH2O; - similarity coefficient: 0,01…1000; - length of a pipe: 1…40 m; - local pressure drop coefficient: 0…2,5; - air velocity: 1…40 m/s.

Fig. 1. Total air pressure loss calculation using the equivalent resistance

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3. Example of using the nomogram for air pressure loss calculation Following, we give an example how to use this nomogram. If it is known that a ventilation

duct having a length of 30 m transports 3000 kg/h air with total pressure losses of 70 mmH2O and the local pressure drop coefficient is 1,05, find the equivalent diameter of the duct and air velocity through the duct. Using diagram no. I (Figure 1), for D=3000 kg/h (point A) and the head loss 70 mmH2O (point B), we obtain point C, from which results the equivalent coefficient µ=10. From the same diagram, for � =30 m and D = 3000 kg/h, we obtain point D, which is projected on the vertical axis to the right of the diagram, in point E. In diagram no. II (Figure 1), joining the point E with point F (corresponding to local pressure drop coefficient �ξ = 1,05), we obtain on curve µ = 10,

the G point which gives the ventilation duct diameter, d=240 mm. From diagram no. III (Figure 1), for D=3000 kg/h (point I) and d = 240 mm (point H) results the air velocity in the duct, v=18,5 m/s. CONCLUSIONS

This calculation method allows a more simple and rapid determination of total loss of pressure in a ventilation duct network, just by using a nomogram. Is a very useful tool for the designers of ventilation systems in order to make an accurate calculation of ventilation ducts and total pressure drops along the air route.

The method is easier to use than the classical method of calculation (pressure balancing method) and therefore is recommended for air ducts quick calculations and for making technical and economic estimates of the ventilation systems.

It can be used both to calculate the longest and loading air route and to balance the secondary branches of the ventilation networks. Pressure losses in branches will be determined by imposing the pressure balance in the junctions. Knowing the head loss for each individual duct, the length of the ducts, the necessary air flow volume and choosing the sum of local pressure drop coefficients, the design diameter of the duct may be determined.

The method respects the romanian regulations regarding the designing and the execution of ventilation duct systems [8], [9], therefore it can be easily used. REFERENCES 1. POPOVICI T., DOMNI�A F., HO�UPAN A. (2010), Instala�ii de ventilare �i condi�ionare (Ventilation

and air conditioning systems), Vol. I – Ed. UTPress Cluj-Napoca. 2. ASHRAE HANDBOOK (2008), HVAC Applications. 3. CHRISTEA A., NICULESCU N. (1971), Ventilarea �i condi�ionarea aerului (Ventilation and air

conditioning); Vol. I – Ed. Tehnic� Bucure�ti. 4. DU�� G., COLDA I., STOIENESCU P., ENACHE D., ZGAVAROGEA M., HERA D., DU�� A.

(2002), Manualul de instala�ii; Instala�ii de ventilare �i climatizare (Building Services Handbook; Ventilation and air conditioning systems) – Ed. Artecno Bucure�ti.

5. ETHERIDGE D., SANDBERG M. (1996), Building ventilation. Theory and measurement, Ed.Wiley. 6. NICULESCU N., DU�� G., STOENESCU P., COLDA I. (1983), Instala�ii de ventilare �i climatizare

(Ventilation and air conditioning systems), Ed. Didactic� �i Pedagogic� Bucure�ti 7. GRIMM N.R., ROSALER R.C. (1997), HVAC Systems and Components Handbook – Ed. McGraw-Hill. 8. *** I-5 (2010), Normativ privind proiectarea �i executarea instala�iilor de ventilare �i climatizare

(Regulations regarding the design and the execution of ventilation and air conditioning systems), Indicativ I 5/1.

9. *** STAS 9660 – Instala�ii de ventilare �i climatizare. Canale de aer. Forme �i dimensiuni (Ventilation and air conditioning systems. Air ducts. Dimensions and shapes).

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31

CALCULATION OF KINEMATIC CHARACTERISTICS THROUGHOUT THE RANGE OF CONDITIONS TO MAXIMIZE THE POWER

EXTRACTED FROM WIND IN THE CASE OF SMALL SIZE TURBINE

DUB�U C�lin Gavril, University of Oradea, Faculty of Environmental Protection, e-mail: [email protected]

A B S T R A C T This paper aims at achieving a high degree of reaction, around 0.8, to which the theoretical power coefficient should reach values

over 0.8; by using these guidelines one have analyzed different types of distribution throughout the range of kinematic values, � , kV3, �r and v, resulting in kinematic and geometric details for the conditions to maximize the power extracted from wind; thus one obtained 16 calculation variants which were twofold synthesised.

Keywords: stage of reaction, kinematic values, power coefficients, section calculation


INTRODUCTION

We have studied the optimal conditions of maximizing the power extracted from wind. Thus we have aimed at achieving a high stage of reaction, around 0.8, to which the theoretical power coefficient (8) can reach value above 0.8.

By using these guidelines we have analyzed different types of distribution over the range of kinematic values� ,

�� , �r. which represent the designer’ options resulting from the design strategy

adopted. After solving six equations and nine variables calculation we have obtained the kinematic and geometric details for various conditions to maximize the power extracted from wind.


The following calculation algorithm solves kinematic values along the radius of turbine moving blade. Velocity (speed) in the inlet area entry v1 is constant throughout blade length range, and normal to the tangential path ( 1 = 90°), tangential (conveying) velocity is calculated for each computing section provided by the current coordinate r by the relation ur = uR(r/R). For each calculation section r one computes all velocity triangles in the inlet (1) and outlet (3) areas are calculated, and for asymptotic conditions ( ), by usual trigonometric / circular relations.

In developing this algorithm one have used optimum conditions related to maximizing the power extracted from wind. Thus one aims at achieving a very high degree, around 0.8 - 0.9, to which the theoretical power coefficient reaches a value above 0.8.

H2500 turbine areas were structured by type turbine - �0, v1 - operating velocity of the turbine, uR - tip speed, coupled with the maximum speed required by generator (n = 250 rpm, respectively � = 26.2 rad./s).

Analytical relations presented within this system are valid for a control volume associated with the turbine located in the closest proximity (reduced as axial extension downstream), therefore they are strictly valid for the turbine model itself.

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2

k1k 3

T

VV

+= (1)

2r

2V2

t�

k1)R(1)R(1k 3

−+−+−−= (2)

t2rp kR2k

s⋅⋅⋅= λ∆ (3)

t2rVP kk2C

T⋅⋅⋅= λ (4)

T

a

V

PF k

CC = (5)

r

PM

CC

λ= (6)

The system allows to identify the values associated to a basic calculation section of the

turbine located at a current radius r; setting the overall values for the whole turbine is done by integrating relations across blade (from 0 to r = rmax).

RESULTS AND DISSCUSIONS

Data on designed turbine, with the wind exposed area of 7.5 m2 which were used to set the number of variants examined must meet the following parameters:

� Turbine type �0 = 2...3, � Calculation wind speed 5 m/s and 12 m/s.

Turbine operating velocity field falls within the range v = 2-15 m/s; the maximum speed of power unit is n = 250 rpm, � = 26.2 rad./s tip speed uR � 40 m/s (at maximum).

� ��

� = 0.8…0.9,

� Reaction stage � = 0,8...0,9.

Based on these four parameters encoded as follows x1 = �0, x2 = v1, x3 = ��

� , x4 = � ,

calculation variants were obtained starting from the following configuration:

Table 1. Variants of calculation Variants : x1-x2-x3-x4

= 2 1 x1 � �0 = 3 2 = 5 1

x2 � v1 = 12 2 = 0.8 1 x3 �

��

= 0.9 2 = 0.8 1

x4 � � = 0.9 2

��

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33

This paper has analyzed different types of distribution the kinematic values of � , kV3, �r and v along the radius, resulting in kinematic and geometric details for the conditions to maximize the power extracted from wind; 16 computing variants were obtained and which were embodied in the following syntheses.

Table 2. Synthesis 1a Variant Constant ��

1-1-1-1 �0 = 2 ; v1 = 5 ; ��

� = 0.8 ; � = 0.8 516.123 0.893 0.8

1-1-1-2 �0 = 2 ; v1 = 5 ; ��

� = 0.8 ; � = 0.9 649.733 1.124 0.9

1-1-2-1 �0 = 2 ; v1 = 5 ; ��

� = 0.9 ; � = 0.8 338.480 0.586 0.8

1-1-2-2 �0 = 2 ; v1 = 5 ; ��

� = 0.9 ; � = 0.9 455.586 0.788 0.9

1-2-1-1 �0 = 2 ; v1 = 12 ; ��

� = 0.8 ; � = 0.8 7134.881 0.893 0.8

1-2-1-2 �0 = 2 ; v1 = 12 ; ��

� = 0.8 ; � = 0.9 8981.911 1.124 0.9

1-2-2-1 �0 = 2 ; v1 = 12 ; ��

� = 0.9 ; � = 0.8 4679.151 0.587 0.8

1-2-2-2 �0 = 2 ; v1 = 12 ; ��

� = 0.9 ; � = 0.9 6298.017 0.788 0.9

2-1-1-1 �0 = 3 ; v1 = 5 ; ��

� = 0.8 ; � = 0.8 631.006 1.092 0.8

2-1-1-2 �0 = 3 ; v1 = 5 ; ��

� = 0.8 ; � = 0.9 865.990 1.499 0.9

2-1-2-1 �0 = 3 ; v1 = 5 ; ��

� = 0.9 ; � = 0.8 395.748 0.685 0.8

2-1-2-2 �0 = 3 ; v1 = 5 ; ��

� = 0.9 ; � = 0.9 585.403 1.013 0.9

2-2-1-1 �0 = 3 ; v1 = 12 ; ��

� = 0.8 ; � = 0.8 8723.033 1.092 0.8

2-2-1-2 �0 = 3 ; v1 = 12 ; ��

� = 0.8 ; � = 0.9 11971.444 1.499 0.9

2-2-2-1 �0 = 3 ; v1 = 12 ; ��

� = 0.9 ; � = 0.8 5470.823 0.685 0.8

2-2-2-2 �0 = 3 ; v1 = 12 ; ��

� = 0.9 ; � = 0.9 8092.617 1.013 0.9

Table 2. Synthesis 1b. Power coefficients

Reaction stage � = 0.8 � = 0.9 v1 �0 ��

�

�� 0.8 0.893 1.124 2 0.9 0.586 0.788 0.8 1.092 1.490

5 3

0.9 0.685 1.013 0.8 0.893 1.124 2 0.9 0.587 0.788 0.8 1.092 1.499

12 3

0.9 0.685 1.013 CONCLUSIONS

The syntheses reflect the influence of the series of parameters used in the analysis based on the new calculation model on the achievable power coefficient. Power coefficients analysed are the

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maximum theoretical values. It seems that high reaction stages are very effective in obtaining higher power coefficients. The values represent the maximum possible quantities and they show that the new evaluation method of possibilities is grounded. The calculation was done for two wind speeds in the field of interest of the turbine surveyed; these speeds directly influence the theoretical power that can be extracted from the wind.

REFERENCES 1. BEJ A. (2003), Turbine de vânt (Wind Turbines), Colecia “Energetica” Editura Politehnica Timi�oara,

ISBN 973-625-098-9, 85-90. 2. DUB�U C, (2007), Utilizarea microagregatelor eoliene în componen�a unor sisteme complexe (The use

of small wind turbines in some complex systems), Editura Politehnica Timi�oara, ISSN: 1842-4937, ISBN: 978-973-625-408-6, pp. 235-239.

3. GYULAI F. (2000), Curs de specializare în tehnologii energetice durabile. Modulele: Instala�ii Eoliene �i Agregate Eoliene (Specialization course in sustainable energetic technologies. Modules: Wind Installations and Wind Turbines), 42-45.

4. GYULAI F. (2003), Contributions on horizontal axis wind turbine theory, A V-a Conferin� Internaional� de Ma�ini Hidraulice �i Hidrodinamic�, oct. 2000, Timi�oara, România, pp. 5.

5. GYULAI F. (2003), Vocational Traning in Sustainable Energy-Course Wind Energy, pp. 2-3. 6. GYULAI F., A. BEJ (2000), State of Wind Turbines in the End of 20th Century and Proposals for

Romanian Options, Buletinul �tinific al Universit�ii “Politehnica”, Timi�oara, România, Tom 45(59), 2000-ISSN, 1224-6077, pp. 10-12.

7. GYULAI F.,(1992), Ecological arguments for the Wind Farm Semenic, SDWE Timisoara. 8. HARISON E. (1989), Study on the next generation of large wind turbines, EWEC. 9. ILIE V., ALMA�I L., NEDELCU �. (1984), Utilizarea energiei vântului (The use of the wind energy),

Editura Tehnic� Bucure�ti, pp. 118-120. 10. SPERA D. A. (1994), Wind turbine technology – Fundamental concepts of wind turbine engineering,

ASME PRESS, New York.

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APPLICATION OF X RAY DIFFRACTION (XRD) AND SCANNING ELECTRON MICROSCOPY (SEM) METHODS TO THE PORTLAND

CEMENT HYDRATION PROCESSES

JUMATE Elena*, MANEA Daniela Lucia, Tehnical University of Cluj-Napoca, *e-mail: [email protected] (corresponding adress)

A B S T R A C T This paper presents a study performed on type I Portland cement with respect to the cement hydration processes performed at various time intervals. The methods used concern X-ray diffraction and electronic microscopy applied to define materials and to understand the changes occurring in mineral compounds (alite, belite, celite and brownmillerite) during their modification into hydrated mineral compounds (tobermorite, portlandite and etringite).

Keywords: tobermorite, portlandite, ettringite, mineral component


INTRODUCTION Knowing the chemical composition of the raw materials, of intermediate products and of the

final product represents an aspect of major importance in the fabrication and use of a product with expected specifications. The reactions occurring in the cement hydration process have, consequently, been of high interest for the researchers who studied them, among others, by means of the X ray diffraction (XRD) and scanning electron microscopy (SEM) methods.

The use of the mentioned methods allows more accurate information regarding the behaviour of the Portland cement paste during hydration, and a more realistic knowledge of the mechanisms that generate new properties such as strength and durability, which are among the most important in the selection of cement for a specific application.

The Portland cement represents a mixture of clinker and finely ground gypsum, where the clinker is made up of four main mineral components [1] (Table 1), at a maximum temperature of up to 1450°C. In the clinker, the calcium silicates represent 75 - 80 %, hence the name of silicatic cements, while calcium aluminates and calcium aluminoferrite form only 20 – 25 % [2].

Table 1. Main mineral components in Portland cement

Name of the mineral component

Chemical name Oxidic composition Abbreviated formula

Alite Tricalcium silicate 3 CaO • SiO2 C3S Belite Dicalcium silicate 2 CaO • SiO2 C2S

Celite I Tricalcium aluminate 3 CaO • Al2O3 C3A Celite II or Brownmillerite Tetracalcium aluminoferrite 4 CaO • Al2O3 • Fe2O3 C4AF

The Portland cement mixed with water forms a plastic paste or slurry that stiffens as time

goes on and then hardens into a resistant stone. When water is present, the mineral compounds undergo hydration and hydrolysis reactions, which are followed by the appearance of new hydrates. The hydro derivatives exhibit a colloidal structure, which, in time, is concentrated as a gel, and makes the cement paste more and more consistent. The gels in question and the mineral

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components hydro derivatives respectively, start losing hydration water and crystallise. As a result, the cement paste turns into a rigid body with very high strengths.

The cement hydration products are poorly soluble in water [3], as practically the hydration of cement particles is never completed, whatever the precipitation system of the hydration products as shown by the hardened cement stability in water. The hydration speed gradually diminishes. It was found out that, at 28 days of contact with water, the cement grains were hydrated only in a depth of 4µm, and after one year, the hydration reached a depth of 8µm [3, 4].

Alite (C3S) is the mineral component to be found in cement in the largest ratio (50%), under the form of colourless and equisized grains [3]. This is the calcium silicate with the highest hydrolysis reaction which very easily reacts to water. C3S hydration defines to a large extent the behaviour of the cement, though its rate is not constant and not even the rate changes are constant. C3S rapidly hydrates and hardens the cement slurry and enforces high initial (1-3 days) and final mechanical strengths [5].

Belite (C2S) is the mineral component that exhibits three or even four polymorphous forms [3], easily reacts with water and turns into a hydrated dicalcium silicate. This slowly hydrates and hardens the cement slurry and improves the cement mechanical strengths after 7 days. After 28 days, this mineral compound hardens and its mechanical strength will be very close to that of the calcium silicate hydrate originating in C3S [5].

The hydration and hydrolysis reactions of the two mineral compounds above also produce hydrosilicates that initially have a gel structure similar to that of the natural mineral called tobermorite. The calcium silicate hydrates form the majority of the hydration products, present a gel structure, where the solid phase is made up of a lattice of microcrystals, initially of angstrom size with eyes filled with a saturated composition of components: in a later stage, the crystals develop, age and strengthen, leading to the increase of the mechanical strengths [2, 4, 5].

Celite (C3A) is the mineral component that has the form of crystals in the clinker when slowly cooled down or as vitreous mass, when the clinker is cooled down fast. In this case it fills in the voids between the alite and belite crystals. Pure celite has a violent reaction with water; the slurry hardens instantaneously, which requires the addition of gypsum (CaSO42H2O) when the cement clinker is ground [3]. The amount of gypsum added to the cement clinker should be carefully controlled as a too large an amount leads to expansion and hence, to the damage of the hardened cement paste. The optimal gypsum amount shall be defined by observing the hydration heat release. The gypsum also reacts with brownmillerite (C4AF), forming calcium sulphoferrite hydrate and calcium sulphoaluminate hydrate, whose presence can accelerate the hydration of calcium silicates. The two calcium aluminate hydrates act as flux and diminish the clinker burning temperature [3, 5].

MATERIALS AND METHODS 1. State-of-the-art methods used to define cement hydration 1.1. X Ray Diffraction (XRD)

Diffraction is a physical phenomenon that consists in electromagnetic waves avoiding obstacles if the size of the obstacles compares to the wavelength. This phenomenon can be applied to the analysis of materials as the atom plans are placed at comparable distances to X ray lengths. X rays are electromagnetic waves similar to light, but whose wavelength is much shorter (� = 0,2 - 200 Å ).

XRD is produced as a reflexion at well defined angles. Every crystalline phase has its own diffraction image. The diffraction image contains a small number of maximum points that is not all

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the families of crystallographic planes give maximum diffraction points; all the crystalline phases with the same type of elementary cell will exhibit the same succession of Miller indices for the crystalline planes families giving a diffraction maximum points. For the XRD analysis we use diffraction devices (diffractometers), mainly according to the Bragg–Brentano system (Figure 1) (the sample rotates at a diffraction angle ”θ”, while the detector rotates at the angle ”2θ”. In Figure 2, the X ray diffractometer (Bruker) is shown.

Fig.1. The basic layout of an X ray diffractometer [5] Fig.2. X ray diffractometer (Bruker) [5] The diffractogram is made up of a succession of diffraction maximum points, showing the

intensity of the diffracted X radiation on the ordinate measured in pulses/second, and the angle ”2θ” on the abscissa, where ”θ” is the Bragg angle, measured in degrees. The diffraction image depends upon the material structure.

The diffraction methods allow for the performance of the following studies: the determination of the crystalline structures, the phase quantitative and qualitative analysis, the study of phase transformations, the study of the crystallographic texture, the size of the crystallites, the internal stresses in the sample, etc.

The identification of the crystalline phases can be carried out with the X ray diffraction method if the respective phase represents more than 3 - 4% mass. The identification can be made by calculation with Bragg’s relationship or computer-based, using Match, XpertScore software, in the PDF (Powder Diffraction File) database, where identification files for about 200,000 metal crystalline phases, alloys, oxides, salts, etc. are found [5, 6, 7].

1.2. Scanning electron microscopy (SEM)

SEM represents a high performance method used to investigate the structure of the materials. It is defined by: easiness to prepare samples to be tested, large diversity of information reached, good resolution associated with high field depth, large and continuous range of magnifying, etc. The examination of microstructures with SEM offers two benefits as compared to optical microscopy (OM): much more resolution and magnification, as well as very large field depth giving the impression that images obtained are outstanding. Thus, the field depth in OM when magnified 1200 times is 0,08 µm, while in SEM at 10.000 times magnifying, the field depth is 10 µm.

The scanning electron microscope SEM of type Jeol 5600 LV (Figure 3) presents the specifications:

- resolution 3,5 nm (35 Angstroms), with secondary electrons;

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- 300,000 times magnification; - the examination of nonconductive samples (ceramic, biological, medical, etc.) can be

made in reduced vacuum (up to 130 Pa) with backscattered electrons [8]; local quantitative chemical analyses can be made, based on the characteristic spectrum of X rays for the elements contained between Boron and Uranium, at a detection limit of 0,01 %.

Fig.3. The scanning electron microscope SEM of type Jeol 5600 LV

4. The present stage of the Portland cements hydration processes studied with XRD and SEM

The evolution of the Portland cement in the hydration process was investigated by Manuel A.M. Giraldo, Jorge I. Tobon, and other researchers as well, with the help of the materials definition approaches that use XRD and SEM. They identified the modifications that occur in the mineral compounds (alite, belite, celite I and brownmillerite) during the hydration processes, wherefrom calcium silicate hydrates and calcium aluminate hydrates appear (tobermorite, portlandite and ettringite).

2.1. X Ray Diffraction (XRD)

The conclusions of the studies carried out by the researchers mentioned above pointed out the following results, with reference to the hydration processes investigated through X ray diffraction:

After three days, the largest peaks of the of the diffractograms correspond to the tobermorite gels, the second peak corresponds to portlandite, while ettringite exhibits the smallest values. The most abundant phase at three days age is that of tobermorite.

After seven days, the model resembles to the model after three days. The gels of tobermorite and the gels of portlandite have higher values that the ettringite phase.

After 28 days, tobermorite forms a mass which is more dense, more compact and continuous, but where still non-hydrated belite grains can be met and ettringite is difficult to recognise.

As time goes on, in the hydration process there occurs a diminution of values between the tobermorite and portlandite phase, between days three and seven, as well as between days seven and 28. This development can be explained by the fact that during the first three days of hydration tobermorites originating in the alite are more predominant, alite is present in larger amounts than belite which is the source of portlandite that is developed more slowly. As time goes on, a larger amount of portlandite is produced, so that at seven days, the difference between tobermorite and portlandite diminishes. After 28 days, the difference between the two compounds increases as

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tobermorite develops very much. While hydration continues, ettringite becomes more and more evident [9].

2.2. Scanning electron microscopy (SEM)

In the case of research performed with SEM, we present the conclusions of the studies performed by the researchers mentioned above, as follows.

After three days, while alite is hydrated, tobermorite develops to a higher extent and forms a continuous matrix of layers joined together. Portlandite in its hexagonal shape is also identified. Grains of still non-hydrated alite and belite can also be found. Ettringite starts appearing in small amounts. Due to hydration heat, cracks visible at the scanning electron microscope can also be seen.

After seven days, the hydration process has occurred to a higher degree, tobermorite forms and goes on developing the continuous matrix which had appeared at the age of three days. Portlandite can be identified in its characteristic hexagonal shape and the needle shaped ettringite can be seen quite well.

After 28 days, tobermorite forms a mass that exhibits more density, more compactness and continuity and where belite grains that have not yet hydrated can be identified. 5. Personal contributions regarding the study of the hydration processes of the Portland

cement by means of XRD and SEM The Portland cement slurry was prepared in the laboratory of the Faculty of Civil Engineering

of Cluj-Napoca. Tests were performed on the slurry, according to SR EN 196-3/A1:2009, the final formula being: 300g Portland cement and 96ml water. Cakes were made from the normal consistency slurry and were kept in relative humidity of 55 % at a temperature of 20°C, for 28 days. Samples were taken from the cakes at 1, 3, 7, 14 and 28 days, to be examined by means of the two methods mentioned that is XRD and SEM.

RESULTS 1. X Ray Diffraction (XRD)

The analyses made by XRD used a DRON 3 diffractometer, with an angular range of 2θ = 10 – 70 degrees, at a radiation of λ=1,54182 Å, voltage of 25 kV and intensity of 25 mA, in a Bragg – Brentano scheme. The diffraction samples were either powder (found by pestle milling) or cakes.

The diffraction spectra show that alite is found mostly in the sample, that part of it is hydrated and the calcium silicate hydrate is produced. Ettringite and portlandite are present in all hydration stages. Changes in the mineral compounds during hydration were highlighted where from silicate hydrates and aluminate hydrates appeared (tobermorite, portlandite and ettringite), Figure 4.

After three days, the highest peaks in the diffratograms corresponded to alite and tobermorite gels. After seven days, the spectrum is similar to the spectrum after three days. During the hydration process, values between tobermorite and portlandite phases are seen to diminish both between three and seven days and seven and 28 days.

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Fig.4. X-rays pattern of Portland cement (cs-alite; csh- hydrated calcium silicate (tobermorite); p-portlandite; e-ettringite, b-belite; c-celite; bw-browmillerit) [5]

The phenomenon can be explained by the fact that in the first three days of hydration

tobermorites originating from alite are predominant, that alite is present in larger quantity than belite which is the source of portlandite that, in turn, is formed more slowly. More portlandite is produced in time, so that at seven days, the difference between tobermorite and portlandite reduces. At 28 days, the difference between the two compounds increases as tobermorite develops very much [5].

2. Scanning electron microscopy (SEM)

To analyse the morphological evolution of the Portland cement samples at various time intervals (after 3, 7 and 28 days) we used the scanning electron microscope SEM of type JEOL JSM5600LV, equipped with a EDX Oxford instruments spectrometer, INCA 200 Soft. As the microscopic analysis requires samples that are electrically conductive, all the samples were covered by a layer of gold, deposited by spraying. The samples were then broken and the analysis concerned the breaking surface. After three days, while alite is hydrated, tobermorite develops more and more, forming a continuous matrix in the form of layers joined together (Figure 5). Portlandite is also identified, in a hexagonal shape. Alite and belite grains not hydrated yet are still present. A more magnified image shows varied areas in the sample. These areas correspond to the various constituents of the cement, which appear after three days following the mixing of the cement. Thus, one of the constituents in these samples is ettringite (Figure 6), platelike structure is visible almost everywhere.

Fig.5. The tobermorite at 3 days Fig.6. The ettringite at 3 days

If the cement slurry is kept up to seven days, its structure will be more modified, so that its surface, under a small magnification, shows to be more compact (Figure 7). On the other hand,

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cracks were found (Figure 7) as a result of the hydration heat, produced by the exothermal hydration reactions. After seven days, the slurry exhibits a high degree of hydration, tobermorite forms and develops its continuous matrix evident after the third day. Portlandite is also identifiable in its characteristic hexagonal shape and ettringite in the forms of needles can also be seen at this age (Figure 8).

Fig.7. The morphology of cement surfaces at 7 days

Fig.8. The ettringite at 7 days

After 28 days, tobermorite forms a more dense, compact and continuous mass, with identifiable though still non-hydrated belite grains (Figure 9).

Fig.9. The morphology of cement surfaces at 28 days

In the case of portlandite, no notable change was found after seven days. At this age, the crystals are well developed, with well-defined sides and the characteristic hexagonal plates. Filaments of ettringite can also be seen among the particles (Figure 10).

Fig.10. The ettringite at 28 days

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CONCLUSIONS By means of the two methods DRX and SEM, the presence of hydration compounds as well

as of mineral compounds that are to hydrate in time was highlighted. The interpretation of the results with the DRX enables a more precise identification of the mineral phases, in all samples under investigation. Alite, belite and brownmillerite were well maintained after 28 days inclusively, while the calcium aluminate was not identified after the first day. There are three main hydrated compounds: tobermorite, ettringite and portlandite. The phases occur during the test. The study of cement hydration by means of SEM identifies the modifications that appear in the cement mineral compounds as follows. After 3 days the most abundant phase is that of tobermorite. Portlandite and ettringite are also present. After 7 days, the tobermorite gels and the portlandite gels present a higher value that that of ettringite. After 28 days, tobermite forms a much more compact and continuous mass where still non-hydrated belite grains can be found, while ettringite is difficult to recognise.

In the course of the time, during the hydration process, the values between the tobermorite and portlandite phase diminished. After 28 days, the difference between the two compounds becomes larger as tobermorite increases very much. Both methods applied in the cements hydration processes highlighted the same order for the appearance of the hydration compounds, namely fist appears tobermorite, followed by portlandite and ettringite. The advantage of using SEM lies in the possibility of pointing out other aspects, such as cracks, while the advantage of using DRX lies in the possibility of showing other hydration compounds that the ones mentioned. However, the cement hydration process is never fully completed. REFERENCES 1. IONESCU, I. and ISPAS, T. (2008), Propriet��ile �i tehnologia betoanelor (Concrete properties and

technology), Editura Tehnic�, Bucure�ti. 2. SERBAN, L. (1998), Materiale de construc�ii (Building materials), Matrix, Bucure�ti. 3. NEVILLE, A.M. (2003), Propriet��ile betonului (Concrete properties), Editura Tehnic�, Bucure�ti. 4. MOLNAR, L., MANEA, D. and ACIU, C. (2010), The study of hydration processes of cement based on

latest generation methods, Proceedings of the Internaional Scientific Conference, CIBv 2010, Vol. 1, (November 2010), pp. 215-221, ISSN 1843-6617, Transilvania University Press.

5. JUMATE, E. and MANEA, D.L. (2011), X-Ray difraction study of hydration processes in the Portland cement, Journal Of Applied Engineering Sciences, Vol. 1(14), Issue 1, pp.79-86.

6. GHEORGHIE�, C. (1990), Controlul structurii fine a metalelor cu radia�ii X (Control the fine structure of metals with radiation X), Ed.Tehnic� Bucure�ti.

7. ARGHIR, G. (1993), Caracterizarea cristalografic� a metalelor �i aliajelor prin difrac�ie cu raze X (Crystallographic characterization of metals and alloys by X-ray diffraction), Editura U T Press, Cluj Napoca.

8. VIDA-SIMITI, JUMATE, N., CHICINAS, I. and BATIN, G. (2004), Applications of scanning electron microscopy (SEM) in nanotechnology and nanoscience, Rom. Journ. Phys., Vol. 49, Nos. 9–10, , Bucharest, pp. 955–965.

9. GIRALDO, M.A. and TOBON, J.I. (2006), Evolucion mineralogica del cemento Portland durante el proceso de hidratacion (Mineralogical evolution of Portland cement during hydration processes), Medelin, Colombia.

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PROBLEMS CONCERNING IN SITU BEHAVIOUR OF COMPLEX STRUCTURES

KOPENETZ Ludovic Gheorghe, C�T�RIG Alexandru*, LI�MAN Drago� Florin, Technical University of Cluj-Napoca, *e-mail: [email protected] (corresponding address)

A B S T R A C T Complex structures have two main common characteristics; they have both large horizontal and/or vertical spans. Special care has to be taken by structural designers in order to satisfy the strength and stability criteria for static and especially dynamic actions. The in situ structural behavioural parameters are usually: displacements, deformations, velocities and accelerations. The accuracy of results concerning the assessment of the safety of complex structures depends on the degree of approximation which exists in the theoretical model used during assessment. The selection of the mathematical model for approximation and its pertaining structural schematization along with the interpretation of results constitute the most difficult phases in the process of confirmation of the actual, real behaviour of these structures. In the paper, the authors present a set of solutions concerning the assessment of in situ behaviour of structures, using advanced investigation methods and concepts of system theory.

Keywords: in situ, behaviour, assessment, system theory Received: March 2012 Accepted: March 2012 Revised: April 2012 Available online: May 2012

INTRODUCTION Complex structures generally have large horizontal and/or vertical spans. The tendency to use

these high rise constructions types, with large and very large spans, having lateral surfaces solved using different solutions (flat, curved, twisted, etc.), can be considered one of the main features and trends in today's development of the structural science and technique.

The practical application of these trends is implemented along with the development of new structural concepts combined with innovative structural materials and with new specific advanced technologies.

Recently developed complex structures are characterized by the use of high strength materials and slender, if not bold, constructive solutions.

In many situations these structures withstand large displacements, which have implications both on the structural geometry and on the structural safety. These implications can pass unnoticed during the design phase [1].

For the new structural tendencies in the field of complex structures, the definition given by the architect Ada Luise Huxtable from U.S.A stands perfectly, according to her: a modern construction is that construction, which could not have been developed in the centuries preceding our century. There exist numerous examples of twisted buildings like: The Turning Torso from Malmo (Sweden), Chicago Spire from Chicago (U.S.A), dancing buildings like: West Bay Lagoon Plaza from Doha (Qatar), super skyscrapers with more than 150 stories like: Burj Khalifa from Dubai (U.A.E.), one storey buildings with spans over 200 meters (stadiums from Singapore, Fukui, Namihaya, Nara Hall, etc.)

The paper presents a series of problems concerning in situ behaviour of these complex constructions both from the point of view of the aspects that derive from the structural composition concepts and from the point of view of the design codes [2].

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MATERIALS AND METHODS 1. New concepts that have an influence on in situ behaviour of complex structures

It has to be taken into account that any construction with impressive dimensions has an extraordinary impact on the environment and on natural resources. As a natural necessity, it appears the need to highlight and encompass the aspects of desired in situ behaviour into the building's design phase.

Every complex structure has to satisfy the following general criteria: • the building must be safe and must have sufficient constructive strength; • the building must be durable; • the building must be aesthetic and economic; • the building must correspond to the designed functionality.

In addition to these classic basic mandatory criteria, new structural robustness criteria are

imposed, like: • The avoidance of progressive or chain collapse caused by certain local or global

structural creeping, due to seismic action, severe wind action or due to the effects of shockwaves produced by terrorist attack explosions;

• Preservation of structural integrity and the adaptability to extremely severe actions. As a result the following new aspects arise as requirements for the designers in order to

assure the safety and the sustainability of complex structures: • quick construction time with minimum consumption of structural material; • the distribution of structural elements must not cause functional impediments; • the assurance of adequate natural lighting; • reduced maintenance costs; • the use of ecologic cements; • the use of natural or artificial fibre reinforced concretes; • construction techniques that use the existing in situ soil; • the usage of aerogel-type thermal insulators; • the extended use of lightweight mixed structures composed of steel-concrete,

concrete-wood, etc.; • the placement of cristal unit electric systems for energy generation by using the effect

of pedestrian traffic on the bottom coverings; • mass scale use of solar and photovoltaic panels.

It can be observed that sustainability requirements for complex structures are tightly

connected to structural composition criteria, safety, robustness, economic criteria, problems of environmental impact (natural, social or economic environments) [3].

In order to ensure the sustainability of complex structures, the designers and developers must strive to significantly reduce the quantities of consumed structural materials, nevertheless assuring the above safety criteria.

According to the above, the authors state the following principle: it is always recommended to use statically indeterminate structures.

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Moreover, the promotion of lightweight complex structures is desirable if it correctly solves the problems connected to the building's lateral stiffness requirements. For example, we consider a general economic truth the fact that frame trusses are more economic than Vierendeel trusses. Nevertheless, there are situations when from the manufacturing or from the functional point of view, the Vierendeel trusses are more advantageous and better serve the purpose than regular frame trusses do [4].

2. Design and composition aspects of complex structures related to adequate in situ behaviour

In the case of complex constructions, the bearing structure's design is tightly connected to the chosen constructive solution.

In essence, complex structures are space systems having different structural forms like: bidirectional rigid frames, bidirectional wind-braced trusses, cores and peripheral pillars, tube, tube inside tube and multiple tubes.

The bearing structure is chosen taking into account the function of the building, the shape and the dimensions of the building, the occupancy degree of the available terrain, the climatic and seismic zone, geotechnical conditions, financing possibilities, structural materials, construction technology, height over building height ratio, envelope type, heating systems, electrical energy systems, air conditioning systems, water supply systems, sewage systems, elevators, signalling systems, etc. [4], [5].

The design and the structural composition has to allow the safe appropriate overtake of permanent actions, variable actions and especially actions that create dynamic stresses like severe wind or seismic actions. In addition, it must create a stabile behaviour for exceptional actions that appear as a result of explosions, fire, aviation induced shockwaves, human error, terrorist attacks, etc. [6].

As a result it is of utmost importance to adopt a structural composition that is able to prevent the chain collapse of high profile buildings, even if some structural elements of parts of the building are accidentally destroyed or damaged.

The correct structural composition of a complex bearing structure involves an appropriate fixation of the ensemble of composing elements with respect to the foundation media (existing foundation media or another stabile construction). Thus, the interconnected elements must form a geometrically fixed system, an invariable geometric system in concordance with the hypothesis of Euclid's model.

The structural engineer swill avoid the cinematic chain character or the mechanism character of the structural ensemble by correctly placing the connections, both between the interior elements and the connection between the elements and the foundation basis (exterior connections or supporting connections). Abiding the laws of nature is a prerequisite even in the design phase of structures, given that the nature strives to achieve equilibrium state with the smallest investment of structural material consumption [7].

Professor Eduardo Torroja, in “Philosophy of Structures”, phrases this requirement as follows: we must strive for minimum stress on substance. Design of structures is more like science and technology, it has a lot in common with art, realist-minded, with sense and intuition, the gifted, the joy of creating broadly, creation to which the scientific calculation contributes as a finishing touch, certifying the structure's health and the fact that it corresponds to the functionality.

This finding is entirely true, if we think of how many structures are rated in the stage of imminent collapse in terms of static analysis, nevertheless are still standing because they have an appropriate structure and thus the structural elements help one another.

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As shown, the proper structure design and composition is crucial, structural analysis, static and dynamic and the process of sizing have a subordinate role. From the in situ inappropriate behaviour analysis of large complex structures appears an unfortunate finding that, missing out on this vision today, many architects and engineers are trying to create something new without taking into account the conditions for composition in conjunction with sustainability criteria.

The great engineer-architect P.L. Nerves, in “L'architecture d'aujoud'hui XII-1961” makes the remark that the abundance of ways of structural solutions must not lead to unnatural static models, the structures arising from non-personal laws of statics and do exhibitionist acrobatics with forces. This is now, in his opinion, the greatest danger of engineering construction. In this context, to achieve stabile behaviour of complex structures in situ, that is in the line of sustainable development, engineering work at the highest level is of several options, the best alternative. This operation cannot be done automatically, even though currently there are many expert systems programs, because of the too many involved parameters.

In the consistency and design of today's complex structures, modern architects adopt various structural concepts, with many elements from the world of biostructures. Thus, the architect - engineer Santiago Calatrava, for the project of the "Turning Torso" building in Malmo, used nine blocks (each having six floors) that twist from the bottom to the top 90 degrees, like the human spine. The project “Chicago Spire” building, Calatrava imagined a spiral form, taken from a rattlesnake or form a twisted tree. In this case each of the 150 floors rotates about two degrees, enforcing to the top floor of a turn of 360 degrees.

Another example of complex structures in the landscape of tall buildings are made in zigzag by architect - mathematician Zaha Hadid, the West Bay Lagoon Guard. These buildings, with a height of 143 m, were able to create a sensation of dance structure, thus making the concept of choreographic fluidity in structural architecture. This way, a new class of complex buildings was created, called “dancing towers” (Dancing Towers). The “Absolute Tower” (1 and 2) buildings, due to the adopted shape were called "the couple that is complete" or “Yin and Yang”, the South Tower portraying a woman and the North Tower in the role of a powerful men.

A new category of complex buildings is called "Green Structures" or “Ecological”, designed to be rotated by the wind's action. Rotation may be required for each level or group level. In this way appears, for complex structures, the dynamic architecture concept, in addition to structural architecture already well crystallized [8]. By analyzing in situ behaviour of these complex structures, designed and made correctly, it appears that these meet, in many respects, the proposed sustainable development requirements.

3. Structural changes that characterize in situ behaviour of complex structures

The importance of in situ behaviour concepts of complex buildings as well as structural robustness issues came to the attention of engineers after a series of events caused by terrorist attacks and after the manifestation of extreme actions (tsunami, fire, severe winds, etc.).

Observation of structural issues, in the case of existing complex buildings, appears as an important need, given the fact that the vast majority of them do not provide the necessary structural robustness requirements. The structural robustness of an existing complex construction can be based on the analysis of accident scenarios and taking into account the vulnerability and the actual degree of degradation or damage observed in situ.

In the event of major deficiencies in structural robustness, whose removal requires great costs with suspending the operation of the construction, a continuous or periodic monitoring is recommended.

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The concept of monitoring complex structures differs fundamentally from the current special concepts of monitoring highlighted by national code P130-99 and similar codes seen abroad. Monitoring should be able to capture the following aspects:

• The emergence of structural discontinuities (holes, open or partially open joints, even constructive damage to joints in certain propagation conditions can lead to catastrophic failure or even the collapse of the structure);

• Breaking, the separation of structural material of current applications; • The appearance of cracks and crack propagation with a certain speed and within a time

period. Here the observation that the crack propagation speed depends on the internal defects

(flaws, inhomogeneous inclusions of various materials, etc.) and the existence of macro-cracks (fractures, holes, joints, etc.). The observation of these phenomena can be microscopic or macroscopic.

The monitoring concept will follow, besides acquiring with proper safety the quantities to be monitored, the peculiarity derived from the structural material used. The choice of the model, the properties that are monitored as well as the interpretation of results are the most difficult phases of structural monitoring. CONCLUSIONS

Complex structures, characterized by large dimensions (spans, heights), require special attention from designers, reviewers and experts, in order to satisfy the requirements of strength and stability in situ, especially static and dynamic actions.

Current trends for complex structures falling within the concept “The sky is the limit”, launched in 1900, after the first modern skyscraper (Home Insurance Building in Chicago) requires a permanent activity of in situ performance monitoring of these structures.

The complexity of in situ behaviour for these buildings appears as a consequence of both the size and nonlinear behaviour (geometric, physical).

Specifying the issues that must be pursued in situ was made by considering the notions of interaction between structure and concepts of sustainability or sustainable development. REFERENCES 1. KOPENETZ, L.G., KOLLO, G. (2004), Problems of Applied Statics, International Conference of Civil

Engineering and Architecture, �umuleu-Ciuc, Romania, pp.243-246. 2. KOPENETZ, G. and IONESCU, A. (1985), Lightweight Roof for Dwellings, IAHS, International Journal

for Housing and its Application, Florida, U.S.A, volume 9, no 3, pp. 213-220. 3. KOPENETZ, L., C�T�RIG, A. (2006), Teoria structurilor u�oare cu cabluri �i membrane (Light

structures with cables and shell theory). Editura UT Press, Cluj-Napoca, Romania. 4. C�T�RIG, A., KOPENETZ, L., HODI�AN, T. (2002), Immaterial Structures in Real Structures.

Proceedings of the XXX IAHS World Congress on Housing, Coimbra (Portugal), Vol.1, pp.187-192.

5. C�T�RIG, A., KOPENETZ, L., ALEXA, P. (1996), Rehabilitation on Structures via Membranes. Proceedings of the Eleventh World Conference on Earthquake Engineering, Acapulco, Mexico, pp.1201- 1208.

6. C�T�RIG, A., KOPENETZ, L., ALEXA, P. (1996), Analysis problems of tubular offshore structures, Proceedings 7th International Symposium on Tubular Structures, Miskolc, Hungary, A.A.Balkema Rotterdam Brookfield, pp. 415-420.

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7. LEE, L.T., COLLINS, J.D. (1977), Engineering Risk Management for Structures, Journal of the Structural Division, ASCE 103, No.ST9, September 1977, pp.1739-1756.

8. KOPENETZ, L.G., PRADA, M.F. (2011), Introducere în teoria structurilor speciale (Introduction in special structures theory) , Editura Universit�ii din Oradea, Romania.

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REALTIME BEHAVIORAL MONITORING OF CABLE TRANSPORT STRUCTURES

KOPENETZ Ludovic Gheorghe, LI�MAN Drago� Florin*,

Technical University of Cluj-Napoca, *e-mail: [email protected] (corresponding address)

A B S T R A C T The paper presents two techniques for in situ monitoring of the status of transport structures based and sustained by bearing cables. These transport structures are met both in civil structures and at structures pertaining to mining, military and marine applications. The first technique presented is based on building a 3D cable model by using images acquired with the help of several digital cameras. The 3D model is inspected for damages and degradations by using a database of possible cable degradations and damages. Decisions concerning the existence and degree of degradations made by using fuzzy-logic based reasoning. The system performs detections with accuracies over 70 percent and the database is updated by adding new degradation models after each cable monitoring. The reader will be presented in the second part of the paper with a proposal for a new monitoring technique based on sonic scanning. This technique functions by capturing the sound frequencies produced by the rupture of wires or wires strands during partial structural failure. Work on the implementation of this method is still in progress and no final results can be presented.

Keywords: 3D, sonic scanning, fuzzy-logic, defects, degradations, detection Received: March 2012 Accepted: March 2012 Revised: April 2012 Available online: May 2012

INTRODUCTION Bearing cable transport structures are a category of flexible structures which sustain and

transfer predominantly stretching forces. This is possible by using the strength of the material through intermolecular cohesion. These categories of transport structures based on bearing cables can be found both in civil and military, mining and marine applications. The cables are manufactured using high quality steel and their use is, in principle, unlimited taking into account that their profitability increases with the increase of the spanning distances. Due to the fact that these cables are subject to considerable efforts and also withstand the diverse actions of natural elements, they have to be subject to careful monitoring and any degradation or defect must be identified as soon as it may develop or even predicted, if possible. The image scanning of the outer surface can be successfully performed using the 3D imagining technique presented by the authors along with some captions of the test results performed in situ. In addition to this innovative technique, a second method based on sound scanning is proposed. The aim of this technique (under development) is to be able to predict any inner or outer degradation or damage that may appear in the cable's structure by recoding and interpreting the sound frequencies produced by breaking wires or strands. MATERIALS AND METHODS 1. Necessity for safety and monitoring

Transport structures sustained by bearing cables such as cableways, telepher line ways, cable cars, funitels, gondola lifts, etc., once entering service are subject to actions that in the majority of cases, are significantly different than the loads taken into consideration by the design codes. As a result, the paper focuses, besides in situ behavioural aspects, on aspects concerning the monitoring of these structures in order to confirm their safety in service [1]. The diameter of the transport

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cables is tightly bound to the forces that appear both in static and dynamic load conditions. Any change in diameter or loading conditions must be immediately detected and accounted for.

Gondola lifts are a modern and efficient solution for transportation in areas with rough or hilly terrain. Their length varies from 2 km up to 8 km, but designs with lengths of up to 36 km are under development. The sequence of the spans is chosen taking into account the possibility of pillar placement, such that all edifices, structures, buildings, water beds, high voltage lines and streets are avoided. In addition, the necessary expropriations should be kept to a minimum. Box pattern frames (closed frames) are mainly used for building the pillar, but rarely truss structures can be met too.

The advantages of using such transport structures are: • The versatility of their usage both for person transportation and for goods

transportation; • They can be in service 12 months a year, both under very cold and very hot

temperature conditions. The extended use of these transportation means allows for abandoning building terrestrial roadways which require expensive infrastructures. Bearing cable transportation structures require mandatory current monitoring according to existing national regulations, namely:

• Law no. 10/1995 - The law of quality in constructions; • Normative P130/1999 concerning construction's behaviour in time.

The conformation of safety for these structures is based on structural analysis performed both in static and dynamic load configurations, using real geometry, which implies that current observations are not sufficient [2]. The real geometry of bearing elements depends on many aspects including, but not restricted to: the intensity of the load, the type of material and the exploitation conditions. The main building materials used througout the times for building cables were: papyrus, camel's hair, flex and hemp. Starting with the year 1834, the first cables using steel wires were manufactured. This construction element became indispensable in many domains of structural engineering due to its singular properties such as high rupture strength compared to its sole weight, high flexibility and high durability. The cables used for bearing structures can have classical or special construction using metallic insertion elements with three or five edges. Steel cables are manufactured using high quality carbon steel, having a carbon content of 0.5% and rupture strength of approximatively 60 daN/mm2. By a procedure called wire drawing, the steel bar, having a circular cross-section, is transformed into wire [3]. Along with this procedure the rupture strength of the wire increases to approximately 120 - 200 daN/ mm2. After wire drawing, the wires are subject to a thermal treatment and thus the material regains its plastic properties. The individual wires are twisted around a central wire, in one or several layers forming strands. In turn, strands are twisted around a central core forming the final cable [4].

Lately, in highly corroding environments, cables manufactured using polypropylene, fibre-reinforced polymers (FRP), polyesters and nylon are used. The main characteristics of polypropylene and nylon are the facts that the ratio between the specific weight of polypropylene and the specific weight of water is 0.91 and the ratio between the specific weight of nylon and the specific weight of water is 1.14.

In the case of cable bearing structures manufactured using polymeric materials, the behaviour in time and the behavioural monitoring generates a delicate problem, due to the complexity of the elements involved and due to the aging and degradation of the structural polymeric materials.

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The elements involved in the aging and degradation of polymeric bearing cables are: • The content of atmospheric CO2 and SO2; • Precipitations and precipitation pH; • UV - rays; • The variation of the solar ray energy; • Temperature differences (day - night, summer - winter).

The objective of monitoring is to achieve a correspondence between the values of the control input and the values of the output. Monitoring can be of open or closed type. The open monitoring system does not perform a check on the relation between the two types of values (input vs. output). The monitoring systems described in the paper are of closed type, performing verifications between the values.

Generally, for in situ monitoring the following methods are used: • Visual observations; • Electrochemical methods such as the voltage method, magnetization method or

mechanical sound method; • Non-electrochemical methods such as microscopic observation, gravimetry, infrared

thermography, gammagraphy, radiography in the radar method [5]. 2. 3D scanning In order to perform visual diagnosis structural scans are necessary. It is recommended to perform structural scanning using a system of sliding digital cameras. This system was named MOV_CAM by the authors and the other researchers involved in the development. The system is pending patenting with the national Romanian authorities. The working phases of the system are:

• Image acquisition; • Image pre-processing; • Image processing; • Image analysis; • Pattern recognition; • Interpretation and management of results.

For implementing the first phase, a system of two cameras is used in conjunction with interior and exterior holograms. Three-dimmensional scanning is an operation of measuring the real object by acquiring images and 3D geometric data. After image acquisition, in the pre-processing phase noise is cleared out from the images. During the pre-processing phase, based on the 2D coordinates of the image points and using a reference system, the 3D coordinates are obtained. In the processing phase operations involving border detection and edge detection are performed. These operations are realised in order to obtain information about the location and the magnitude of existing degradations and defects. The main degradations that can be detected are:

• Loss of cable section, the loss of metallic cross-sectional area (LMA), through corrosion, wear and plastic deformations;

• Changes of the geometrical form; • Cup and cone fractures of wires [6]; • Shear wire breaks with diagonal ends [6]; • Tensile wire breaks with tapered ends [6];

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• Local faults (LF). Generally, for detecting LMA and LF-type degradations and defects, electromagnetic investigation procedures are used; these procedures are based on a magnetic head equipped with strong permanent magnets and sensors with distance counters. The software application allows the recording and display of LF and LMA degradations in a synthetic manner. It is mandatory that the images are acquired in real-time, continuously, in order to allow precise positioning of measurement points. Even if data acquisition through telemetry appeared in the 1980s, it gained momentum just after the widespread use of GPS systems. In (fig. 1) a sample of local fault (LF) detection is presented. The fault is enclosed in the rectangle.

Fig.1. Local fault detection snapshot

3. Sound scanning The rupture of composing wires of bearing cables can generate a high level of sound and can be a very important trigger for detecting the degradation of bearing capacity. The sound frequencies propagate through the cable material with a velocity of approximatively 5000m/s towards the anchoring points of the cables [7]. A design for the data acquisition components of the sound scanning system is presented in (fig. 2).

Fig.2. SONOCAB System - Data Acquisition

The intensity of the sound signal is proportional to the number of broken wires or strands, thus having the possibility of developing an efficient structural control system. The sensors that will be used in the test implementation of the system have to be robust, easy to mount and must have electromagnetic screening. The sensors are chosen in such a way that vibrations, environmental

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temperature and the effects of electric or magnetic fields have a very small influence on their functioning. The authors have named SONOCAB the bearing cable monitoring system based on sound scanning. The in situ data acquisition equipment must contain: sound sensors, filtering and amplification systems and a recording system. The calibration tests for the sensor require intense work in order to obtain the required transfer function. The sound taken into account when measuring is the sound transmitted through the cable from the sound source, namely the breaking point of one or several elementary wires, to the sound sensors. For evaluation purposes, only the direct sound generated by the rupture, will be taken into account. The filtering of refractory components is assured through electronic screening. For the directly transmitted sound, attenuation created by the cable mass has to be accounted for by using the mass laws specified in dB by equation (1):

TL = k log MS + r (1)

4. Interpretation of data using fuzzy logic Even though, monitoring and scanning problems are generally dealt with by a series of

techniques, the same cannot be said concerning the interpretation of the data resulting from monitoring. The problems posed by the interpretation of data resulted from monitoring bearing cable structures are of great complexity, due to the multiple random variables that intervene [8]. As a result we found our work on using fuzzy-logic based reasoning for interpretation and quantification of data resulted from monitoring [9]. The use of fuzzy-logic based reasoning can be successful in the case of large structural systems, where the human factor, the expert, appears as a main component in the concept of decision. The clustering of information is based on the following criteria: the nature of the monitored system, the purpose of the identification and the type of data that is to be processed. The main property of the real structural system, considered as an informational system, is uncertainty [10]. The main uncertainties for a bearing structure are: the connections between wires/strands and structural elements, the nature and magnitude of degradations, the position and the types of wire ruptures, the crack off, splitting up and erosion of wires, the existing corrosion type, etc. These uncertainties amplify exponentially with the increase in dimensions and the complexity of structures.

Uncertainties can be grouped in two main classes: imprecise information uncertainties and vague information uncertainties. As a result, the main problem resides in quantifying these classes of uncertain information [11]. The management of uncertainties in our system uses fuzzy logic. Following, the uncertain information is represented using fuzzy subsets with linguist descriptors [12]. Generally, the logic model is expressed using implications having the form of conditions that imply consequences, namely, if a predetermined model "A" exists, then it results a consequent model of "A" [8]. In this way, model "A" can be determined if fuzzy values are associated with fuzzy variables. In order to operate on uncertainties, having vague information, a confidence factor CF is used. This factor will have numerical values between 0.000 (false) and 1.000 (true). Mainly the structural damage of existing cable structures can be classified in categories like: no damage, slight damage, moderate damage, severe damage and imminent failure. The positioning of the degradations is obtained by using a database of models containing descriptive information or 2D and 3D models for real world situations like: isolated breakings, horizontal breakings, inclined breakings, masked breakings, etc. Corresponding attributes are created for each type and level of degradation and the models from the database will be tagged accordingly [13].

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CONCLUSIONS With the increasing degree of complexity of constructions (high-rise and with large-spans)

using bearing steel cables, an increase in the need of evaluating the associated risks appears. Continuous, periodic or continuous-periodic monitoring helps analyze and mitigate these risks. The results of monitoring constitute a starting point when taking decisions concerning interventions to the structure in order to avoid local or chain collapse. Even though monitoring and interpretation of structural monitoring of bearing cable structures implies whatever is best in the science and art of structural engineering, the number and the magnitude of uncertainties is very large. We have developed techniques based on quantifying these uncertainties using intelligent control systems with fuzzy logic reasoning. The usage in the survey activities of decision systems based on fuzzy logic allows the correct evaluation of the technical state of the bearing structures and helps take immediate measures for improving the safety levels of the monitored structure when it exhibits dangerous degradations.

We must consider that although we use fuzzy logic concepts during monitoring, fundamental aspects associated with structural survey remain valid, our techniques offering another viewing angle and another method for information processing. Our monitoring system allows autonomous monitoring and diagnostics of cable bearing structures by using real-time data transmissions from image or pressure sensors to a central processing facility by using wireless media. The system also allows focus on certain spans of a given cable structure. Alarm management in case of control parameter threshold value exceeding must be correlated in such a way that the users of the monitoring system are able to individually deactivate certain features of the decision process based on the consent of multiple experts in the domain. REFERENCES 1. STAHL, F.L., GAGNON, C.P. (1996), Cable Corrosion in Bridges and Other Structures: Causes and

Solutions, ASCE Press, ISBN: 978-0784400142, New York (NY), USA., pp.87-97. 2. C�T�RIG, A. and KOPENETZ, L.G. (1991), Time Surveillance and in Situ Testing by Dynamic

Methods of Steel Structure, Steel Conference, Timisoara, Romania, volume 3, pp. 259-265. 3. *** (2004), http://www.unionrope.com/Resource_/TechnicalReference/1164/Techreport%20101.pdf,

viewed at 05/03/3012. 4. ***(2011), http://www.unionrope.com/Resource_/TechnicalReference/1302/Wire-Rope-Users-Hand

boo k.pdf, viewed at 12/03/2012. 5. KOPENETZ, L.G. and GOBESZ F. ZS. (2007), The Structural Expertise of Steel Cables,

Transportation Infrastructure Engineering, edited by Intersections, Iasi, Romania, volume 4, pp. 49-62. 6. *** (2004), http://www.unionrope.com/Resource_/TechnicalReference/1166/Techreport%20103.pdf,

viewed at 20/03/2012. 7. PELLERIN, G. (2006), Acoustique architecturale: Théories et pratiques, electronic version,

http://assoacar.free.fr/archives/Cours/Ac%20des%20salles/Acoustique_Architecturale_CNAM.pdf, viewed at 05/01/2012.

8. KANDEL, A. (1986), Fuzzy Techniques in Pattern Recognition. John Wiley and Sons, New York, USA. 9. BOTHE, H. (1994) Fuzzy Logic, Springer Verlag, Berlin, New York, London. 10. DUBOIS, D. and PRADE, H. (1980), Fuzzy Sets and Systems: Theory and Applications. Academic

Press, London. 11. BOCKLISCH, F. (1987), Prozeßanalyse mit unscharfen Verfahren. VEB Verlag Technik, Berlin, Germany. 12. BANDEMEK, H. and TOTTWALD, A. (1990), Einführung in Fuzzy Methoden. Academic Press, London, UK. 13. LAZAR-MAND, F. and LISMAN, D.F. (2011), Comparative aspects between linear and non-linear

analysis of cable structures, Proceedings of the 9th International Symposium "Computational Civil Engineering", Iasi, Romania, ISBN: 978-606-582-006-7.

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ADVANCED IN SITU MONITORING TECHNIQUES FOR THE BAHAVIOUR OF HERITAGE STRUCTURES

LI�MAN Drago� Florin*, KOPENETZ Ludovic Gheorghe,

Technical University of Cluj-Napoca, *e-mail: [email protected] (corresponding address)

A B S T R A C T The necessity of historical building monitoring arises in the context of national and international existence of a high number of historical structures whose health state has to be preserved for a long period of time. Monitoring operations are vital taking into account the fact that the locations of the majority of these structures are in seismic areas, which have safety and reliability requirements listed at the highest levels. In the paper, two advanced non-destructive testing and monitoring systems are presented, which is based on a wireless sensor networks and 3D laser scanning. Due to the fact that reasoning processes are complex, there exists a strong need for knowledge and information specific to historical buildings, which can only be supplied by a structural expert. Besides the structural expert, it is mandatory to have an engineer with strong knowledge of advanced hardware and software systems required by the design, implementation, deployment, exploitation and servicing of structural monitoring systems.

Keywords: heritage buildings, 3D laser scanning, wireless sensor networks


INTRODUCTION Starting with the 1990's the importance of heritage structures and their true value is

reconsidered. In order to safely have these structures in service, it is not enough to execute rehabilitation works for them, it is also mandatory to perform a continuous monitoring activity. Monitoring existing heritage buildings with the purpose of having a general survey of the situation is a very expensive and difficult operation. The evaluation of the results obtained from monitoring existing structures is essential in determining their safety levels [1]. Nowadays, the most frequently used method includes periodic observations which usually consist of visual inspections followed by the evaluation of the monitoring results. The expert survey involves non-destructive testing besides the routine visual inspection, in order to determine both the characteristics of the used structural materials and to establish the types of in situ supporting structures [2].

In the paper the authors present monitoring techniques and methods using tools that can be implemented with a high degree of automation. Modern transducers and latest information processing techniques help performing monitoring tasks with a very good cost to efficiency ratio. One method that is in particular appealing is the innovative 3D image scanning technique. By using this method structural geometry and possible existing degradations, deformation, cracks and fissures can be determined. This technique is not singular to civil engineering; it is successfully used in medical or space engineering. MATERIALS AND METHODS 1. Monitoring concepts for heritage structures

Monitoring can be performed continuously or periodically and can have as purpose the global structure or just specific structural elements. Depending on the objectives, one can monitor the stress configuration, existing loads, etc. The monitoring technique one uses greatly depends on the existing state of the heritage building. When a certain technique is used, it has to be taken into

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account the fact that it is mandatory to perform a structural integrity observation both in border conditions and in normal service conditions. This has to be done in order to be able cu quantify the degree of structural safety [2]. For the validation of the efficiency and the accuracy of a monitoring technique the aspects presented above must be closely correlated both to the cost-results ratio and to aspects concerning the data interpretation obtained from real-time acquisition sources (sensors, transducers, etc.). It is of utmost importance to compare actual monitoring data to data determined by survey using experimental test methods [3]. Every type of monitoring has to have a preliminary in situ inspection phase of the placement conditions, the influence of environmental factors and the execution of specific tests and measurements. In the case of heritage constructions, special attention has to be given to all types of changes that were performed on the construction in the interval of time since it entered service.

2. In situ specific tests

Testing of structural materials involve core control operations, examination of existing internal and external cracks, fissures, penetration resistance, compression resistance determinations based on concrete test hammer recoil measurements, pull-off, pull-out and break-off tests.

External and internal crack or fissure monitoring can be performed using one of the latest technologies in the field of wireless networks: wireless sensor networks (WSN). WSNs can be designed and implemented by using autonomous nodes equipped with their own continuous voltage power supply, wireless communication transmitter/receiver for data dissemination and a very large range of sensors for data acquisition [4]. A test deployment solution under development by our research team is based on Waspmote technology developed by a spin-off company from the University of Zaragoza, Spain. This solution based on Waspmote uses several devices positioned in predetermined locations on the heritage structure [5]. Each node is equipped with crack detection, crack propagation and linear displacement sensors. Besides these sensors the mote has classical temperature, humidity and dust sensors. The crack detection sensor consists of a very small conductive strand with a very low resistance value embedded in a fibre-glass film. In the case of a crack development, the film will crack, sending a signal to the microcontroller embedded in the node. For installing the sensor, a special adhesive must be used in order to fix the fibre-glass film on the monitored surface. In long term installation, the manufacturer recommends the use of a protective coating [5].

In the case of the crack propagation sensor of the node, the operating mode resembles the operation mode of the crack detection sensor, the difference being that it is composed of several resistive strands placed in parallel whose ruptures cause a variation in the total resistance of the sensor. The modified resistance is measured with a voltage divider and a voltage proportional to the number of broken strands is obtained at the input of the mote's microcontroller.

The linear displacements are measured with the help of two sensors mounted on the node. One of them is the accelerometer and the other one is a potentiometer whose wiper moves along an axis guided by the sensor body. By fixing both ends on the structural element, we can measure the displacements and possible vibrations. The sensor has an accuracy step of approx. 10 �m and a total range of 10mm with a linearity of ±0.5% [6].

Depending on the type of construction there are specific tests for heritage constructions built using stone or brick masonry, concrete, metal (iron, steel) or wood. Thus, for stone and brick masonry and concrete based heritage constructions specific tests are performed in order to determine the hardness degree of the visible surface, in depth inspections using ultrasounds, radiographs, neutron absorption, permeability degree, humidity, the assessment of the type of

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plaster and its composition, as well as a petrography study. For heritage structures built using reinforced concrete tests for determining the depth of carbonation, chloride concentration and sulphates content have to be run. For heritage structures built using metal, tests are performed for the specific resistance value, linear polarization resistance, coating thickness, corrosion degree and the magnitude of fatigue cracks, collisions and over-loading.

3. Monitoring using terrestrial laser scanning

Since the development of the global positioning system in the 1970s there is a quick solution for the problem of three-dimensional point positioning on the terrestrial surface. On this basis was built the method of rapid determination of spatial positioning of objects using the laser scanning technique. 3D laser scanners resemble photo cameras by collecting information only from the visible points using a conic visual field. This information is in turn used to extrapolate the objects shape through the reconstruction process. Current systems allow acquiring information concerning the colour of the objects surface, too [7]. This concept was adapted for the heritage structure monitoring technology by digitally determining the geometry of a certain spatial object with high speed and accuracy. The results of the measurements are recorded in a database as a multitude of points, cited as a cloud of points in the technical literature [8].

The 3D scanning equipment is composed of the hardware 3D scanning system and the software applications as well as the transfer and data processing components of an editing software application. The laser scanning technology can be applied both statically and dynamically.

Dynamic 3D scanning requires that the measuring instruments are mounted on a moving platform placed on a terrestrial vehicle (terrestrial scanning) or in a flying object (aerial scanning). In the case of static scanning, the position of the measuring instruments is fixed during the entire data acquisition time interval. For very large heritage structures, terrestrial or aerial dynamic scanning is used. Dynamic scanning operations involve high costs due to the fact that they use inertial navigation systems (INS) and GPS systems with very high accuracy. For regular heritage structures static scanning on medium to long distances is used in order to obtain accurate results. High accuracy is reached due to the higher number of acquired points. In the case of singular structural properties of heritage structures it is recommended to use short distance static scanning.

In the test setups high accuracies were obtained using the RIEGL LMS-z390i high resolution 3D terrestrial laser scanner in conjunction with the RiSCAN PRO processing software application [9]. The output data from the RiSCAN PRO software was fed into our software application developed for post-processing of data.

4. Structural analysis problems solved using laser scanning

The quick monitoring technology of the shape of a heritage structure by using a single scan or several scans at different time intervals allows the real-time survey of deformations, cracks and stresses. It is also the perfect support information for taking rapid structural intervention decisions [10]. In the case of the single scanning procedure, the displacements and deformations of the structural elements are traced by measuring the characteristic points of the structural elements against the certain reference planes. Multiple scans performed at certain time intervals allow the visualisation of the evolution of degradations by recording the results and then comparing them against consecutive scans. The scanning interval depends on the type of structure and can span from a few days to several months or years. By interpreting the records in the database, a software application can predict an evolution model for the degradations and the necessary interventions. In addition, the time frames when these interventions are to be performed can be established

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CONCLUSIONS For heritage constructions having a high degree of complexity both caused by their height

and by their span appears the necessity for the assessment of risks associated with their exploitation under safety conditions. Continuous, periodic and periodic-continuous monitoring allows analyzing and evaluating the risks. Using an advanced monitoring technique based on wireless sensor networks allows a radically different approach to the monitoring problem by eliminating issues concerning mounting wires and energy sources for the hardware data acquisition components. Another advantage is the small dimension of the sensing elements which is especially welcome in the case of heritage buildings having valuable works of art on their structural elements. This compactness assures a small visual impact for the visitors. In addition, in allows direct accurate monitoring of environmental parameters such as temperature, humidity, dust or direct detection of more complex parameters such as cracks, crack propagations or displacements. The laser scanning technique has several advantages besides the determination of the existing geometric shape at a certain moment in time, such as the possibility of estimating the evolution of the deformation or displacements and the assessment of cracks or fissures in real-time.

Nevertheless, both methods exhibit a certain number of uncertainties during the interpretation of the measured parameters, which have to be dealt with. The solution that we base our research on, involves the use of fuzzy-logic reasoning for uncertainties quantification.

REFERENCES 1. ARMER, G.S.T. (2001), Monitoring and assessment of structures, London, UK, Spon Press, ISBN:

978-0419237709. 2. HEJLL, A. (2007), Civil Structural Health Monitoring: Strategies, Methods and Applications, Doctoral

Thesis, Lulea University of Technology, Department of Civil and Environmental Engineering, Division of Structural Engineering, Lulea, Sweden, ISBN: 978-9185685080.

3. MALHOTRA, V.M. and CARINO, N.J. (2004), Handbook on Nondestructive Testing of Concrete, 2'nd ed. Florida, USA, CRC Press LLC, ISBN: 0-8031-2099-0.

4. CERIOTTI M., MOTTOLA L. and others (2009), Monitoring Heritage Buildings with Wireless Sensor Networks: The Torre Aquila Deployment, Proceedings of the 8th ACM/IEEE International Conference on Information Processing in Sensor Networks, San Francisco, CA, USA, April 13-16.

5. ***, (2012), http://www.libelium.com/documentation/waspmote/smart-cities-sensor-board_eng.pdf, technical guide electronic version, viewed at 03/04/2012.

6. ***, (2012), http://www.libelium.com/documentation/waspmote/waspmote-datasheet_eng.pdf, electronic version datasheet, viewed at 02/04/2012.

7. KUTULAKOS, K.N. and STEGER E. (2008), A Theory of Refractive and Specular 3D Shape by Light-Path Triangulation, International Journal of Computer Vision, Vol. 76, No. 1, pp 13-29.

8. FECHTELER, P. and EISERT, P. (2008), Adaptive color classification for structured light systems, IEEE Conference on Computer Vision and Pattern Recognition, CVPR 2008, Anchorage, USA, ISBN: 978-1-4244-2339-2.

9. ***, (2010), http://www.riegl.com/uploads/tx_pxpriegldownloads/10_DataSheet_Z390i_20-04-2010.pdf, viewed at 07/01/2012.

10. GRUSSENMEYER, P., LANDES, T., VOEGTLE, T. and RINGLE, K. (2008), Comparison methods of terrestrial laser scanning, photogrammetry and tacheometry data for recording of cultural heritage buildings, XXIst ISPRS Congress: Commission V, WG 2 , Beijing, P.R.C., pp.213-218.

11. REMONDINO, F. (2011), Heritage Recording and 3D Modelling with Photogrammetry and 3D Scanning, Remote Sensing 2011, Vol. 3, No. 6, pp. 1104-1138.

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LINEAR THERMAL BRIDGES AT VERTICAL ELEMENTS OF BUILDING STRUCTURE

LUPAN Lidia-Maria *, MOGA Ioan,


A B S T R A C T This paper presents comparative study on energy efficiency between constructive system consisting of steel columns with closure of masonry and the classic system with concrete colums embedded in masonry, the first ones proved to be more efficient in terms of heat loss through thermal bridges, reducing the energy consumption and emissions in atmosphere. The results were obtain by comparation of linear thermal transmittance coefficients, which were calculated manually as they are not found in any catalogs or design standards in the country or other catalogs treating this problem.

Keywords: energy eficiency, steel column, linear thermal transmittance, heat loss

Received: February 2012 Accepted: March 2012 Revised: April 2012 Available online: May 2012

INTRODUCTION The implementation of the European Directive on the Energy Performance of Buildings

(EPBD) is a milestone towards the improvement of energy efficiency in the building sector [1, 2]. Despite the insulation requirements specified by national regulations, thermal bridges in the building's envelope remain a weak spot in the constructions [3, 4]. Heat loss through thermal bridges can reduce thermal resistance of the current field wall by up to 45%. Thermal bridges are areas of the construction elements, which due to structural or geometric composition have a greater permeability that the rest of the thermal elements. In adition to higher heat losses, thermal bridge formation may have negative consequences in terms of interior comfort. Low surface temperatures in the region of thermal bridges can lead to surface condensation if they are below the dew point of the air.


Preventing development of thermal bridges are important points not only for costs reduction or energy but also prolonging the life of a building and preserving its value.

In calculation, the effect of thermal bridges is taken into acount by introducing linear thermal transmittance coefficient “�”, which brings a unidirectional correction calculation, taking into acount both the presence of constructive thermal bridges and the behavior of real two-dimensional thermal flow in the areas of inhomogeneity construction elements and is given by the relation:

�=L2D-Ui*li (1)

where: � - is the linear thermal transmittance (W/mK) ; L2D - is the thermal coupling coefficient obtained from a two-dimensional calculation of the component separating the two environments being considered (W/mK) [5]; Ui - is the thermal transmittance of the one - dimensional component i which is separating the two environments being considered (W/m2K) ;

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li - is the length within the two-dimensional geometrical model over which the value of Ui applies (m) [5].

Linear thermal transmittance is the rate of heat flow per degree temperature difference per unit length of the thermal bridge [6]. The units for “�” values are Watts per meter Kelvin (W/mK) [7]. Linear thermal transmittance “�” does not differ according to climates zones. For more usual details it is often convenient to use tabulated default values that can be found in thermal bridge catalogues.

Because the values of linear thermal transmittance “�” for these details are not in the thermotechnics design standards in our country, results were based on manual calculation using the relationship mentioned above (1).

To calculate the linear thermal transmittance “�”, in the present paper, it were made constructive details with three types of steel columns: HEA 240 (fig.1-a), HEB 240 (fig. 1-b), HEM 240 (fig. 1-c) which were embedded in masonry Porotherm 25cm type, autoclaved aereted concrete of 25 cm and solid brick of 24 cm.

Fig. 1. Horizontal section through exterior wall and steel columns with 10 cm isulation

a) HEA 240 , b)HEB 240, c)HEM 240

Keeping same characteristics of walls, linear thermal transmittance coefficients were calculated in case the steel columns were being replaced with concrete columns and results were compared, establishing which of the two solutions are more energy efficient. There were being considered solutions where the walls were assumed to be thermally non-isulated, isulated with 10 cm, 15 cm, 20 cm EPS (expandable polystirene).

a b

c

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Fig. 2. Horizontal section through exterior wall and columns with 10 cm isulation

a) Column23x25 , b) Column 24x25, c) Column 25x27

Calculation was performed by a simplified method taking into acount following assumptions: - apply the stationary temperature conditions; - all physical properties are independent of temperature [8]; - there is no heat source inside the element studied..

Adiabatic (zero heat flow) sectioning plans were places at 1.00 m from the core. Thermal conductivity coefficients were given for materials in condition of normal humidity during opeation: - interior plaster �=0.87W/m2K ; - exterior plaster �=0.93W/m2K ;

- expandable polystyrene �=0.044W/m2K ; - steel �=60W/m2K ; - concrete �=1.74W/m2K ; - rigips plates �=0.30W/m2K ; - Porotherm mansory �=0.19W/m2K ; - autoclaved aereted concrete masony �=0.32W/m2K ;

- solid brick masonry �=1.61W/m2K [9] ; Surface heat transfer coefficients were:

i =8 W/m2K ; e =24 W/m2K

Fixing polystiren dowels were made of plastic and they didn't influence the thermal resistance of the wall.

a b

c

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RESULTS AND DISCUSSIONS According to the method describe, linear thermal transmittance coefficients were calculated, and they were centralized in the table bellow (Table 1.)

Table 1- Linear thermal transmittance coefficients (W/mK)

*) Because the geometric features of HEM 240 exceed the width of the solid brick, this solusion was not considered. **) When we use solid brick mansory the dimensions of concrete column are 23x24 cm, 24x24 cm, 24x27 cm.

Thermal resistances in current field obtained in each cases are:

For Porotherm mansory: Without isulatioan: R=1.506m2K/W 10cm of isulation: R=3.778m2K/W 15cm of isulation: R=4.915m2K/W 20cm of isulation: R=6.051m2K/W

For Autoclaved aereted concrete mansory: Without isulatioan: R=0.971m2K/W 10cm of isulation: R=3.244m2K/W 15cm of isulation: R=4.380m2K/W 20cm of isulation: R=5.517m2K/W

For Solid brick masonry: Without isulatioan: R=0.339m2K/W 10cm of isulation: R=2.612m2K/W 15cm of isulation: R=3.748m2K/W 20cm of isulation: R=4.884m2K/W To underline the obtained results, for each type of masonry were made graphs accorrding to

expandable polystirene thickness.

Type of columns

Type of mansory EPS thickness HEA

240

Concrete Column 23x25

HEB 240

Concrete Column 24x25

HEM 240

Concrete Column

25x27 **) Without is. -0.069 +0.498 -0.056 +0.519 -0,047 +0,584 10cm -0.026 +0.027 -0.025 +0.028 -0,020 +0,032 15cm -0.018 +0.015 -0.017 +0.015 -0,015 +0,017

Porotherm mansory

375x250x238 20cm -0.013 +0.009 -0.013 +0.009 -0,011 +0,011 Without is. -0,150 +0,416 -0,141 +0,435 -0,048 +0,489 10cm -0,036 +0,017 -0,036 +0,018 -0,032 +0,020 15cm -0,024 +0,009 -0,023 +0,009 -0,021 +0,010

Autoclaved aereted concrete mansory

600x250x100 20cm -0,017 +0,005 -0,017 +0,005 -0,015 +0,006 Without is. -0,521 +0,020 -0,514 +0,021 *) +0,024 10cm -0,052 +0,004 -0,051 +0,0004 *) +0,0004 15cm -0,032 +0,0008 -0,031 +0,0002 *) +0,0002

Solid brick masonry 240x115x63

20cm -0,022 +0,0001 -0,021 +0,0001 *) +0,0001

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Fig.3. Graph of “�” values for Porotherm masonry

Fig.4. Graph of “�” values for Autoclaved Aereted Concrete masonry

Fig.5. Graph of “�” values for Solid Brick masonry

Linear thermal transmittance cofficients values of all three masonry types:

Porotherm 25cm, autoclaved aereted concrete 25cm, and solid brick 24cm (fig.3, fig.4, fig.5) in the cases where we used steel columns had lower and negative values comparing with the cases where we used concrete columns. It means that the heat flows through the system composed of steel columns were lower than the ones through the system composed of concrete columns.

The sign (+) represents a reduction in thermal resistence corrected R’, to the unidirectional thermal resistance R , the sign (-) has a lower frequency and means an increase in the value of R’ to the value of R. The sign for linear thermal transmittace coefficient values shows the heat transfer direction from warmer to colder [10,11]. In steel columns cases the sign was „minus” , meaning

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that heat transfer direction was from steel columns area to current field area. When we used concrete columns, thermal transmittance coefficients had „plus” sign and the heat transfer was from current field area to concrete columns area.

Energy efficiency of steel columns was lower, the bigger was the section of europrofile. From all these three types analyzed the most energy efficient was HEA 240, having an egual section S=76.8cm2, compared to HEB 240 with S=106cm2, or HEM 240 with S=200cm2 [12]. CONCLUSIONS

Even if steel has a high thermal conductivity (�) compared with many other construction materials, systems that include steel columns, proprely designed can lead to low heat flows [13]. One efficient way of reducing thermal bridging in steel construction is to eliminate the thermal bridge by keeping the steelwork within the isulated envelope. This areas require careful consideration.

Steel is a 100% recyclable material which has a positive impact on energy saving.

REFERENCES 1. THEODOSIOU, T.H. and PAPADOPOULOS, A.M (2008), The impact of thermal bridges in the energy

demand of buildings with doulbe brick wall construction, Energy and buildings, Vol. 40, Issue 11, pp. 20083-20089.

2. *** http://www.cenerg.ensmp.fr, accessed at 05.04.2012. 3. LITTLE, J. and BEÑAT, A. (2011), Thermal bridging- Understanding its critical role in energy

efficiency, Construct Ireland, Issue 6, vol. 5. 4. *** http://www.sciencedirect.com , accessed at 05.04.2012. 5. *** http://www.tud.ttu.ee, accessed at 05.04.2012. 6. *** SR EN ISO 10211-2:2005 Thermal briges in building construction –Calculation of heat flows and

surface temperatures –Part 2:Linear thermal bridges 7. *** SR EN ISO 10211-1:2005 Thermal briges in building construction –Heat flows and surface

temperatures – Part 1: General calculation methods. 8. *** http://www.eng.uci.edu, accessed at 05.04.2012. 9. MC 001/1-2006, Methodology for calculating the energy performance of building envelope .Partea I-a-

Building anvelope 10. *** SR EN ISO 10211-1:2005, Thermal briges in building construction –Heat flows and surface

temperatures – Part 1: General calculation methods. 11. *** EN ISO 13789-99, Thermal performance of buildings – Specific transmission heat loss –

Calculation method. 12. *** http://www.europrofile.ro/date_tehnice.php , accessed at 05.04.2012. 13. *** http://www.steelconstruction.org/, accessed at 05.04.2012.

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CORRELATIONS BETWEEN GEOTECHNICAL PARAMETERS OF TRANSILVANIAN COHESIONLESS SOILS BASED ON TRIAXIAL

LABORATORY TESTS RESULTS

MOLNAR Iulia, Technical University of Cluj-Napoca, e-mail: [email protected]

A B S T R A C T The current rhythm for construction works and their complexity raise various geotechnical problems. One of the key issues in both the execution and design stage is to ensure stability while building especially when unfavorable soil conditions are present. Accurate information regarding geotechnical parameters represents an important subject in solving various problems using computer programs used worldwide. Most times the number and complexity of parameters needed as input to specialized computer programs is very high and requires advanced knowledge of soil mechanics science. Because of this, very often complete data for such analysis are not included in geotechnical studies. As an immediate consequence of this problem, determining correlations between geotechnical parameters both mechanical and physical in order to facilitate the input process for computer analysis represents an actual necessity. The paper presents ways of processing parameters determined based on laboratory triaxial test and correlations between them. The study was conducted on a cohesionless soil from Transylvania Region near Baia Mare City.

Keywords: geotechnical parameters, triaxial laboratory tests, cohesionless soil, correlations between parameters


INTRODUCTION The purpose for field and laboratory tests is to provide the needed data for optimal design

process of soil constructions, foundation systems calculation, bearing capacity calculation, tunneling and slope stability analysis. A complete characterization detailing the intricate and complex response of soils remains a challenging task that can only be realized through careful drilling and sampling program coupled with detailed laboratory testing [1]. However, the situation when complete data for computer analysis is not included in the geotechnical studies is very often encountered. The determination of different correlations between the geotechnical parameters will simplify the input process for computer analysis and will also facilitate de process of selection for the correct parameters needed for the analyze [2]. The main purpose of this study is about to determine the required parameters from the geotechnical parameters that can easily be determined from laboratory tests. However relations between the geotechnical parameters need to be accurately calculated in order to precisely represent the true behavior of the soil.


The study for this paper is made based on the results from triaxial compression tests. Triaxial tests is the most complex test procedure that can be performed in geotechnical laboratories. It implies a complex procedure that requires advanced knowledge of soil mechanics, laboratory testing procedures and sample preparation.

6. Triaxial compression laboratory test principles

The triaxial test is the laboratory test that provides the most accurate results because of its complexity and also because it manages to simulate the best the situation on the site by creating the

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same stress state for the tested samples. [3]. The triaxial test is performed on a cylindrical sample, usually having the diameter 100 mm, and the height twice as big as the diameter.

Triaxial apparatus is a complex system made of several components so that during the test sample can be subjected to various triaxial controlled tensions. Also throughout the test data regarding: the value of pore water pressure, volume deformations, vertical displacement and axial stress are provided (fig.1) [4]. Triaxial shear testing can be performed in several ways depending on the situation in the field [5]:

� The UU (unconsolidated-undrained)test. This type of test is usually performed when the speed of execution for the construction is faster than the consolidation speed for cohesive soils.

� The CU (Consolidated-undrained)test. This type of test is usually performed when, after the consolidation of the foundation soil under the existing construction new loads appear and there is no water drainage possibility.

� The CD (Consolidated-drained)test. This test is usually done when there is water draining conditions as consolidation pressure increases.

Thus the complexity of data obtained from the triaxial tests allows different and various possibilities for processing [6]. Due to extensive areas that triaxial tests are used it is very important that data provided to be accurate. Possible errors should be anticipated and a careful analysis must be made on the values of correction factors and how to apply them [7]. Triaxial compression tests are also used as a component in wellbore stability, sand production and subsidence calculations and also for mine and excavation design.

Fig.1. Triaxial apparatus

7. Triaxial compression tests performed on cohesionless soil extracted from Baia-Mare Area

Triaxial shear tests on cohesionless soil present certain particularities especially in preparation of samples. Before their construction, in order to ensure accurate test results, pore indexes describing the maximum and minim density were determined ( mine and maxe ). With the pore indexes values the density index DI of the samples subjected to the triaxial compression test were calculated [8]. If field conditions allow the initial density index should also be determined. In this case the sample subjected to triaxial compression can be created at the same density index as that cohesionless soil had in the field before extraction. The triaxial tests discussed in this paper were executed on a cohesionless soil extracted from a site near Baia-Mare City from Transylvania Region.

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Regarding the geomorphological aspect of the tested soil the location where samples were extracted belongs to the Somes Depression from the right side of the river Somes. The area is characterized by alluvial material with chaotic bedding. The formations are composed of quaternary deposits represented by clays and gravels and sands. Based on the granulometric analysis the material tested in the laboratory is a medium sand with round particles and it was extracted from a depth of about 4m [9]. During the laboratory testing, nine triaxial tests were conducted in consolidated-drained system (table 1). The tested samples were constructed at three different density indexes. The samples with the same density index were tested at three different confining pressures.

Table 1. Density Index and confining

pressures or tested samples Sample Number

Pore Index

initiale Density Index

DI Confining pressure

[kPa] 1.1 0,50 0,95 100 kPa 1.2 0,50 0,95 200 kPa 1.3 0,50 0,95 400 kPa 2.1 0,67 0,15 100 kPa 2.2 0,67 0,15 200 kPa 2.3 0,67 0,15 400 kPa 3.1 0,55 0,71 100 kPa 3.2 0,55 0,71 200 kPa 3.3 0,55 0,71 400 kPa

RESULTS

Based on the values recorded after the failure stageof the triaxial compresion tests the effective stresses 3 'σ and 1 'σ were calculated. Based on the calculated values of the effective stresses the effective shear strength parameters were determined by using the Mohr-Coulomb Model (fig.2), (fig.3), (fig.4).

This model is the most common procedure used for determining the effective shear strength parameters from triaxial laboratory tests. Mohr-Coulomb model relies on a line defined by the Coulomb failure stress and the stress circles of Mohr. The field of failure is given by the cohesion and internal friction angle [10].

Fig.2. Determination of effective shear strength parameters

using Mohr-Coulomb model for the samples 1.1., 1.2, and 1.3

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DISCUSSIONS For a certain type of soil the purpose is to establish relations between the geotechnical

parameters both mechanical and physical in order to facilitate the input process for computer analysis. Based on the initial density index of the samples and on the processed data from the triaxial tests a linear relation between the initial density index and effective shear parameters was established.The frictional angle for the tested sand can be calculated with the following relation (fig.5): ' 3.2587 35.784DIϕ = + (1)

Fig.5. The influence of the density index on the effective frictional angle

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If the density index directly influences the effective frictional angle than it will also influence its residual value. Similar with the previous determination, the relation between the residual frictional angle and the density index was expressed in the following way (fig.6):

' 2.8022 34.656DIϕ = + (2)

Fig.6. The influence of the density index on the residual value of frictional angle

In the case of sands with maximal densification and minimal porosity, the particles slip

tending to the state of maximal loosening. In this way, the soil is loosed by shearing. The loosening phenomenon of dense sands by shearing is named dilatancy.

The effective frictional angle value is afferent to the maxim value of the shear strength. After the maximum value for the shear strength is reached it starts to decrease until it reaches the residual value that is afferent to the residual frictional angle. The diference between the effective frictional angle and the residual value of the frictional angle represents the dilatancy angle. The dilatancy angle is a very used parameter in geotechnical engineering problems especially as an input parameter for computer analysis. Based on the sample principle a relation between the density index and the dilatancy angle was determined (fig.7):

0.4565 1.1279DIψ = + (3)

Fig.7. The influence of the density index on the dilatancy angle

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CONCLUSIONS A complex analyze of the behaviour of a cohesioless soil and also the analyze of its

geotechnical parameters represents a challenge task and it requires advance knowledge of soil mechanics. Also accurate results must be ensured by a careful analyze of each step of the geotechnical parameters determination process. Determination of correlations between geotechnical parameters that represent the true behavior of a certain type of soil studied is an important task that will facilitate the next steps of modeling and design of various structures. Also in terms of modeling, the use of correlations between the geotechnical parameters subject may be carried out further in order to evaluate different computational models used in geotechnical engineering in order to highlight the advantages and disadvantages of the used model for a particular type of soil. Another directions of the use of correlations between the geotechnical parameters is represented by the study for simulation the triaxial compression test in a finite element program which represents the subject of a further study.

ACKNOWLEDGMENTS

This paper was supported by the project “Doctoral studies in engineering sciences for developing the knowledge based society - SIDOC” contract no. POSDRU/88/1.5/S/60078, project co-funded from European Social Fund through Sectorial Operational Program Human Resources 2007-2013.

REFERENCES 1. BRINKGREVE, RBJ. (2005), Selection of soil models and parameters for geotechnical engineering

application, Geotechnical Special Publication No.128, ACSE, pp.69-98. 2. I.SOKOLIC, Z.SKAZLIC (2011), Triaxial test simulation on Erksak sand using hardening soil model,

Proceeding of the 15th European Conference on Soil Mechanics and Geotechnical Engineering, 12-15 September, Athens, Greece.

3. HUNT ROY E. (2005), Geotechnical engineering handbook, Taylor and Francis Group Ltd. 4. *** BS 1377-8:1990 Methods of test for Soils for civil engineering purposes Part 8: Shear Strength

tests (effective stress), Bechtel Ltd. 5. STANCIU A., LUNGU I. (2006), Funda�ii (Foundations), Tehnica Ltd. 6. N. BELHEINE, J.P.PLASSIARD (2009), Numerical Simulation of drained triaxial test using 3D

discrete element modelling, Computers and Geotechnics, Vol. 36, pp.320-331. 7. MAYNE P.W., COOP M.R. (2009), Geomaterial behaviour and testing, 17th International Conference

on Soil Mechanics and Geotechnical Engineering Alexandria, Egypt, pp.2777-2872. 8. ROMAN F. (2011), Aplica�ii de inginerie geotehnic� (Geotechnical Engineering Aplications), Papyrus

Print Ltd. 9. *** STAS SR EN ISO 14688-1/(2004), Cercet�ri �i încerc�ri geotehnice.Identificarea �i clasificarea

p�mânturilor. Partea 1: Identificare �i descriere (Researches and geotechnical tests. Identification and clasification of soils. Part 1: Description and identification).

10. NETO E.A. DE SOUZA, PERIAE D., DRJ Owen (2008), Computational Methods for Plasticity: Theory and Applications, John Wiley and Sons Ltd.

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SEISMIC ENERGY DISSIPATION IN STRUCTURAL REINFORCED CONCRETE WALLS WITH STAGGERED OPENINGS

MO�OARC� Marius*, STOIAN Valeriu,

“Politehnica” University of Timisoara, *e-mail: [email protected] (corresponding adress)

A B S T R A C T Reinforced concrete structural walls are used for buildings in seismic areas because they satisfy the resistance, rigidity, stability and ductility requirements imposed by earthquakes on buildings. Among the factors which influence the capacity of seismic energy dissipation of RC shear walls, an important role is played by the door and window openings by their dimensions and positioning. If until 1985 the scientific literature recommended shear walls with staggered openings for buildings with high settlements, the earthquake from Chile showed that they are able to dissipate a large amount of seismic energy. This article presents a comparison between the dissipated seismic energy determined by theoretical studies and experimental tests performed on five lamellar structural reinforced concrete walls: one without openings, one with vertically ordered openings and three with staggered openings, differentiated by the position of the openings. The results presented in this article confirm the capacity of walls with staggered openings to dissipate seismic energy, in function of the vertical positioning of the door openings, lacking special reinforcing measures.

Keywords: RC shear walls, staggered openings, dissipated energy, earthquake, pushover analysis, experimental test


INTRODUCTION The seismic behaviour of structural walls with staggered openings started being studied after

the 1985 earthquake in Chile, when high rise buildings with these types of walls had a very good behaviour, recording limited damages. Since then, several researches on this type of walls had been performed by Aejaz & Wight [1], Wallace [2], Marsono & Subedi [3][4], Hui & Bing [6]. These researches provided great contributions to the calculus and dimensioning method, ductility, degradation of rigidity, dissipated seismic energy and the failure modes of these walls in function of the positioning of the openings, the direction of action of the forces, reinforcing details and the positioning of the bulbs and web in this type of walls. The authors present within this article recent information regarding the dissipated seismic energy of lamellar walls with identical amount of reinforcement, but differentiated by the staggered positioning of the openings. Results were obtained based on theoretical and experimental researches performed within the Civil Engineering Department at the Politehnica University of Timisoara [7], [8], [9]. Research results show an increase in rigidity and seismic energy dissipation of these walls, lacking special measures of reinforcing and assembly which should be clearly stated in the design codes. MATERIALS AND METHODS

Research has been conducted on four types of walls of shear walls with opening: three staggered shear walls (SW23, SW45, SW67), and one with vertically ordered openings (SW8) and one without openings (SW1). The studied walls with openings were differentiated by α angle values. The notation of walls with openings, depending on the location of openings and direction of seismic load is presented in Fig.1. Physical-mechanical The D dimensions of the and forces acting on shear walls were chosen so as to study the influence of openings on shear walls stiffness. with average height with the ratio of sides hw / lw > 2 and to allow the casting and handling of

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experimental models. All the shear walls were subjected to vertical constant loads actions (N=50000 daN) and monotonically increasing horizontal forces actions applied only to the top. Material characteristics for linear and nonlinear analysis are presented in Table 2.

Fig. 1. Notations and loading condition for the experimental models

Models dimensions are ¼ the size of real walls. In Table 1 are indicated the dimensions of the

real wall and of the experimental models.

Table 1. Dimensions of the real structural wall – and of the experimental model

Dimension Notation Shear wall [mm] Experimental model [mm] Wall height hw 10400 2600 Wall width lw 5000 1250

Wall thickness bw 250 80 Storey height hs 2600 650 Doors height hd 2000 500 Doors width ld 1000 250

Table 2. Material characteristics of

structural shear walls Rebars

Rebar diameter 6 mm Characteristic strength fsk = 355 N/mm2

Modulus of elasticity Ea = 210 kN/mm2 Concrete

Average tensile strength fctm = 3.0N/mm2 Average compressive strength fcm = 50 N/mm2

Modulus of elasticity Eb = 34 KN/mm2 Compressive strain 3,5‰

1. Theoretical models

Post elastic behaviour analysis of concrete shear walls at horizontal loads was performed using the computer software BIOGRAF [10]. The computing software performs a nonlinear 2D analysis based on load increments proposed by the user. It is used for surface element loaded in their plane, meshed as anisotropic finite reinforced concrete elements in a plane stress state. The analysis results were the stress and strain state in concrete and reinforcement and the physical

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condition of the element (cracked or not, plasticized, crushed) for each load level. The studied shear walls were meshed into triangular finite elements, their size being imposed by the position of reinforcements.

All the walls were reinforced with the same amount of horizontal and vertical reinforcement. They are distinguished by vertical reinforcement casing position, which changes depending on the location of openings. Horizontal and vertical reinforcements should be provided on both sides of the walls, both at the ends of the piers and in the span. Around the openings in the walls should be arranged casings of reinforcements. The casings were made of four bars provided with stirrups. In Fig. 2 is presented graphically: the FEM models, cracks and crushed concrete in ultimate limit state.

Fig. 2 a. FEM model; b. Cracks in ultimate limit state;

c. Damage in concrete at ultimate limit state 2. Experimental models

Wall model has the height of 2400 mm, width 1250 mm and storey height 600 mm. In order to avoid failure due to the loss of lateral stability induced by the missing of floors and bulbs, a wall thickness of 80 mm was adopted. The dimensions of the openings were 250mm x 500mm. Experimental models have been provided with a block foundation with height of 400 mm, width 350 mm and a length of 1750 mm. The specimens were cast together with the foundation block. According with Prof. V. Stoian, experimental models were reinforced with the same amount of reinforcement [8]. In order to ensure adherence of the concrete, were used PC 52 profiled 6 mm diameter reinforcement bars. The reinforcement was arranged as mesh on both sides of the walls. Around openings were placed reinforcement chassis made up of four bars of 6 mm diameter, tied with stirrups of the same diameter of OB37. In order to avoid joints and achieving good compaction of the concrete, model walls were cast and vibrated horizontally in metal casing. Experimental models reinforcement is presented in Fig. 3. The coating with concrete of all reinforcements was 10 mm. To avoid local crushing of concrete in the areas of application of horizontal and vertical forces at the top of the walls, were placed steel U8 profiles, throughout their length, and fingerprinted metal plates 8 mm thick ,on the lateral sides. Vertical spacing between reinforcement bars were kept the same for all experimental models, while the horizontal distance between reinforcement bars varied from one model to another, depending on the position staggered openings. Reinforcement

a b c

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percentages of experimental models are presented in Fig. 4 and Table 2. Due to the cyclic alternating applications of seismic loads and to knowledge needs how the staggered openings influence the failure mechanisms of the walls, at all models with openings were applied cyclic alternating horizontal forces. The methodology used for loading of these walls is the same as that used in the experimental tests by Wight and Aejaz [1]. It is based on controlling the horizontal displacements at the top of the experimental models and consists in recording of all the stresses and strains that occur in concrete and reinforcement in critical areas, for certain values of horizontal displacements. Seven values of horizontal displacements were chosen in Fig.5. These data were recorded and were made observations on cracking, distribution of cracks, displacements and stress in the concrete and reinforcement.

Fig.3. Reinforcement arrangement at experimental models

Fig. 4. Reinforcement percentage: a. vertical reinforcement;

b. horizontal reinforcement

Table 2. Reinforcement percentage of experimental walls

Vertical Reinforcement Horizontal Reinforcement Wall pv1[%] pv2[%] pv3[%] po1[%] po2[%] po3[%] po4[%] SW1 0,85 - - 1,01 - - -

SW23 1,56 2,15 0,31 0,79 1,61 0,31 0,79 SW45 1,57 1,57 0,31 0,79 1,61 0,31 0,79 SW67 2,15 1,40 0,31 0,79 1,61 0,31 0,79 SW8 1,07 0,31 - 0,79 1,61 - -

a b

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Fig. 5. Number of cycles and values of horizontal imposed displacements

Fig.6. The test bench

As shown in figure 6, the experimental stand was composed of two horizontal hydraulic

cylinders fixed on two sided metal frame. They were acted at the top of the experimental models and were able to develop horizontal forces up to 12 kN; and a vertical hydraulic cylinder fixed at the top of the wall which maintained a constant pressure of 50 kN. Alternating horizontal force was applied (in the East or West) through two hydraulic cylinders. There were used the following measuring devices: two displacement transducers - to measure horizontal movements of the models at the top and middle, a topographical total laser station - to measure horizontal displacements at each floor and the vertical at the foundation level, electrical resistively transducers placed on the concrete and reinforcement for recording specific strains. The physical state of the walls in ultimate limit states is presented in Fig. 7.

Fig. 7. Experimental models with door openings in ultimate limit state

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COMPARING THE RESULTS Scientific literature states that the ductility of a structure isn’t the clearest indicator of the

ability of the structure to adapt to seismic solicitations; hence the recommended calculus of absorbed energy of the structure is through post elastic deformations. Very little differences between the theoretical and experimental force-displacement P-� curves can be observed (Fig. 8). This is due to the way they were acted upon by seismic forces: the theoretical models were solicitated monotonically ascending while the experimental models cyclic alternating.

Fig. 8. Comparative P-� curves a. SW1; b. SW23; c. SW45; d. SW67; e. SW8

The capacity of the models to absorb the seismic energy was determined, by adding the areas

within the force-displacement diagrams, with the specification that the area between the last recording (which produces the crushing of concrete) and the displacement axis were not taken into account. Figure 9 shows the values of the dissipated energy function of the direction of the seismic action.

a b

c d

e

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Fig. 9. Comparative results of the dissipated seismic energy

Comparing the results of the dissipated seismic energy, obtained by theoretical and

experimental testing, the following conclusions can be made: (i) there are no major differences between the values of the dissipated seismic energy of the

theoretical and experimental models; (ii) the highest energy is dissipated by SW1 wall, followed by the models with staggered openings

SW23, SW45, SW67 and the wall with vertically ordered openings SW8; (iii) by positioning the openings towards the ends of walls and reducing the values of the angle,

the structural walls with staggered openings present a progressive decrease of the seismic dissipation capacity. The largest amount of energy in walls with staggered openings was dissipated in model SW23 ( = 45°);

(iv) all the experimental models have absorbed small amounts of energy by cracking of the concrete in the elastic domain. In the post elastic domain the largest amount of energy was consumed by yielding of the vertical reinforcement rebars from the post extremities;

(v) the experimental models SW8 and SW67 dissipate almost the same amount of energy. In the SW8 model this occurred due to the fact that the coupling beams didn’t have special reinforcement in order to avoid brittle failure due to shear forces, having the same amount of reinforcement in this area as the models with staggered openings. Model SW67 dissipates a reduced amount of energy due to the brittle failure induced by crushing of the concrete in the post at the base, before yielding of the vertical reinforcement. Increasing the dissipation capacity can be made by confining the concrete in the post from the base of the wall;

(vi) the energy dissipated by the experimental models has started to increase from a horizontal relative displacement value of 0.5%. But the coupled wall dissipated the same reduced amount of energy on each cycle of loading, after reaching a relative displacement of 0.5%, as a result of the fast degradation of the concrete at the ends of the coupling beams and yielding of the reinforcement in these zones, leading to a reduced number of recordings in the post elastic domain. Crushing of the concrete occurs at the base for 1,0% maximum relative horizontal displacements, while crushing of the concrete in the coupling beams is recorded at a maximum relative displacement of 0.6%;

(vii) the quantity of dissipated energy in the walls with staggered openings is influenced by the direction of the seismic action. Smaller differences between these values at the same wall are recorded for values of the angle between 32º and 18°.

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CONCLUSIONS Theoretical analysis and experimental test performed on structural reinforced concrete walls

with staggered openings show a high capacity of seismic energy dissipation, lacking special reinforcing detail for the same amount of reinforcement. The dissipated energy was evaluated based on the force-displacement curves obtained from numerical analysis on models subjected to monotonic loading and from test on experimental models subjected to alternating cycling loading.

The walls with staggered openings are able to dissipate almost the same amount of energy as the coupled walls with openings, if the plastic zones from the coupling beams ends aren’t appropriately reinforced for shear forces. The source for the energy dissipation is the vertical reinforcement from the wall extremities. It is necessary that the reduced sections from the base to be confined in order to avoid crushing of the concrete, buckling of the reinforcement and the horizontal reinforcement from the floor levels to remain in the elastic domain.

From the point of view of the positioning of the openings, the capacity of energy dissipation is reduced by placing the openings close to the wall extremities. From the studies performed until now, values higher than 45° are recommended for the angle for walls with staggered openings situated in high seismic areas with a high number of load-discharge cycles, where is necessary a high ductility for structural walls. High dissipation of the seismic energy for walls with staggered openings is due to a high post elastic deformation capacity.

REFERENCES 1. AEJAZ, A. and WIGHT, J. K. (1991), RC Structural walls with staggered door openings, Journal of

Structural Engineering, May, vol. 117, no. 5, pp. 1514 – 1531. 2. WALLACE, J. W. (1994), New Metodology for Seismic Design of RC Shear Walls, Journal of Structural

Eingineering, ASCE, V.120, No. 3. 3. MARSONO, A. K. and SUBEDI, N. K. (2000), Analysis Of Reinforced Concrete Shear Wall Structures

With Staggered Openings Part I: The Total Moment Concept, University of Dundee, Scotland, internet. 4. MARSONO, A. K. and SUBEDI, N. K. (2000), Analysis Of Reinforced Concrete Shear Wall Structures

With Staggered Openings Part II Non-Linear Finite Element Analysis (NLFEA), University of Dundee, internet.

5. HUI, W. and BING L. (2003), Parametric study of reinforced concrete walls with irregular openings, Pacific Conference on Earthquake Engineering, paper no.124.

6. BING, L. and QIN, C. (2011), Initial stiffness of reinforced concrete structural wall with irregular openings” Eartquake Engineering and Structural Dynamics, pp. 397 - 416, DOI 101002/eqe.946.

7. MO�OARC�, M. (2004), Contributii la calculul si alcatuirea peretilor structurali din beton armat, Doctoral thesis, ‘Politehnica” University Timi�oara.

8. MOSOARCA, M. and STOIAN, V. (2012), Modelling by theoretical and Experimental Analysis of RC Shear Walls with Staggered Openings Subjected to Seismic Actions. Reduction of rigidity, International Conference on Modeling and Simulation ICMS 2012, World Academy of Science and Technology, Issue 61, January 15-17, 2012, Zurich, Switzerland, pp. 687 - 697, pISSN 2010-376x; eISSN 2010-3778.

9. MOSOARCA, M. (2012), Failure Modeling using Simplified Computational Methods of RC Shear Walls with Staggered Openings Subjected to Seismic Actions, International Conference on Modeling and Simulation ICMS 2012, World Academy of Science anf Technology, Issue 61, January 15-17, 2012, Zurich, Switzerland, pp. 970 - 978, pISSN 2010-376x; eISSN 2010-3778.

10. STOIAN, V. and FRIEDRICH, R. (1992) BIOGRAF – Software for Nonlinear Biographical Analysis of Reinforced Concrete Elements in Plane Stress State, Civil Engineering Department, Politehnica University of Timi�oara, Romania.

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MODAL ANALYSIS WITH RESPONSE SPECTRUM FOR AN UNREINFORCED MASONRY STRUCTURE USING SEISMIC ROMANIAN

CODES

PEREDI �tefan*,CHIRA Alexandru, Technical University of Cluj-Napoca, *e-mail: [email protected] (corresponding adress)

A B S T R A C T This study presents several analysis regarding the seismic behaviour of the typical unreinforced masonry buildings from Romania. The study was conducted using a modal analysis with response spectrum with for all the romanian codes starting from 1963 until nowadays.The main goal of this study was to observe what changes were made by the seismic codes in terms of forces,tensions and displacements.

Keywords: modal analysis, response spectrum,unreinforced masonry,seismic codes


INTRODUCTION Over the years the seismic codes have changed taking into account several aspects like: new

seismic records, scientists results in the seismic area, evolution of computers and others. The safety of old structures that were designed using old seismic codes comes into question. In Romania many residential buildings of four story have an unreinforced masonry structure so their behaviour was interpreted using the results of the seismic analysis from each code . The building is considered to be situated in area with an horizontal acceleration of 0.24g. MATERIALS AND METHODS 1. Building model. Analysis, loads and materials

The analysis was conducted using Sap 2000 software using beam and shell elements. The typical unreinforced masonry building is presented in figure 1. The structure has four story with a height of 2.80 m and seven openings in the long direction with 3.60 m four of them and 2.60 in the middle area for three of them and in for the short direction the building has two openings of 5.75 m.

The structure consists of masonry walls with precast slabs. For this type of seismic analysis the response spectrum from all the seismic codes that were used are presented in figure 2.

Permanent loads were taken without the weight of the structural elements as it follows: - slab: 1.3 kN/m2; - roof slab :2.3 kN/m2; - exterior walls : 3.10 kN/m2; - partition walls : 0.5 kN/m2; Live loads were taken in respect to STAS 10101/OA-75 [1]: - rooms: 1.5 kN/m; - stairs: 3 kN/m2; For the seismic analysis were used all the seismic romanian codes starting from 1963: P13-63 [2], P13-70 [3], P100-78(81) [4,5], P100-92 [6], P100-2006 [7].

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Fig.1. The model of the structure

Fig.2. The seismic response spectrum of Romanian seismic design codes

The masonry properties are: - Young modulus E=2300 N/mm2; - Poisson’s coefficient �=0.2 ; - characteristic compression limit fk=2.3 N/mm2; The concrete used is C16/20 and his properties used in the analysis are:

- Young modulus E=27000 N/mm2; - Poisson’s coefficient �=0.2.

2. Results The results from the seismic analysis are briefly presented in the following figures.

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Fig.3. Shear base force [kN]

Fig.4. Story drifts [cm]

Fig.4. Compression stress [kN]

Fig.5. Tangential stress [kN]

CONCLUSIONS

The compression limit is exceeded in all the presented cases wich clearly indicates the need of retrofitting. Taking into account the specifications of the seismic code P100-2006 regarding the story drifts that indicates an admisible value of 0.005h( h being the story height), we obtain 1.4 cm wich is lower that almost the values calculated from the seismic analysis.

In conclusion the seismic capacity of the unreinforced masonry structures needs to be taken into account for a future earthquake and methods of rehabilitation to be considered.

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ACKNOWLEDGMENTS All the Romanian seismic codes were used for this study. Because the aim of the article was

to evaluate the results of the behaviour of a structure regarding the seismic codes no new references from the existing literature could be used.

REFERENCES 1. *** STAS 10101/2-75, Ac�iuni în construc�ii. Înc�rc�ri datorit� procesului de exploatare (STAS

10101/2-75 – Actions in constructions. Exploatation loads). 2. *** P13-63, Normativ conditionat pentru proiectarea constructiilor civile si industriale din regiuni

seismice (P13-63, Seismic code for civil and industrial buildings). 3. *** P13-70, Normativ condi�ionat pentru proiectarea construc�iilor civile �i industriale din regiuni

seismice (P13-70, Seismic code for civil and industrial buildings). 4. *** P100-78, Normativ privind proiectarea antiseismica a construc�iilor de locuin�e, social-culturale,

agrozootehnice �i industriale. (P100-78, Seismic code for civil, social,culture,agricultural and industrial buildings).

5. *** P100-81, Normativ privind proiectarea antiseismica a construc�iilor de locuin�e, social-culturale, agrozootehnice �i industriale (P100-81, Seismic code for civil, social,culture,agricultural and industrial buildings).

6. *** P100-92. Normativ privind proiectarea antiseismica a construc�iilor de locuin�e, social-culturale, agrozootehnice �i industriale (P100-92, Seismic code for civil, social,culture,agricultural and industrial buildings).

7. *** P100/1-2006. Cod de proiectare seismic� - Partea I: Prevederi de proiectare pentru cl�diri (P100/1-2006, Part I: Building design elements).

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SHEAR CAPACITY FOR PRESTRESSED-PREFABRICATED HOLLOW CORE CONCRETE SLABS,WITHOUT SHEAR REINFORCEMENT

PINTEA Augustin,

Technical University of Cluj-Napoca, e-mail: [email protected]

A B S T R A C T Eurocode 2 presents a design method for formulating the capacity against shear effort, a method that standard EN 1168:2005 + A3:2011 puts to use for preventing web shear failure of hollow core prestressed-prefabricated concrete slabs. But this method has the drawback of ignoring the shear efforts owing to the transfer of the prestressing force. The results gathered from testing FGP 200 and FGP 320 prefabricated hollow core slabs indicates that this method should not be used when designing prestressed-prefabricated hollow core slabs that lack shear reinforcement, due to the large overestimating nature of this practice when considering web shear resistance.

Keywords: Eurocode 2, shear capacity, pretensioning, critical point Received: January 2012 Accepted: January 2012 Revised: March 2012 Available online: May 2012

INTRODUCTION

As mentioned in European Standard EN 1168:2005 + A3:2011 [1], adopted by CT321-Concrete and prefabricated concrete products technical committee to be the equivalent romanian standard; these type of slabs can be used as structural elements within buildings or other types of civil engineering jobs,excepting bridges. In case of structural elements for buildings, they have perfect usability as floors, roofs, and also type F and G vehicle zones which are not subjected to strain stress and are in accordance with EN 1991-1-1 standard.The certification process applied for assessing the conformity of prestressed-prefabricated hollow core concrete slabs, regarding essential characteristics, 2 +, states that: For the intended use of these slabs, the certification process has to be based on the conformity evaluation procedure resulted from the application of articles within this European standard. This procedure forsees that, the resistance to shear failure obtained from calculus has to be confirmed afterwards by physical testing of full scale models conformable to article 4.3.3.3, and in accordance with standard appendix J. Article 6.2.2. – Initial type testing try-outs, specifies the following observation; to confirm the good working order of production equipment, the criteria checking from J5, needs the calculation of shear stress capacity, indifferent of the presence or absence of mechanical resistance properties declaration by the manufacturer that intends to introduce these products to market.


The mechanical resistance of hollow core prestressed-prefabricated concrete slabs can only be checked at this stage of European standardisation process by using the calculus method; nevertheless a good working order of the production equipment is imperative, because the concrete properties used as entry data for the calculation of the resistance to shear failure depends on it.

Due to certain product specifications (the lack of transversal reinforcement) , a few additional calculus rules are needed to EN 1992-1-1: 2004[2] method. On top of that, researching done on extruded hollow core prestressed-prefabricated concrete slabs led to specific calculus rules, used on a large scale, but not yet integrated in EN 1992-1-1:2004[2].

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1. Shear Capacity For extruded hollow core prestressed-prefabricated concrete slabs without web shear

reinforcement, the resistance to shearing stress of cracked sections resulted from slab deflection, are calculated based on the following expressions (6.2a) and (6.2b) taken from EN 1992-1-1:2004[2]:

( ) dbkfkCV wcpckcRdcRd ��

��

� += σρ 131

1,, 100 (1)

with a minimum value of: dbkVV wcpcRd )( 1min, σ+= (2)

In the uncracked areas after the bending force is applied (where tensile strength resulted from

deflection is smaller than c

ctkfγ05,0 ), the shearing force capacity is limited through the tensile

strength of concrete. In these areas, the shearing force is calculated using the following expression (6.4.) from EN 1992-1-1:2004 [2]:

( ) ctdcpctdw

cRd ffS

IbV σα1

2, += (3)

But, expression (3) has the drawback of not taking in consideration the shearing stress resulted from transfer of prestressing force. Such a stress cannot be ignored and for this reason the figure presented below ilustrates its effect. If there is no contact between A and B:

a) before release b) after release

Fig.1. Detaching manner for the two pieces of a hollow core slab (bottom and upper pieces), when shearing stress is missing at the release of the prestressing force [9]

The bottom part of the hollow core slab tends to contract at the release of the prestressing

force; and because the bottom piece of the slab is connected to the upper one, there has to be a shearing stress which keeps them together. This means that the design method for shearing stress capacity presented in Eurocode 2, (which deals with web shear failure prevention due to applied shear stress on the prestressed-prefabricated hollow core concrete slabs without a shear reinforcement) is ignoring the shearing stress at the release of the prestressing force. However, in the web of a prestressed hollow core slab the nature of stress is essentially two-dimensional; and because the compressive principal stress being relatively small, its effect on the transversal tensile strength is also small. Due to a short design length, lack of shear reinforcement and absence of stiffening to prevent flexural strain; in these prestressed-prefabricated hollow core concrete slabs, a diagonal shear crack in the web close to the support zone principially needs a failure.

Thereby a failure criterion is formulated like this: ctI f=σ (4) where: Iσ is obtained from the following expression:

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22

22τσσσ +�

�

��

�+=I (5)

The vertical normal stress resulted from support pressure is taken into consideration in such a way, that the applied failure condition is not to close to the support area, where vertical stress component is effective and reduces Iσ . When choosing the failure criterion, beside the maximum principal stress, we have other principal types of stresses too, so the question about calculating the maximum principal stress arises. Two-dimensional elastic linear analyses entails knowing how the transfer of prestressing force is done from the tendons to the concrete. 1.1. Shear capacity, according to CP 110

The British Code of Practice CP110, [6], recommends the use of following expression, when calculating the shear resistance of a web:

ctcpctw

wwc ff

SIb

V σ8,02 −= (6)

AP

cp

−=σ (7)

1.2. Shear capacity according to Walraven and Mercx Walraven and Mercx proposed a similar formula for calculating web shear resistance, their

work entitled “The bearing capacity of prestressed hollow core slabs” [7] addresses this subject in more detail:

ctcpctw

c ffS

IbV σ−= 275,0 (8)

with 0,75 being the calibration factor. The difference between these two approaches, is that I and S are calculated for the whole

transversal cross-section, taking the prestressing force at the inner edge of the support.

�=

−−��

�

��

��

� −−+=n

tc

Edxt

tcc

icp yY

lM

lPl

YpYyYA

y1

)()())((1

)(σ (9)

- possitive if there is compression, while Y is the height of the critical point situated on the fracture line.

Eurocode 2 has adopted expression (8) under the following form:

ctcpctw

cRd ffS

IbV σα1

2, )( += (10)

Expression (10) is equivalent with expressions (4) and (5), when the shearing effort in the considered point , cpτ is calculated this way:

VI

S

bcp

wcp

1=τ (11)

According to this expression the maximum shearing effort, and as a result the maximum principal stress have the highest values when:

w

cp

b

S has the maximum value

In the case of hollow core slabs with rounded or oval inner voids, the maximum principal stress is obtained at the centroidal axis of the transversal cross-section, or very near to it.

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1.3. Shear capacity according to Yang In the case of one tendon layer, Yang [8] proposed the following expression for the calculus

of shearing stress:

��

��

�+��

�

��

�−= V

I

S

dxdP

I

eS

A

A

bcpcpcp

w

1τ (12)

The first term of the afore mentioned expression:

��

��

��

��

�−=

dxdP

I

eS

A

A

bcpcp

wt

1τ (13)

it’s attributable to the transfer of prestressing force. And if this term leads to the expression (11), then expression (13) suggest the fact that the maximum shear stress doesn’t necessarily have the positioning in the vicinity of centroided axis.

The maximum principal stress may be positioned in other places on the slab. As a consequence the horizontal normal stress, for one tendon layer slabs, is calculated as follows:

zI

MPeAP +−+−=σ (14)

According to SR EN 1168:2005 + A3:2011 standard, the critical point is situated on a straight line that forms a o35=β degree angle with the horizontal axis. This line has its origin at the edge of the support flange and cRdV , expression has the lowest value.

Concerning the FGP 320 hollow core slabs(with non circular voids), the critical point was situated a little lower than the point of intersection between the flat web and the inferior support flange.

Fig.2. Location of critical point for FGP 320 slab

Geometrical symbols for the FGP 320 slab are illustrated in next figure:

Fig.3. The illustration of geometrical parameters of the considered

transversal cross-section for the FGP 320 slab

The effective web width depends on cpz and it is obtained from:

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�=i

cpwiw zbb )( (15)

RESULTS

The slab try-outs were executed in order to obtain experimental data,which will be compared afterwards with the data from design stage. All this effort is done in order to certify tested specimens (by certain certification organizations) to become fully legal and market available products. During these try-outs it has been carefully observed the way in which elements behave at limit states like: SLEN, SLR, SLU. Within those limit states, the moment when the first cracks appear, the closure and reopening of these cracks, the equivalent deflection and the manner of breaking (slippage and breaking of inner strands/tendons or because of web failure due to crushing of concrete) are of critical importance to this study. Analysing the resulted experimental data, the following observations have been reported:

- under working loads, the deflection at the middle of the span, has smaller values than those accepted;

- after unloading the slabs, the reopening of initial cracks only occur after adequate loading at resistant limit state; because under working loads the cracks remained closed;

- under working loads, the slabs showed no evidence of fissurations; the cracks only appeared after exposing the slabs to greater load values than those taken from calculus stage;

- the critical point was situated a little lower than the point of intersection between the flat web and the inferior support flange(where the web has the minimum thickness).

In Romania, an extruded hollow core prestressed-prefabricated concrete slabs production unit

was homologated and put into service, at S.C ASA CONS ROMANIA S.R.L. TURDA.

Fig.4. Manufacturing stages for the extruded FGP 200 and FGP 320 slabs

(Photo: Pintea Augustin, 2011)

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CONCLUSIONS Equation (6.4) from Eurocode 2 presents a method for designing a slab against web shear

failure, but just for web elements that don’t have a shear reinforcement. This method is also utilised for hollow core prestressed-prefabricated concrete slabs lacking transversal reinforcement.

Based on the results from testing,Yang’s design method for web shearing capacity of hollow core prestressed-prefabricated concrete slabs is considerable better fitting testing results, compared to Eurocode 2 method. Due to this substantially improved method of designing a slab against web shear failure, and because of its accuracy, Yang’s method should altogether replace Eurocode 2 method. Eurocode 2 method should never be used without a reduction factor in the case of hollow core slabs with flat webs, and its applicability on other types of hollow core slabs should be checked vigorously before using them, either numerically, or experimentally.

REFERENCES 1. *** SR EN 1168:2005 + A3:2011 – Prefabricated concrete products. Hollow core slabs. 2. *** SR EN 1992-1-1:2004 – Eurocode 2: Design of concrete structures Part 1-1: General rules and rules

for buildings. 3. *** SR EN 1992-1-1:2004/AC:2008 – Eurocode 2: Design of concrete structures Part 1-1: General

rules and rules for buildings. 4. *** SR EN 1992-1-1:2004/NB:2008 – Eurocode 2: Design of concrete structures Part 1-1: General rules

and rules for buildings national attachment. 5. *** SR EN 1992-1-1:2004/NB:2008/A91:2009 – Eurocode 2: Design of concrete structures Part 1-1:

General rules and rules for buildings national attachment. 6. *** CP 110 – The British Code of Practice. 7. WALRAVEN, J.C. & MERCX, W.P.M. (1983), The bearing capacity of prestressed hollow core labs.

Heron. 8. YANG, L. (1994), Design of Prestressed hollow core Slabs with Reference to Web Shear Failure. ASCE

Journal of Structural Engineering. 9. MATTI PAJARI (2005), Resistance of prestressed hollow core slabs against web shear failure. Espoo.

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MANAGEMENT OF STRUCTURAL MONITORING USING OPTICAL FIBERS

R�DULESCU Adrian T. G., R�DULESCU Gheorghe M.T.*,


A B S T R A C T Virtually everything is made by man at investment level: construction and means of communications, equipment, transportation but also the natural elements of the land related to relief, must be observed over time. In both cases we must ensure, through monitoring the activity in time that their evolution cannot lead to dangers to human life. For those structures mentioned, the results from a design activity and observation of behavior over time, not only are regulated and legally required, but are also a challenge for designers, the monitoring results being able to validate, or not, the design solutions. This paper presents the conceptual elements of the management of structural monitoring using optical fibers. Although the model is designed for construction, there are elements of adaptability to any structure of the kind mentioned.

Keywords: monitoring the activity in time, optical fibers, environmental conditions


INTRODUCTION

Development of completion of special constructions of more and more performing dimensions and constructive features, experienced an unprecedented expansion in history, during the last 30 years. Against this background, the monitoring activity of the time behavior of the constructions became a stage in the life of these structures, started since the completion of foundations, and running their entire life cycle [1].

Once in use, each structure is subject to evolutionary patterns of loads and other actions. The sum of these uncertainties created during design, construction and use is a big challenge for those involved and responsible for the safety, maintenance and operation of the structure. The monitoring of the structural condition aims to provide more solid information about the real state of a structure, observing its evolution while detecting new degradation. Monitoring is a new security and management tool which complements the traditional models ideally, such as visual inspection, conventional methods of tracking and visual modeling [6].

MATERIALS AND METHODS 1. New structures, construction and testing, placing sensors in the life of structures Optical fiber sensors have important advantages compared to traditional measurement methods, including low cost, versatility in measuring various parameters, insensitivity to electromagnetic influences and corrosion, their small size and high density of information than they can carry far away. The inclusion of sensors in the construction tracked led to the concept of “smart structure” [4]. Its application to specific needs of civil engineering opens new perspectives in the long-term monitoring of all major construction works like structures, tunnels, dams, airport runways, domes but also unstable soils and rocks, and last but not least large industrial structures: machinery, airplanes, trains, ships, cranes, etc.

2. New monitoring needs, replacement or improvement of conventional instrumentation

The new monitoring needs in engineering can be subdivided into two broad categories [1]:

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1. Some old methods of monitoring can be replaced with modern equipment, which responds better to the needs of today;

2. New techniques can achieve measurements in structures that traditionally lack adequate monitoring systems.

Many monitoring systems used now, considered traditional and classic, do not satisfy user needs. In the case of conventional strain sensors, some of the disadvantages mentioned by experts include:

1. Difficulty of use, requiring specialized operators. They are slow and inefficient; 2. Difficult or impossible automatic/remote measurements; 3. Calibration or recalibration are necessary operations; 4. Sensitivity to temperature, humidity and other environmental variations; 5. Sensitivity to electromagnetic fields caused by lightning storms, railways or power

lines and stray currents; 6. Susceptibility to corrosion; 7. Large size; 8. High operational costs, including basis cost or for pre-measurements and maintenance.

Any monitoring system must solve at least some of these problems to achieve replacement of existing equipment. In this context, techniques and instruments that are considered classics have been partially replaced, appearing new methods, new equipment and also genuine monitoring manager systems of land and constructions [6]. 3. Validation instrumentation

Some structures and materials are not sufficiently monitored because there is no appropriate measurement system [4]. Some applications that could benefit from a new monitoring system (based on optical fiber sensors) include:

• Monitoring of concrete structures; • Geometric monitoring in structures. Many structures such as buildings, towers, special

constructions like silos, tanks, etc. and others can be monitored in terms of geometry, by measuring the distance between fixed points and the structure;

• Monitoring with large temperature variations. Structures such as bridges, high buildings, reservoirs, boilers, tanks or space farms may suffer large and spontaneous temperature variations;

• Measurements with curved shapes as traditional sensors do not allow measurement of curved surfaces;

• Measurements of industrial nature, identifying construction defects, determining the effects of oscillations and vibrations on the functional capacities.

The sensor is part of the measurement system which is installed in or on the structure and transforms the movement in a change in imbalance between the fibers and the route [2]. Because different types of structures and different materials require specific sensors, this subsystem is to be adapted to most particular applications [3]. The sensor must meet different requirements in terms of optical and mechanical soundness and transmission of movement from the structure to the fiber [4].

• Optical requirements. The sensor must encode the structure’s displacement in a change of state of one of the measurement fibers. This is possible by using a interferometer which converts the mechanical phenomenon into an optical one.

• Mechanical requirements The measurement fiber must be in mechanical contact with the host structure. All axial movement must be transferred from the host structure to the fiber.

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• Environmental requirements. The sensor must survive the construction and, if possible, the entire lifespan of the structure. During the construction phase, the sensor is exposed to a hostile environment and thus must be protected from external agents.

• Financial requirements. As the number of sensors needed to monitor a large structure, like a bridge, can be tens or even hundreds, the price of each sensor should be as small as possible.

4. Fiber optic sensors; quantities measured by optical methods

The six forms of exploitable energy in the sensor field determine the quantities categories measured by optical methods, as follows: mechanical quantities (displacement, rotation, force, speed, acceleration, effort, pressure, flow, vibration, acoustic field); electrical quantities; magnetic quantities (magnetic field); thermal quantities (temperature); chemical quantities (pH, chemical species, humidity); radiant type quantities (optical radiation). Optical fiber [2] is a cylindrical dielectric waveguide, which consists of a central region called the core whose refractive index is greater than the one of the dielectric material surrounding the core and forming fiber’s casing. The core and the casing are surrounded by a shell with a protective role against external forces and provide a good mechanical resistance. In the field of optic fiber sensors there is, at present, a large amount of information, because this area has gained a great expansion. Findings have been uncoordinated, and the result is a mosaic of fiber optic sensor solutions of the most diverse, for applications just as diverse. Systematization of the field is now a necessity. Defining meaningful classification criteria for all specialists connected with the optic fiber sensors can help develop it faster [3].

5. Intelligent detection with optical fiber, part of optical metrology

Although the concepts behind intelligent detection are not necessarily related to optical methods, detection is usually implemented in conjunction with fiber optic sensors. Intelligent detection is also linked to structural monitoring and normally is considered the first construction of small structure; the others are processing and acting [5]. This field is so young that the community for “smart structures” has yet to find a generally accepted definition for the term “intelligent detection”. Intelligent detection can be seen as a combination of technologies (sensors, transporters of information, information processors, interfaces) allowing sensitive structures to be executed. Combining these detection capabilities with a series of ad hoc actuators it could be possible to create a structure with auto-repair, with shape control or vibration damping capacity for an intelligent structure. Sensors used to monitor different parameters needed to quantify the status of a given structure could be of any type [4, 5]. Other technologies, such as electrical and electromechanical sensors (force, position, angle, acceleration, temperature) or special systems like GPS can be used in addition with FOS to convey information about the structure and its environment. Even if the structure is a singular case, some key decisions are common to most applications and are summarized in the following paragraphs. Movement and tension are the most important parameters to be monitored in a structure and a great variety of sensors have been designed for this purpose, requiring special attention in their choice. Measurements of displacement, strain and tension are thus the first step in the design of intelligent sensing structures in the analysis of parameters that need monitoring in choosing the best technology for detecting and assessing the number and positions of measurement points required [5]. Other types of sensors include temperature, pressure and chemical sensors. They are the most interesting parameters monitored in the vast majority of structures. Tension, strain and displacement sensors (sometimes based on the same technology) can be used together for complex structures to get a

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complete understanding of their global and local behavior. Choosing a sensory technology usually leads to an estimate of the number and placement of sensors needed to monitor a certain structure.

6. Sensor selection

A measure of absolute tension gives a relative value to the non-tensioned status of materials and is thus useful for determining the load status of the structure. Most sensors will provide a value that is relative to the one measured at the time the sensor was installed on the host material [4]. Once the type of measurement required to monitor a structure was established, it is necessary to quantify the values to be measured. We distinguish three categories of measurements resolved in time [6]:

• Dynamic measurements require readings every second, even millisecond. They are usually related to vibration measurements in a resonative condition;

• Short-term measurements (quasi-static) can extend from a few seconds to a week; • Long-term measurements (static) require stable sensory techniques that guarantee

sufficient precision for readings that can be extended from one month to years or even decades.

As a general rule, any sensor is also a temperature sensor. Simultaneous measurement of tension and temperature usually consists of a sensor that responds only to variations in temperature and one that is sensitive to tension and temperature in a linear way.

Fig.1. Subsystem with smart sensors and sample of implementation for intelligent bridges [4]

Sub-system sensor

Sub-system transporter information

Reads the sub-system unit

Sub-system processing

Sub-system interface

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7. Selection of sensory technology Selection of measurement techniques, to be developed in a research project, must follow two

main objectives [6]. 1. A new system must meet the real needs of end users and, 2. Be an innovative approach in metrology showing originality compared with the results of

other active research laboratories with the same concerns. Requirements of a strain sensor for short or long term monitoring (not dynamic) show that

interferometry with low coherence to optical fiber sensors meets all these requirements [6]: • Detection of strain - a strain sensor has to measure the variation between the sensors of two

fixed points on the structure. When possible, the sensor should be embedded in the building materials to provide more representative measurements for the behavior of the structure when compared with surface mounted sensors.

• The length of the sensor - because of the variety of structures found in civil engineering it is impossible to find a standard length for sensors to suit all applications. A sensor dedicated to all goals should allow measurements from a few inches long, up to a hundred meters or more. The active region can be sometimes up to several kilometers away from the reading unit.

• Resolution and accuracy – the needs of resolution vary with every application. If the sensors are considered to be the replacement of conventional techniques, such as dial comparators, it is important to ensure at least the same resolution. An accuracy of 1% of measured strain should be considered sufficient.

• Stability - Since the aim is to achieve long-term applications, the resolution and accuracy cited above should remain valid even for measurements with an interval of several years between them.

• Detection of temperature - All strain sensors are, at some level, temperature sensors. It is interesting to study the influence of temperature in strain measurements. This helps define the type of compensation of the suitable temperature for a given application. Any sensor will turn a strain into a change of a certain X quantity.

CONCLUSIONS

Monitoring new and existing structures can be approached in terms of material or structural. • In the first case, monitoring will focus on local properties of the materials used in the

construction (concrete, steel, wood), and it will be observed the behavior under load or the aging effect. If large number of sensors were installed in different parts of the structure, it would be possible to extrapolate information about the behavior of the whole structure, using these local measurements.

• In the structural approach, the monitored element is geometrically observed. Using the tracked length strain sensors, with bases measuring from one meter to several meters, it is possible to gather information about the strain of the structure as a whole, and to extrapolate the overall behavior of construction materials.

REFERENCES 1. CHRZANOWSKI A., SZOSTAK-CHRZANOWSKI A., BOND J. (2007), Increasing public and

environmental safety through integrated monitoring and analysis of structural and ground deformations, in Geomatics Solutions for Disaster Management (eds: J. Li, S. Zlatanova, A. Fabbri) Springer, pp. 407- 426.

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2. CULSHAW B., DAKIN J. (1984), Optical fiber sensors. Systems and application,vol. 2, Artech House, London.

3. ERIC UDD. (ed.) (1993), Fiber Optic Sensor – An introduction for Engineering Mechanics, American Society of Civil Engineering, New York.

4. GLIŠI�, B., INAUDI, D. V. (2007), Fibre Optic Methods for Structural Health Monitoring, John Wiley & Sons, Inc., Chichester (ISBN: 978-0-470-06142-8), 2007 (Europe) / 2008 (USA).

5. INAUDI D. (1998), Fiber Optic Sensor Network for the monitoring of civil engineering Structures, presentee au departement de genie civil Ecole Polytechnique Federale de Lausanne pour l’obtention du grade de docteur es sciences techniques, E.P.F. Lausanne.

6. R�DULESCU A.T.G.M. (2011), Tehnologii topografice moderne utilizate la urm�rirea comport�rii în timp a construc�iilor situate în perimetrele miniere (Modern surveying technologies used for tracking the time behavior of constructions within mining perimeters), Phd Thesys, University of Petro�ani.

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THE MONITORING OF BESKA DANUBE BRIDGE, NOVI SAD, SERBIA, IN A PERMANENT QUASI-STATIC REGIME

R�DULESCU Adrian T. G., R�DULESCU Gheorghe M.T.*,

Technical University of Cluj Napoca, *e-mail: [email protected] (corresponding adress)

A B S T R A C T The analysis of a building’s construction must also include a good knowing of the characteristic environmental factors in the location area, but in order to optimize the design solutions, the behavior models provided by the data banks (future) become a must. In setting up these data banks the main information is provided by the topo-geodesical activity, which must coordinate the entire monitoring process for the tracking of the behavior after the action of environmental and exploitation factors. The case study, in continuous quasi-static condition, was performed on Beska Danube Bridge, located near Novi Sad, Serbia. The tracking period started on January 9, 2009, on November 14, 2011 the summary results were uploaded at the company' Vienna Consulting Engineers Company in Vienna.

Keywords: general behavior models, optical fibers, Beska Danube Bridge, topo-geodesical activity


INTRODUCTION Security of the civil engineering works requires regular monitoring of the structures [1]. The current methods are often difficult applications, the resulting complexity, dependency from the condition of the atmosphere, and also the costs, limiting the applicability of these measurements. Special attention is therefore focussed on maintaining them in a serviceable condition. The problem is quite complicate as it is function of their age, variety of structural types, different processes of deterioration and increasing volume and composition of traffic. Mostly developed in the last 10-15 years, this type of approach even not common practice, has been and is used on both new and existing structures to keep under control structures of strategic importance or very deteriorated structures whose critical conditions may require continuous attention [3].The emergence of new methods and technologies for structural monitoring has occurred slowly, until two decades ago, while developing new methods, tools and also conventional techniques, nowadays the information explode, appearing practically endless combinations in shaping the time behavior monitoring of an objective [2]. Through this work we proposed, firstly, to fill this informational gap, to seek and know the most relevant achievements in the field, to systemize, categorize and then present them in this thesis. For it we had to reconsider the monitoring activity of the land and constructions, launching the concept of monitoring in three segments (static, quasistatic-quasidynamic, dynamic) [4], relative to the structure response speed to stress. The methods, equipment and monitoring systems for circumstances of continuous taking-over of data are generally valid, only the nature and frequency of the loadings making a distinction between them [5]. As a result, the measurements regarding the case studies for continuous quasi-static were performed inside the VCE-Vienna Consulting Engineers company in Austri a[9], with the headquarters in Vienna, one of the largest companies in the field in the world, which not only has the required equipment but has contracts in the filed which are currently unfolding. Thus, the case study for continuous quasi-static regimen was performed on the Beska Danube Bridge, located nearby Novi Sad, Serbia [4].

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MATERIALS AND METHODS 2. The monitoring of Beska Danube Bridge, Novi Sad

The case study for continuous quasi-static regimen was performed on the Beska Danube Bridge [9], located nearby Novi Sad, Serbia, 45,17° latitude, 20,08° longitude. In the course of the completion of the E75 (part of the European Corridor 10) on 4 lanes the construction of a new bridge over the Danube is required between Novi Sad and Belgrade near Bezka. This bridge will be erected as a so-called "twin"- bridge parallel to the existing bridge, which currently accommodates two lanes in both directions and is to have the same appearance. The Danube is bridged with spans of 60m+105m+210m+105m+60m (without foreland structures). The old bridge, built as prestressed concrete bridge in 1975, is to be maintained and continued to use. November 2008 a BRIMOS© measurement for the assessment of the condition was carried out at this structure. In addition the effects of the existing sliding slope on the structure were analyzed. In January 2009 a BRIMOS© monitoring system for permanent supervision of the bridge pier movements was installed additionally in order to be able to identify negative effects of the foundation works for the new construction on the old structure on time. Furthermore movements of the sliding slope are to be detected by means of the measuring system. The measuring system includes a tachymeter for continuous monitoring of the pier movements as well as acceleration sensors for vibration supervision during sheet pile and bored pile works.

The tracking period started on January 9, 2009, on June 14, 2010 the summary results were uploaded at the company's headquarters in Vienna VCE. The measurements were made for the prism sensors installed on the bridge’s piers, as follows: Pier 42 Up, Pier 43 Up, Pier 43 Down left, Pier 43 Down right, Pier 44 Up, Pier 44 Down, Pier 45 Up, Pier 45 Down, Pier 46 Up, Pier 46 Down, Pier 47 Up, Pier 47 Down, Pier 48 Up, and Pier 48 Down. Special measurements have been made for the following points: Pier 42 N42NLG, Pier 42 N42NDG, Pier 42 N42ULD, Pier 42 N42UDD, Pier 43 N43NLG, Pier 43 N43NDG, Pier 43 N43ULD, and Pier 43 N43UDD. The results are presented in diagrams, which can be reviewed either on diverse intervals, or along all the recordings, annually, monthly, weekly, daily, hourly. Next, we present examples of these recordings, more examples in the addenda, because of the manner through which VCE provides the results, the example are infinite, one being able to select infinite time intervals for these recordings.

Fig. 1. Beska Danube Bridge, Novi Sad, Serbia

Fig. 2. Old Bridge, structural details of the resistance monitored

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Fig. 3. a. Prism reflector mounted on the structure in the 14 tracked points Sensor, Leica GRZ4 (3-D Dimensional), b. Permanent total station mounted outside the structure, which follows the

14 mobile points

Fig. 4. Print screen of the recordings, Vienna Consulting Engineer Site, Brimos,

Bridge Monitoring System

a

b

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Fig. 5. The situation of the meteorological recordings, for the all monitoring period 14.04.2010-25.10.2011

Fig. 6. The situation of the events,

for the all monitoring period 14.04.2010-25.10.2011

Fig. 7. The behavior under pressure of pier 42NDG, for the all monitoring period 14.04.2010-25.10.2011

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Fig. 8. The behavior under pressure of pier 42NDG,

in the interval 25.09.2011-25.10.2011, the last month of the monitoring


in the interval 18.10.2011-25.10.2011, the last week of the monitoring

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in the interval 24.10.2011-25.10.2011, the last day of the monitoring

Fig. 11. Print screen of the all recordings 1, BriMoS

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Fig. 12. Print screen of BriMoS site, Beska Danube Bridge all all records were completed on 25.10.2011, the monitoring activity is “completed”

CONCLUSIONS Regarding the use of classic methods, in which the studied phenomena are secquentially put

into evidence, the following conclusions can be stated after the studies made [6], [4]: 1. The information regarding the behavior in situations of similar constructions (compatible)

are extremely useful, but unfortunately the results of the measurements are seldomly communicated and so, for now, it is arduous to build a data bank in this field.

2. Aware of behavior model of construction A, with the structural parameters B, located in the C area, characterized by D environment factors, makes the calculation of the worst case possibilities of combining the influence factors of the monitored construction’s behavior possible and the taking-up, within the maximum resistance limits provisioned, of the optimum solution for designing a similar future project.

3. Between sub-sizing and over-sizing, the design variant varies according to the information had regarding the structure’s behavior in real conditions and according to the standard fundamentation degree used, which ultimately is based on this information when configuretion is made.

REFERENCES 1. CHRZANOWSKI, A., SZOSTAK-CHRZANOWSKI, A., BOND, J. (2007), Increasing public and

environmental safety through integrated monitoring and analysis of structural and ground deformations, in Geomatics Solutions for Disaster Management (eds: J. Li, S. Zlatanova, A. Fabbri) Springer, pp. 407- 426.

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2. GLIŠIC, B. INAUDI, D.&VURPILLOT S. (2002), Structural Monitoring of Concrete Structures Procedeengs of Third World Conference on Structural Control, 7-12.4.2002, Como, Italy, pp. 1-10.

3. INAUDI ,D.(1998), Fiber Optic Sensor Network for the monitoring of civil engineering Structures, presentee au departement de genie civil Ecole Polytechnique Federale de Lausanne pour l’obtention du grade de docteur es sciences techniques, E.P.F. Lausanne.

4. RADULESCU, A.T.G. (2011), Modern surveying technologies used for tracking the time behavior of constructions within mining perimeters, PHD Thesys, University of Petrosani, Faculty of Mines, its support 21.01.2011, Scientific Coordinator, Prof. univ. dr. ing. Nicolae DIMA.

5. R�DULESCU, A.T., R�DULESCU, GH.M.T, STEFAN, O. (2005), Execution Analysis and Time Behaviour Monitoring of the Highest Building from Romania for Establishing the Influences of Settling on its VerticalityNinth International Conference on Structural Studies, Repairs and Maintenance of Heritage Architecture-Stremah 2005, Malta 22-24.06.2005.

6. R�DULESCU, A.T.G., R�DULESCU, GH.M.T., R�DULESCU, M.V.T. (2010), Geometric Structural Monitoring in Cinematic Regime- dynamic Surveying as Means to Assure a Structure Safety, PAPER (3945), FIG Congress 2010 – Facing the Challenges – Building the Capacity, Sydney, Australia, 11-16 April.

7. * * * http://www.softrock.com.au/ 8. * * * http://www.Smartec.ch 9. *** Brimos, Bridge Monitoring System, Vienna Consulting Engineer, http://www.brimos.com/DMA/

DMAFrames.aspx, accesed between 2009-2012.

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MGIS (MINING GEOGRAPHICAL INFORMATION SYSTEM) A NEW CONCEPT FOR THE INFORMATIZATION MANAGEMENT ON MINING

COMPANIES. SOME CONSIDERATIONS ON THE MGIS FIELD

R�DULESCU Virgil Mihai G. M., R�DULESCU Corina, Technical University of Cluj Napoca, e-mail: [email protected]

A B S T R A C T MGIS (Mining Data Bank Geographical Information System) is a new concept for the informatization of mining activities by creating a mining database combined with a GIS platform. MGIS includes activities from both the inside of mining activity and from outside it, referring to either the environmental impact on mining but also to the influence of mining on the environment. MGIS is also launched in a Web platform, which is www.mdbgis.ro which will be loaded and updated with data related to the concept and its components. The basic idea of the new concept is geo-referencing all data introduced in the system, starting from finding that any information happened somewhere (in a 3D space) and sometime (T-time component). MGIS is a modular system which, once developed and implemented in an organization, can be supplemented with other modules, databases, software, actors, links, information and decisions.

Keywords: MGIS (Mining Geographical Information System ), mining database combined with a GIS platform


INTRODUCTION

Currently, no industrial sector can progress without a management system based increasingly on the informatization of business and its management and coordination. As the leader of the world’s economy, the extractive mining industry cannot be an exception to this rule in any part of it, from the exploration phase to the delivery of extracted materials to processing industries [5]. One can see that the largest mining companies use specialized software for certain activities, that very few companies have implemented a GIS system, including only some activities, that there is mining software, some very powerful and widely recognized by major companies in the industry and that, at conceptual level, large GIS software manufacturers are prepared to enter the mining market, to the extent that it is ready for major changes in organization management.

MATERIALS AND METHODS 1. Short analysis of the current computerization state of mining activities

The most complete mining information system was designed by IBM (Indian Bureau of Mines) at the order of the Indian government, originally entitled Mineral Information System (MIS) [8], supplemented subsequently and renamed Technical Management & Information System (TMIS), which contains a number of databases among which also GIS. Niche information systems in the mining field or in complementary fields were created and work very well, for example in the geological field GeoGRAFX GDMS [9], in the mining field Mining Information System, created by Trimble. Currently, in mining companies in Australia and Canada, the most computerized organizations in the industry, over 60% of information is analog, while in mines in Romania, for the activity still surviving, the rate is much higher, more than 90-95%.

All these data must be digitized, and here is where the material and time costs to create a mining databases, with GIS, will be, as the main operating axis or not. Each company will decide the form and time frame through which this expensive information operation shall be activated and

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solved. In this case, this paper cannot provide solutions, but simply suggest instruments, because the aim is to create a new concept for mining computerization, the so-called MGIS [3].

2. MGIS, Mining Geographical Information System – some considerations

MGIS history, or its roots, can be assimilated with GGIS history and its achievements, GIS geo-sciences through the contributions of Burroughs 1987, Aranof 1989, Tolin 1990, Maguire 1991, Carter 1994, or with the first ESRI application in geo-sciences in 1993 [2]. As a general theme, name and specific concerns, the field is very new, probably somewhere between 2004 and 2006. Based on the few achievements to date in MGIS, we shall present some considerations on the field.

1. Support system in underground mining [4], it has three components that must be considered when setting MGIS:

a. Surface support system, b. Underground support system, c. Relation between the surface and the underground support system.

The purpose of the third component is to ensure compatibility between the first two

components; if this stage is appropriate the two systems merge into one, the surface one. A first conclusion here would be that the transmission of the surface system into the underground must be made while ensuring that the surface system extends vertically, so that virtually the entire mine has a single coordinate system.

2. The support system in underground mining can be the national one, namely the Cartographic stereographic projection system 1970, for Romania, or it may be a local system. As in our view MGIS will include not only mining information layers but also others, referring to mining environmental influences, such as a layer on pollution or environmental influences on mining such a layer with the mine power supply, using a local coordinate system is out of the question. A second conclusion, because it works with information from various industries, related to the direct mining activity, MGIS will operate only in the national coordinate system of the state in which it is applied, coded STEREO70 (below) for our country.

3. Elevation - in the underground, as in surface topography, a single elevation system is used, namely the national Black Sea in 1975 system. A third conclusion is that all rates from the surface and from the underground shall relate to that system mentioned, coded as RMN - BSM, (Reper Marea Neagr� - Black Sea Mark 1975).

4. Mining topography operates with two basic layers: the one for the underground called the general layer of the mine and the surface one called for the situation layer (surveying)[6]. The first one, encoded TOP012 and drawn to scale 1:1000, and the second encoded TOP010, will represent the origin information floors (layers) for MGIS. The fourth conclusion is that, unlike the traditional GIS, MGIS will operate with two "0" layers: Layer TOP010 for surface and Layer TOP012 for underground. Layer TOP010 will include all data with respect to the surface, both for mining and for those related to mining, from networks and infrastructure to vegetation or urbanism. Layer TOP012 will include underground information, from geological data to mining horizons.

5. Underground mining involves understanding the deposit, the railways and all the mining works in detail, both horizontally, operation solved by reference to the TOP012 Layer and vertically under TOP014 codes, Longitudinal sections and cross sections through the deposit at intervals not exceeding 100 m, scale 1:500 - 1:5,000; TOP018, Bank layer for every layer, stock, vein, scale 1:500 or 1:200; TOP020, Longitudinal profile of the main transport routes, horizontal and inclined, supplemented periodically according to the needs of operation; TOP021, Longitudinal profile of the wells, complete with cross sections, indicating the installations in the well and in the ramp for each

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horizon, updated after periodic checks provided in this Regulation; TOP022, Layer and sections of water hollows, pump installations, underground chambers, underground deposits; some of the documents displayed and viewed vertically. Although they are all perpendicular to the TOP012 layer, and although they are vertical layers, they are found in different orientations, or in the case of the CF longitudinal profile, which is a vertical surface, although transverse profiles are vertical layers. Optionally, one may choose two vertical surfaces intersecting at a point considered the center of the mine, with the orientation of the longitudinal layer either zero either a direction considered important for that mine, which we encode VER001, for the longitudinal layer and VER002 for the vertical layer, the other vertical layers receiving codes following their appearance in the MGIS design: Layer VER003, etc.; curved vertical surfaces can be coded by Layer SUP001, etc. and those inclined by Layer ICL001, etc. It is possible to give up the vertical calibration, each layer above being analyzed in relation to the information in the horizontal layers with which they are connected. Fifth conclusion: at the two basic information layers – origin – Layer TOP01 and Layer TOP002, two more, possible, Layer VER001 and Layer VER002 are added, thus defining a system with four layers, two horizontal and two vertical in MGIS; depending on the needs of each mine one can add an infinite number of other layers or reference surfaces (fig. 1). This will be the real challenge in MGIS design for each case.

Fig. 1. Layer structure in MGIS

6. Structuring of information entered in MGIS) will be done, as a first classification [5]: A. Thematic structuring of information:

1. Direct information 1.1. Structure and content of topo-geodetic information; 1.2. Structure and content of geological, hydro-geological information; 1.3. Information on the current context of exploitation.

2. Information on the influence of mining on the environment

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2.1. Structure and content of environmental monitoring information; 2.2. Management of mining waste; 2.3. Management of surface emergencies; 2.4. Stability studies of the land above and surrounding the mining area, subsidence

analysis, earthquake history, landslides; 2.5. Register of the rehabilitation and consolidation works of the areas above and

surrounding the mining area, embankments ; 2.6. Situation of groundwater, discharge of groundwater in surface waters.

3. Information about external influences on the mining industry 3.1. Situation of general, mining, agriculture, forestry cadastre, land rentals and

concessions; 3.2. Weather conditions 3.3. Hydrological conditions, waters, underground water; 3.4. Construction and infrastructure works, channels of communication, works of art; 3.5. Utilities, water, gas, electric and telephone networks; 3.6. Urban area, PUG, PUZ, PUD, RGU, RLU.

Information which is usually included in topographical designs enters this classification. A

second classification of information is: B. Structuring the information in the report according to the area they come from:

1. Information regarding surface data; 2. Information regarding underground data.

The third classification concerns the free movement of information. C. Structuring information in relation to their free movement:

1. Public information; 2. Restricted information; 3. Secret state information, classified information.

A fourth classification is made according to the nature of information and how they are

integrated (or not) in the system. D. Structuring information in relation to access and destination into MGIS:

1. Non-graphical analyzed information which will not enter into MGIS (neither in MBD GIS);

2. Graphical analyzed information which will not enter into MGIS (neither MBD GIS); 3. Non-graphical analyzed information which will not enter into MGIS, but will enter the

MBD GIS databases; 4. Non-graphical analyzed information which will enter into MGIS, as attribute data; 5. Graphical analyzed information which will enter into MGIS, forming layers.

The fifth and sixth classification refers to the degree of information updating in MGIS. E. Structuring information in relation to the novelty it represents:

1. Updated information; 2. Outdated information, which requires refreshing to date; 3. Historical information that is stored in this form.

F. Structuring information in relation to how and when it is updated: 1. Information with continuous synchronous non-stop update; 2. Hourly updated information;

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3. Daily updated information; 4. Weekly updated information; 5. Monthly updated information; 6. Annually updated information; 7. Multi-annually updated information ; 8. Variably updated information; 9. Information that does not require updating.

7. The 1:1000 scale will be only representation scale for the entire system. There may be

documents surveyed, such as the base map at a scale of 1:25000 or smaller scale maps; they are externalized to MGIS, have an advisory nature, and can be analyzed in parts which are surfaces from the perimeter or proximity of the mine.

8. “The Geographic Information System is one of the computer based technologies with the fastest growth. These systems are used in various fields related to the resources field such as resource management, environmental monitoring, utilities, financial planning, transport and market research. The use of GIS has expanded in society over the past decade, faster than any other analytical information technology [1]. This trend was not directly reflected in the underground mining resources sector. Most information technology investigated by mining professionals focuses on the descriptive aspect of the data, although large amounts of mining data can be spatially referenced” [2]. Based on this quote and studying a vast bibliography we have identified the main problems of implementing GIS technologies in the mining industry (becoming MGIS):

1. MGIS can help solve a major current information issue in mining, although most data being used can have a spatial representation; focus is now on the descriptive aspect of the data.

2. MGIS, as an argument, having the opportunity to present data in the form of layers and maps, can provide better aid in decision making for the people responsible in mining, than tabular information or, as stated previously, descriptive information.

3. MGIS has had a very slow evolution in comparison to the explosive development of GIS, because software exists and is applied, basically computer packages for the mining industry, which have sufficient modelling resources for the demand of spatial information in this industry [7].

4. MGIS has had a more difficult entry into the mining industry also because of the CAD technology which was very cheap (some of it), so it expanded, and then operators were familiar with these operating modes which could be interfaced to the aforementioned mining software relatively easily.

CONCLUSIONS

MGIS is currently underused! Following research we have found that, in the field approached here, the situation worldwide was and remains the following:

� Unreasonable, GIS has a hard time getting into the mining industry; � The concept of Mining GIS was defined several years ago, there are several applications,

very different, and it was concluded that a concept was not defined, only the idea that MGIS label mining can be applied to GIS applications;

� There are positive signs: some producers of mining software have gone to configuring GIS interfaces, general software manufacturers have created interfaces with mining software and two or three producers of GIS software have created interfaces with software from different categories.

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REFERENCES 1. BURROUGH, P. A. and MCDONNELL, R. (1998), Principles of Geographical Information Systems.

New York: Oxford University Press. Buttenfield, B. P. (1993): Representing data quality. Cartographica, 30(2/3), pp. 1-7.

2. CARTER, W. (2006), Application of Geographical Information System in Underground Coal Mine to assist Operational Management, A dissertation submitted by Mr Andrew William Carter, In fulfilment of the requirements of Courses ENG4111 and 4112 Research Project University of Southern Queensland, Faculty of Engineering and Surveying.

3. R�DULESCU M.V.G., RADULESCU A.T.G., RADULESCU G.M.T. (2009), GIS in mining & exploration, as a tool to based mining revenue management system, THE NATIONAL TECHNICAL-SCIENTIFIC CONFERENCE „Modern technologies for the 3RD Millenium” – ORADEA, 2009, Analele Universit�ii din Oradea, fascicula Construcii �i Instalaii hidroedilitare, ISSN 1454-4067, Vol.XII, Cod CNCSIS 877.

4. R�DULESCU M.V.G., RADULESCU A.T.G., B�DESCU G. (2010), GIS Applications in the Field of Maramures Mining, 2010 ESRI Survey&Engineering GIS Summit, 10-12 IULIE, 2010, San Diego, California, SUA, http://.esri.com/library/userconf/proc10/vc/papers/pap_1880.pdf-WIE.

5. RADULESCU, M.V.G. (2012), Contributions to the realization of a concept on creating mining data bank, PHD Thesys, University of Petrosani, Faculty of Mines, Scientific Coordinator, Prof. univ. dr. ing. Nicolae DIMA.

6. R�DULESCU M.V.G., RADULESCU A.T.G. (2010), GIS Applications in the field of Subterranean Mining Exploitations 3. The Case Study of the Maramures Mining Industry, The Fifth Session of the International Conference Geotunis 2010, The use of GIS and remote sensing for sustainable development, Tunisia from 29 November to 03 December 2010.

7. R�DULESCU C.,R�DULESCU M.V.G. (2011), Approaches of the management informational systems regarding the implementation of the Geographic Information Systems (GIS) in the mining basins of Romania International Multidisciplinary Scientific GeoConference & expo SGEM, the 12th international geoconference SGEM 2011, 17 - 23 june, 2011, paper 92, Section 7. "Geodesy and Mine Surveying" http://www.sgem.org/sche_pub_schedule.php, scientific data based indexing ISI Web of Science, Web of Knowledge, CrossRef and Scopus.

8. *** MIS-MINERAL INFORMATION SYSTEM (2011), Indian Bureau of Mines, http://ibm.nic.in/ reportch7.13.pdf, accesed at 12.02.2012.

9. *** GeoGRAFX GDMS(2010), ttp://www.geografxworld.com/index.php?option=com_content&view= article&id=22&Itemid=43, accesed at 14.02.2012.

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MGIS (MINING GEOGRAPHICAL INFORMATION SYSTEM) A NEW CONCEPT FOR THE INFORMATIZATION MANAGEMENT ON MINING

COMPANIES. INTRODUCING MINING DATA IN GIS AND ITS TRANSFORMATION IN MGIS

R�DULESCU Virgil Mihai G. M., R�DULESCU Corina, Technical University of Cluj Napoca, e-mail: [email protected]

A B S T R A C T Geo-coding and geo-referencing all information entered in the system is one of the most important aspects to successfully creating MGIS. The creation of MGIS should be regarded as starting from a pre-configured GIS for the location of the mine. The existence of an Urban GIS, Environment GIS, Geology GIS would create an ideal support on which to build the MGIS. However in the earlier phasing, it can be seen that that it is preferable to build an Urban GIS, then to implement MGIS. We think the first benefits of MGIS is the possibility of viewing virtually any information in a graphically digital manner, and then the possibility of simulation and study of the alleged situation’s effects. Another huge advantage is the actuality of data, if the system is working properly, because information can be analyzed after a few moments from running it. But, by currently using MGIS, we do not have to reach such events as the system provides sufficient data to prevent them.

Keywords: MGIS (Mining Geographical Information System ), mining database combined with a GIS platform


INTRODUCTION

The degree of computerization of mining companies is extremely varied from mine to mine, but the before mentioned transnational nature of the extractive industry paradoxically makes some mines in Africa, Senegal, Ghana, South Africa, Botswana or Latin America, Chile, Peru to have a higher rate of computerization compared to mining organizations in countries with a centuries old mining activity like the ones in Europe.

The informatisation ways are different, with some companies opting for mining software, others developing GIS platforms [1]. Others, very rare, have combined the two methods and most have developed custom general software, tailored for the field. MATERIALS AND METHODS 1. Introducing mining data in GIS, it's convert in MGIS

As in any field of activity, in mining data has to be the most important resource, which must be very well managed to serve effectively in organizational management in making sound, timely and efficient decisions [3].

GIS has the ability to graphically display a lot of the information, grouped in issues, going around in mining, and MGIS will become a very useful tool in managing the industry’s activity [2].

Geo-coding and geo-referencing all information entered in the system is one of the most important aspects to successfully creating MGIS and then MBD GIS [4]. The 1:1000 scale will be only representation scale for the entire system. There may be documents surveyed, such as the base map at a scale of 1:25000 or smaller scale maps; they are externalized to MGIS, have an advisory nature, and can be analyzed in parts which are surfaces from the perimeter or proximity of the mine. The creation of MGIS should be regarded as starting from a pre-configured GIS for the location of the mine.

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Fig. 1. MGIS configuration stages

The existence of an Urban GIS, Environment GIS, Geology GIS (or a system such as

Geological Information Register, from the Republic of Moldova [7]) would create an ideal support on which to build the MGIS. However, in the earlier phasing, it can be seen that that it is preferable to build an Urban GIS, then to implement MGIS. The steps to achieving MGIS in its final operational version need to be seen like so [4]: 1. MGIS can be built, following the steps suggested in Figure 1:

a. The graphical topographical-mapping-cadastral data are introduced into the GIS, together with plans and digital maps, or digitized analytical ones, then urban data, starting from PUG, encoded through CAD (cadastral informations) and URB (urban informations) shall transform GIS in URBAN GIS; it is appropriate to use the origin layer as support, a larger topographic plan, at a scale of 1:1000, which also should include the situation plan of the mining enclosure, which becomes LTOP010 in MGIS. The work is obviously carried out in STEREO70, RMN75, thus ensuring compatibility of the system.

b. The reporting origin supports are reconfigured by adding origin layers LTOP012 horizontally and LVER001, LVER002 vertically. It is obvious that both the horizontal and vertical layers can be extended, in fact they are reference layers, not perimeters, and at the same time they are still when it comes to position, but not content. BASIC MGIS is obtained.

c. The graphic data most in use are introduced, generally in the analog version, mining surveying data, coded TOP, geological, coded GEO and the ones from the underground mine structure and the current exploitation activity, data encoded by MIN. We obtain a new version of MGIS, which we called EXPLOITATION MGIS.

d. Introduction of economic-financial supply and sales data and human resources data, information coded FIN, and MRU, can be done only as attribute data, databases, and then they enter the training Chapter MBD GIS; or, the data can be introduced by geo-encoding directly in a

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graphical format, in the STEREO70 system. At least initially, a mixed solution might be used. In any format, a new GIS will be reached, called COMPANY MGIS, as it contains all the data that the mining, company and agency management needs to efficiently lead and manage the organization’s activity.

e. Considering that one of the biggest problems a mining company faces (see the case of Ro�ia Montan�) is the impact of activity on the environment, the introduction in the system, probably from the beginning, divided thematically not chronologically, is essential for proper functioning, and even for the survival of the company. The data in this category, even in digital format, are most easily obtained, since the authorities and civil society ensure that the data, encoded in MED system, are complete and current. Data on climate conditions, meteorological, hydrological, coded CLI, are equally important and equally easy to obtain from the authorities.

A final remark: the offer of GIS configuration and administration software is relatively

small, but each potential MGIS user will choose that software that best suits the actual conditions in which it must operate, the amount of data, the needs to interface with other software, etc.

2. Defining the advantages of MGIS, conclusions on the context

We believe that from the foregoing analysis and presentations results much of the benefits of MGIS and of the management process that may rely on it as a leading provider of information, compared with traditional management methods based largely on the analysis of written reports [6].

We think the biggest advantage offered is the possibility of viewing virtually any information in a graphically digital manner, and then the possibility of simulation and study of the alleged situation’s effects. Another huge advantage is the actuality of data, if the system is working properly, because information can be analyzed after a few moments from running it. In case of underground accidents, knowledge of the effects, if the MGIS system is doubled by a sensory system monitoring the processes and activities underground, can be made spontaneously, and the system will provide the most current and complete information on the organization and coordination of the emergency situation. But, by currently using MGIS, we do not have to reach such events as the system provides sufficient data to prevent them [5]. The biggest gain will be for the daily activity of all those using the information in the current activity, from sector heads to the unit manager, not to mention the managers of upper structures who will be able to oversee the entire activity of all subordinate institutions from their offices.

Another big advantage is the possibility of having synchronous data for important information regarding: state of underground railway, air, gases, dust, water infiltration, situation of perforations and directed explosions, operation of machinery, mining machine, and many other events, obviously depending on the equipment and specific of each mine, but also in relation to the possibilities and willingness to invest in monitoring the underground mining environment. The global crisis did not affect the overall software market and much less the GIS expansion, but in mining, if mining software and the ones with mining applications expand, GIS remains, inexplicably, an untapped opportunity [4]. There are positive signs: some producers of mining software have gone to configuring GIS interfaces, general software manufacturers have created interfaces with mining software and two or three producers of GIS software have created interfaces with software from different categories, all isolated actions, not a trend or a recipe for success. Large multinational mining corporations have operations worldwide, so the implementation of GIS or of efficient software in the field, cited in the paper, has an explanation that is not necessarily related to a local initiative, and here I mean the countries of Latin America or India, Indonesia or even some African countries. The companies being transnational, the system is implicitly applied to

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mines belonging to them, so we find GIS applications in mines in Botswana, Sierra Leone, Indonesia, Malaysia. We believe that at present four countries can be considered promoters (mining companies in these countries) of computerization in mining using (also) GIS technologies: Australia, Canada, China and India. The first two countries were champions in promoting GIS worldwide, since the launch of this important management tool. Understandably, I was refused details on the practical aspects of GIS implementation in business activities in the above category. Public information is that GIS applies to Mine X, but the rest is information that companies consider classified. In environmental monitoring, GIS has become the main operator, mining involving strong environmental influences whose effects must be known;

CONCLUSIONS The application of Geographical Information Systems (GIS) in mining still remains low, and although the mining industry’s share in the global economy is significant, reaching up to 70% in some countries, the sale of GIS licenses is less than 1% of all licenses sold by the largest manufacturers, ESRI and Intergraph. Then, at the suggestion of the scientific leader we went to a more extensive analysis of the process of computerization in mining, further observing that the computerization of mining, especially in large companies, is well established, but through specialized mining software with punctual applications. The concept of Mining GIS was defined several years ago, there are several applications, very different, and it was concluded that a concept was not defined, only the idea that MGIS label mining can be applied to GIS applications. REFERENCES 1. CARTER, W. (2006), Application of Geographical Information System inUnderground Coal Mine to

assist Operational Management, A dissertation submitted by Mr Andrew William Carter, In fulfilment of the requirements of Courses ENG4111 and 4112 Research Project University of Southern Queensland, Faculty of Engineering and Surveying.

2. R�DULESCU M.V.G., R�DULESCU A.T.G.,R�DULESCU G.M.T. (2009), GIS in mining & exploration, as a tool to based mining revenue management system, THE NATIONAL TECHNICAL-SCIENTIFIC CONFERENCE „Modern technologies for the 3RD Millenium” – ORADEA, 2009, Analele Universit�ii din Oradea, fascicula Construcii �i Instalaii hidroedilitare, ISSN 1454-4067, Vol.XII, Cod CNCSIS 877.

3. R�DULESCU M.V.G., RADULESCU A.T.G., B�DESCU G. (2010), GIS Applications in the Field of Maramures Mining, 2010 ESRI Survey&Engineering GIS Summit, 10-12 IULIE, 2010, San Diego, California, SUA, http://.esri.com/library/userconf/proc10/vc/papers/pap_1880.pdf-WIE.

4. R�DULESCU, M.V.G. (2012), Contributions to the realization of a concept on creating mining data bank, PHD Thesys, University of Petrosani, Faculty of Mines, Scientific Coordinator, Prof. univ. dr. ing. Nicolae DIMA.

5. R�DULESCU M.V.G., R�DULESCU A.T.G. (2010), GIS Applications in the field of Subterranean Mining Exploitations 3. The Case Study of the Maramures Mining Industry, The Fifth Session of the International Conference Geotunis 2010, The use of GIS and remote sensing for sustainable development, Tunisia from 29 November to 03 December 2010.

6. R�DULESCU C., R�DULESCU M.V.G. (2011), Approaches of the management informational systems regarding the implementation of the Geographic Information Systems (GIS) in the mining basins of Romania, International Multidisciplinary Scientific GeoConference & expo SGEM, the 12th international geoconference SGEM 2011, 17 - 23 june, 2011, paper 92, Section 7. „Geodesy and Mine Surveying” http://www.sgem.org/sche_pub_schedule.php, scientific data based indexing ISI Web of Science, Web of Knowledge, CrossRef and Scopus.

7. *** http://lex.justice.md/index.php?action=view&view=doc&lang=1&id=336563, accesed at May 2, 2011.

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CONSIDERATIONS ON THE CALCULUS OF CARBON FIBRE REINFORCED POLYMER (CFRP) STRENGTHENED BEAMS

TRIFA Florin Sabin,

University of Oradea, Faculty of Constructions and Architecture, e-mail: [email protected]

A B S T R A C T This work presents a method for complete calculation of the reinforced concrete beams with T-shaped cross-section strengthened with carbon fibre reinforced polymer (CFRP) strips. Applying the classical reinforced concrete beam theory which assumes that plane sections remain plane, that is, strain compatibility is assumed, the proposed calculation method enables to predict the failure mode in flexure of the strengthened element, the failure bending moment and also the load/deflection history until the failure occurs. The method described below may be a useful tool for checking both the ultimate load and the ductility of the strengthened beams.

Keywords: reinforcement, CFRP, polymer, deflection, strengthening


INTRODUCTION If the strength structures are subjected to strong earthquake motions, their main elements are

forced to undertake a number of incursions into the inelastic range of their behavior. In such circumstances, beyond the bearing capacity and the stiffness under the design loads, also comes the demand to ensure of sufficient capacity of inelastic deformation (ductility) needed for the absorbtion of the energy induced by the earthquakes without the risk of reaching the failure state.

In the case of reinforced concrete structures whith members strengthened with carbon fibre reinforced polymer strips, beyond the restoration of the bearing capacity, also the check to ensure a proper inelastic deformation capacity is needed, i.e. the checking of the following requirements:

• to ensure the members of the structure a ductile failure mechanism in bending, i.e. the crushing of the concrete in the compressed area of the cross-section to occur only after large enough strains in the tensiled bars of the reinforcements and in the CFRP strips develop;

• to avoid untimely failure of brittle nature of the elements due to shear force action. The calculation method specified in [1] and [6], although, overrates the safety in terms of the

elastic bearing capacity, by under-rating the value of the resisting bending moment and also, underrates the shear force associated to this for the following reasons:

• ignores the strengthen effect of the reinforcement steel after yielding; • the contribution of the reinforcement bars from the web of the beams cross-section and of

those from the compressed area of the cross-section to the resisting bending moment is neglected. A more accurate calculation method of the bearing flexural capacity of the T-shaped cross-

sections of the strengthened beams with CFRP strips in which the approach to calculate the resisting bending moment of the cross-section removes the above mentioned shortcomings, is presented below.

Also, a method for calculation of the beam displacements, at all loading stages, until the failure occurs, taking into account the action of the bending moment distributed along the beam as shown in Figure 1, is proposed.

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Fig. 1. Beams strengthened with CFRP strips. Moments and curvatures variation

MATERIALS AND METHODS 1. Calculation of the strenthened Beam 1.1. Basic Assumptions

The proposed calculation method is based on the following assumptions: • The working out of the range of the tensiled concrete after cracking; • The linear distribution of the specific longitudinal strains (ε) on the height (h) of the beam

cross-section, at all loading stages, up to failure (Bernoulli's hypothesis), as shown in Figure 2. As shown in Figure 2, at each loading stage, characterized by different values of the specific

strain of concrete in the most compressed fibre (εb), the specific strain of the row “ï” of reinforcement bars, at distance (hoi) from this fibre, is:

x

xhoibai

−= εε (2.1)

Fig. 2. Strain �εεεε distribution on the height of beam cross-section

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Given that the use of CFRP strips for the beam strengthening is made at a certain level of the beam loading due to the service loads, the specific strain of the CFRP strips in the tensiled area of the section is:

0εεε −−=x

xhbf , (2.2)

where 0ε is the strain from the tensiled fibre of the cross-section produced by the moment M0 due to service loads (see Figure 2).

If the section failure occurs through concrete crushing, in (2.1) εb = εbu, is taken, where buε is the ultimate specific strain of the concrete in compression (�bu = 2.5 to 4 ‰, depending on concrete quality).

• The use of the constitutive curve for the concrete in compression (σb-εb) with the parabolic shape, given in [1], taking into account the damage of the concrete compressive strength until the failure occurs (εb is reached). As shown in Figure 3, the compressive stress in concrete is given by:

��

��

�−=

00

2εε

εεσ bb

cb R (2.3)

where: cR is the compressive strength of concrete.

Fig. 3. Constitutive curve in Compression Fig. 4. Constitutive curve in Tensile of CFRP stips of the concrete

For the ultimate value of the concrete compressive strain (εb = εbu = 3,5‰) 5909,1== uαα and cb R65,0=σ are obtained, while the maximum value of ( )cbb R=σσ is reached at

2,20 == εε b ‰, corresponding to 10 == αα , where:

0ε

εα b= (�o=2,2‰) (2.4)

Given the notation (2.4), the concrete compressive characteristic equation becomes: ( )αασ −⋅= 2cb R (2.5)

The use for the constitutive curve (�a – �a) of the reinforcement steel in tension a bilinear diagram (Figure 5), to consider reinforcements strengthening (in tension or in compression) after

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the yield point (�ac), namely: �a>�ac, for any value of the specific strain of longitudinal reinforcement (in tension or in compression) �ai in the range [�ac, �al].

Fig. 5. Constitutive curve for steel reinforcement in tension/compression

acε is the strain corresponding to the yielding of the reinforcing steel and 1aε is the strain

corresponding to the strength of the steel at the breaking value ( )ara σσ = . From the constitutive curve of the reinforcing steel it follows:

- identical behavior in tension and compression of the reinforcements - depending on the specific strain value reached in the reinforcing steel in a certain loading

stage, the stress σa in a reinforcement bar is:

aaa E εσ = , for aca εε ±≤± (2.6)

or

( )aca1

εεεεσσ

σσ −−−

+=aca

acaraca , for aca εε ±≥± (2.7)

1.2. Calculation of the Bending Moment

As can be seen in Figure 6, to achieve the section equilibrium is necessary to satisfy the equations:

0=−−+ �� f

iai

iaib TTCC (2.8)

)()()()()( 00 GfGii

aiiGi

aidrpGbdrGbG hhThhThhCxhhCxhCM −+−+−+′′−−′′−′−′= �� (2.9)

where the following notations are used:

MG = the bending moment at the centroid of the cross-section; Cb bb CC ′′−′= = the compressive resultant force in the concrete from the compressed area of

the cross-section; bC ′ = the compressive resultant force of the compressed area with width (bp) and height (x);

bC ′′ = the compressive resultant force of the compressed area with width (bp-b) and height (x-hp);

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Cai = the compressive force in reinforcement row „i”, located at distance ( oih ) measured from the most compressed fibre of the cross-section;

Tai = the tensile force in reinforcement row „i”, located at distance ( oih ), measure from the most compressed fibre of the cross-section;

Tf = the tensile force in CFRP strip; h = the height of the cross-section; hoi = the distance from the reinforcement row „i”, to the most compressed fibre; x = the height of the compressed area; x = the distance from compressed fibre to neutral axis ( )xx ≤ ;

stx = the distance from compressive resultant force in the concrete to neutral axis of the cross-section;

drx = the distance from compressive resultant force in the concrete to the most compressed fibre of the section.

Fig. 6. The internal forces in the cross-section of the strengthened beam a) Calculation of Compression Resultant Forces in Concrete ( bC ′ , bC ′′ )

From Figure 6 results:

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xx

b

b =εε

(2.10)

yielding:

bb xx εε = (2.11)

the strain biε at the link between the web and the flange is:

bp

bi x

hxεε ⋅

−= (2.11’)

Substituting (2.11) in equation (2.5), stress in the compressed concrete bσ given by:

22

00

2 ��

��

��

��

�−=

xx

Rxx

R bc

bcb ε

εεεσ (2.12)

where, with the notation 0ε

εα b=′ ,

it follows:

( )2

22 ��

��

�′−′=

xx

Rxx

R ccb αασ (2.12’)

Compression resultant forces in the concrete are calculated by integrating the compressive

stress on the surface of the compressed area of the cross-section, as follows: - for the area with width (bp):

( )�=′ x

0 xdxbC bpb σ (2.13)

where (bp) is the width of the flange section.

Substituting (2.12) in (2.13) and performing calculations it comes:

cpcpb xRbxRbC 3

1 ωαα ′=��

��

� ′−′=′ (2.14)

where:

��

��

� ′−′=′

31

ααω (2.15)

For the failure stage through the crushing of the concrete of the compressed area of the cross-section, when 5,3== bub εε ‰, 74725,0=′=′ uωω is obtained. - for the area of width (bp –b):

( )� −=′′ ph- x

0 )( xdxbbC bpb σ (2.16)

In the same way as in the resultant force bC ′ case, the force bC ′′ is: cpb RxbC ⋅′′⋅′′⋅′′=′′ ω (2.17)

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where : )3

1(ααω ′′

−′′=′′ ;0ε

εα bi=′′ ; bbb pp −=′′ ; phxx −=′′

The Centroid positions of compressive resultant forces in the concrete to the neutral axis are calculated with formulae:

( )( ) xdxb

xdxbxx

bp

bp

st

x

0

x

0

σ

σ

�

�=′ (2.18)

�

�−

−

⋅⋅−

⋅⋅−⋅=′′

p

p

hx

bp

hx

bp

stxdxbb

xdxbbxx

0

0

)()(

)()(

σ

σ

( ) xxx stst γαα ′=

′−′−=′

3438

(2.19)

( ) xx stst γαα ′′=

′′−′′−=′′

3438

where: 0ε

εα b=′ �i0ε

εα bi=′′

Taking into account (2.19), drx ′ and drx ′′ are calculated as follows:

( ) xxx drdr γαα ′=

′−′−=′

344

(2.20)

( ) xxx drdr γαα ′′=

′′−′′−=′′

344

For failure stage through the concrete crushing ( 5,3== bub εε ‰), the values are:

57258,0, =′ ustγ and 42742,0, =′ udrγ . b) Calculation of the Compressive Resultant Forces in the Reinforcement Bars ( aC )

In a row "i”, of compressed bars locate at the distance (h0i) from the most compressed fibre, the compressive force is:

aiaiai AC σ= (2.21)

where: aiσ is calculated with one of (2.6) or (2.7) formulae, as acai εε < or acai εε > .

The compressive resultant force in the compressed reinforcement bars is:

ai

k

iai

k

iaia ACC σ��

==

==11

(2.22)

where: k is the number of rows of compressed bars. c) Calculation of the Tensiled Resultant Force in the Reinforcement Bars ( )aT

The tensile strength of the reinforcing steel in the row "i" of tensiled bars, located at the

�

��

120

distance ( ih0 ) measured from the most compressed fibre is:

aiaiai AT σ= (2.23)

where: aiσ has the formulae (2.6) or (2.7), as acai εε < or acai εε > .

The tensile resultant force of the tensiled reinforcement bars is:

ai

j

i

j

iaiaia ATT σ� �

= =

⋅==1 1

(2.24)

where: (j) is the number of tensiled bar rows. d) Calculation of the Tensiled Force in CFRP Strip ( Tf )

Tensile strength of carbon fibre strip is: fff AT σ⋅= (2.25)

Substituting the expressions obtained in a), b), c) and d) in equation (2.9), it follows:

( )( ) 011

=⋅−⋅−⋅+−−′′−′′ ��==

ff

j

iaiai

k

iaiaicppcp AAARhxbbxRb σσσωω (2.26)

After making substitutions in equation (2.26) above, two unknowns are involved: x and bε . e) Practical Calculus of the Bearing Capacity

Giving successively increasing values for the strain ( bε ) of the concrete from the most compressed fibre to the value 5,3=buε ‰, corresponding to failure in compression of the concrete an unknown ( )bε is removed and thus equation (2.8), as given by (2.26) becomes a grade 2 equation with unknown x, which gives the position of the neutral axis at the loading level defined by the value given for bε .

With the value of (x) thus determined, then are calculated the values of aiε with equation (2.1), fε with (2.2) and the corresponding values for aiσ , aiC , aiT , fT , bC ′ and bC ′′ . Introducing them into (2.9), the bending moment M, acting on centroid on the cross-section, coresponding to the introduced value of the strain in the most compressed concrete fibre is calculated.

At each loading stage ( bub εε ≤ ) the strength conditions for stresses in the reinforcement bars, the adhesive and composite strip are checked.

For 5,3== bub εε ‰ the ultimate values for x and M are obtained for the case of the cross-section failure through the compressed concrete.

Practically, the section calculation is carried out by trials, initially estimating the number of tensiled bars (j) and of the compressed ones (k) and also the bars that have exceeded or not the yield stress. After determining the value of x, aiε is calculated and the initial assumption is verified. In case of mismatch, the calculus is redone as many times as needed.

Based on x and M determined as described above, are further calculated for each loading stage (i.e. for each value given to parameter bε ) the corresponding values of bσ , aε , Φ and bbIE as follows:

• bσ to be calculated with (2.5) • aε to be calculated with (2.1) where ( oih ) corresponds to the row of the most tensiled

reinforcement bar

��

�� !� �!��

121

• Curvature � to be calculated [2] with:

Φxbε

= (2.27)

• Bending stiffness modulus of the reinforced concrete cross-section strengthened with CFRP strips subjected to bending ( bbIE ) is computed with:

Φ

= MIE bb (2.28)

1.3. Displacements Calculation A method of displacements (w) calculation of the beam from the Figure 1, strengthened with

CFRP strips is presented below. It is noted that in an actual cross-section of beam, the displacement, the bending moment and

the curvature of the deflected axis vary, i.e.: w = wx , M = Mx and Φ = xΦ , respectively. If the load exceeds a certain level, i.e. if the maximum value of the bending moment Mmax

exceeds a certain value, the Mx - xΦ curve becomes non-linear on the segments of beam having the same length "a" even if the values of M linearly vary.

The actual cross-section curvature formula, determined [2] for the bended bars with homogeneous cross-section, is:

( ) IE

M1

xbb

x−==Φx

x ρ (2.29)

where: ρ x is the radius of curvature of the deflected axis of the beam which also varies along the beam as the bending modulus of stiffness ( EbIb )x of the cross-section does.

The equation of the deflected axis in this case is:

( ) xxbb

xx

IEM

dxwd

Φ=−=2

2

(2.30)

Integrating twice the equation (2.30) to obtain the expression of the deflected axis of the beam produced by the action of the bending moment is:

�� ⋅Φ=x

x

x

x dxdxw00

(2.31)

Seen from Figure 1 that �Φ=⋅Φ

x

xx Adx0

, that is the area of the curvature chart having the

length x (to the cross-section where wx is calculated). It follows:

� ⋅= Φx

xx dxAw0

(2.32)

which is integrated by parts and is results:

�ΦΦ ⋅−⋅=

x

xx

xx dAxxAw00

(2.33)

Applying the theorem of the statical moment to calculate the integral above, the displacement of the deflected axis calculated in the actual cross-section of the beam, is:

( )cxxcxxxx xxAxAxAw −⋅=⋅−⋅= ΦΦΦ (2.34)

where: (x – xcx) is the distance from the centroid Cx of the area ΦxA to the actual cross-section

where the displacement is calculated. For x = l/2 we get the relation for calculating the maximum value of the displacement in the

cross-section at the middle of the beam span:

�

��

122

( )cxlAw −⋅==∆ Φ 2/max (2.35)

where: ΦA = area of the curvature chart from the half of the beam span;

xc = distance from the centroid C of the area ΦxA to the middle of the beam span.

Formula (2.35) set the practical rule to calculate the maximum value of the displacement in the middle span cross-section as the statical moment calculated in this section of the Φ chart area from the half of the beam span. For a row of “n” increasing values of the bending moments in the cross-section where maxMM n = , the moments ni MM ≤ ( )ni ,1= from the beam correspond to the cross-section having the position given by:

aMM

xn

ii ⋅= (2.36)

Since the curvature was calculated with (2.27) for each value iM , it follows that the curvature chart of the varation is known and therefore it can be performed the numerical calculus of the displacement ∆ with (2.35) formula.

CONCLUSIONS

The calculation method presented allows a complete calculus of a reinforced concrete beam strengthened with CFRP strips, i.e. the calculation of the bearing capacity and of the displacements, respectively. The described calculation is a verification calculus of the strengthened beam because it assumes as known the following data: concrete cross-section, the amount and the distribution of the reinforcing steel in the cross-section, the CFRP strip section and the mechanical characteristics of all these materials. Through applying it, is possible to identify the failure mode of the beam and to plot the force ( )P - displacement ( )∆ chart needed to analyze the beam ductility after strengthening.

REFERENCES 1. AGENT, R., POSTELNICU, T. (1982, 1983), Calculul structurilor cu diafragma din beton armat, vol.

I. elastic �i vol. II. Postelastic (Analysis of Structures with Reinforced Concrete Shear-walls), Editura Tehnic�, Bucure�ti.

2. BIA, C., ILLE, V., SOARE, M. V. (1983), Rezisten�a materialelor �i teoria elasticit��ii (Strength of Materials and Theory of Elasticity), Editura Didactic� �i Pedagogic�, Bucure�ti.

3. BRINZAN, I., TRIFA F. (1988), Cercet�ri �i experiment�ri privind zvelte�ea inimii diafragmelor din beton armat monolit cu leg�turi transversale pe contur. Referat cu concluzii. Completare prescrip�ii tehnice �i propuneri pentru proiectarea tip (Researches and Experimentations regarding the Slenderness of the Reinforced Concrete Shear-wall Webs with Boundary Transverse Connections. Report with conclusions. Supplement of Design Provisions and Proposals for the Design of the Type Buildings) – INCERC – LSC, Bucure�ti.

4. TRIFA, FL. S., PRADA, M. (2000), Metod� de calcul a capacit��ii portante a sec�iunilor diafragmelor din beton armat monolit (Calculation Method of the Bearing Capacity of the Reinforced Concrete Shear-walls cross-sections), Analele Universit�ii din Oradea, TOM III. 2000, Fascicula Construcii �i Instalaii Hidroedilitare.

5. *** STAS 10107/0-90, Calculul �i alc�tuirea elementelor structurale din beton, beton armat �i beton precomprimat (Calculus and Detailing of the Reinforced Concrete and Prestressed Concrete Structural Members).

6. STOIAN, V., NAGY-GYORGY, T., DAN D., GERGELY J., DAESCU, C. (2004), Materiale compozite pentru construc�ii (Composite Materials for Constructions), Editura Politehnica Timi�oara.

7. *** (2001), Technical report in the design and use of externally bonded FRP reinforcement for reinforced concrete structure, fib T.G.9.3.

��

�� !� �!��

123

THE INFLUENCE OF SHEAR ON THE INELASTIC DISPLACEMENT OF ECCENTRIC COMPRESSED REINFORCED CONCRETE MEMBERS

TRIFA Florin Sabin,

University of Oradea, Faculty of Constructions and Architecture, e-mail: [email protected]

A B S T R A C T This work presents a method for the calculation of the inelastic deflections of the reinforced concrete members subjected to the combined action of a monotonic increasing horizontal force and of a constant axial load. The method enables a separate evaluation of the deflection due the bending moment and that of the shear force, for each loading level induced by the horizontal force, until the collapse of the member is reached. Performed on the scheme of a vertical cantilever fixed at the bottom, the calculus is based on Bernoulli’s hypothesis, taking into account a linear strain distribution on the height of each cross-section. Two different approaches to predict shear deflections are proposed. The changing of the bending and shear stiffness along the height of the member is taken into account, also regarding the presence of the web vertical reinforcement and the hardening after yield point of the reinforcing steel. This method enables the plotting of the horizontal force - deflection curve of the member which displays the change of its stiffness at horizontal load until the collapse is reached and consists the initial point of the analysis of the inelastic behaviour of any structure.

Keywords: reinforcement, shear, inelastic, displacement, stiffness, compressed


INTRODUCTION To analyze the inelastic behavior of any resistant structure, as fundamental element for

determining its response to seismic action, it is of special importance the accurate assessment of the displacements produced by the horizontal forces, especially of those which exceed the limit of elastic behavior of the structure.

This is the case of the eccentric compressed members, such the columns or the structural walls are, where the values of the bending moments are important and also the shear force is significant that its influence on the stiffness at horizontal forces cannot be neglected anymore.

For example, as shown in [1], in the case of reinforced concrete coupled structural walls stressed in the elastic range, the displacement value due to shear force action in the walls can be up to 70-80% of the total horizontal displacement produced at the top of the coupled walls.

Therefore, taking into account in the inelastic range, even in an approximate manner, the shear effect on total displacement is compulsory. Hence, the total displacement will be: � = �M +

�T, where �M is the displacement due to the bending moment M and �T is the displacement produced by shear force T.

In this paper we present a method to calculate the inelastic displacements of a reinforced concrete member (column, structural wall) subjected to a combined action of a monotonically increasing horizontal force applied at the top and of a constant axial load, as shown in Figure 1.

�

��

124

Fig. 1. Eccentric compressed element. Moment and curvature variation in height

MATERIALS AND METHODS 8. Calculation of the Horizontal Displacements 1.1. Basic Assumptions

The performed on the static scheme of a vertical cantilever fixed at the base and loaded at the top with a constant axial force combined with a horizontal one, monotonically increasing up to the failure, the displacements calculation is based on the following assumptions:

• The working out of the tensiled concrete after cracking; • The neglect of second order geometrical effects; • The linear distribution of the specific longitudinal strains (ε) on the height (h) of the beam

cross-section, at all loading stages, up to failure (Bernoulli's hypothesis), as shown in Figure 2. • The use of a constitutive curve (σb-εb) for the compressed concrete with parabolic shape [1]

to take into account the degradation of the compressive strength of the concrete before failure (Figure 3) as well.

• The use in the angle shearing-deformation calculus of the secant modulus of elasticity of the concrete in compression obtained from the characteristic curve from Figure 3, as follows:

bbb /��E =′ ; • The use for the contraction coefficient of the concrete � (Poisson ratio) a variation law as

plotted in Figure 4;

Fig. 2. Strain � variation and internal forces in the eccentrically compressed cross-section

��

�� !� �!��

125

Fig. 3. Constitutive curve of the Fig. 4. Variation law of Poisson Ratio for compressed concrete the compressed concrete

• The use for the constitutive curve (�a – �a) of the reinforcement steel in tension a bilinear diagram (Figure 5), to consider reinforcement strengthening (in tension or in compression) after the yield point (�ac), namely: �a>�ac, for any value of the specific strain of longitudinal reinforcement (in tension or in compression) �ai in the range [�ac, �al].

• The use in displacements calculus of the secant modulus of elasticity of steel for stresses that exceed the yield point: aaa /��E =′ (for aca �� > ).

Fig. 5. Constitutive curve for steel reinforcement in tension/compression

1.2. Calculation of the Bending Moment Displacements (�M)

Calculation of compressed eccentric member displacements shown in Figure 1 is treated in detail in [8]. For a value of strain (εb) in the most compressed concrete fibre of the cross-section, the internal forces shown in Figure 2 must satisfy the equilibrium equations [7]:

TCCN

iai

iaib �� −+= (2.1)

( ) ( ) hhThhCxhCxhCMi

Goiaii

oiGaidrGbdrGb �� −+−+��

��

� ″−′′−�

�

��

� ′−′= (2.2)

where:

ai

ai CC =� represents the resultant of the forces produced in the compressed reinforcement

bars

�

��

126

� =i

aai TT represents the resultant of the forces produced in the tensiled reinforcement bars

bC ′′−′= bb CC is the resultant of the stresses from the compressed concrete area of the cross-section

bC′ = resultant of the stresses in the compressed concrete area having the width (bp) and the height (x)

bC ′′ = resultant of the stresses in the compressed concrete area having the width ( )bbp − and

the height ( )ph-x Making the substitutions in (2.1), it turns into an quadratic equation with the unknown (x), of

which the position of the neutral axis of the cross-section for the loading stage defined by the considered value of (�b) is determined. The value of (x), thus determined, �ai, Cai, Tai and Cb can be further calculated with formulae from [7] which, substituted in equation (2.2) from above, give the value of the bending moment (Mmax) at the base cross-section of the member. For the same loading stage, the curvature (�) and the bending stiffness modulus (EbIb) of the base cross-section are calculated with formulae ([3], [7]):

M

IE ; x� max

bbb

Φ==Φ (2.3)

With axes chosen as in Figure 1, if Mmax was determined into the base cross-section in the

considered loading stage, then the bending moment can be calculated in any other cross-section on the height of the member (ly = y), as follows:

y/H M/Hl M M maxymaxy ⋅=⋅= (2.4)

Given the variation of the bending moment in the height of the member, it results that the

bending stiffness of the cross-section and also the curvature vary from one section to another, i.e.: EbIb = (EbIb)y and � = �y. Also, given the linear variation of M along the y axis, it can be plotted the curvature � values depending on the bending moment considered on this axis, thus getting the (M – �) curve as shown in Figure 1.

The determination of the horizontal displacement (u) of the member due to the bending moment (M) in the considered loading stage is based on the curvature formula for the cross-section of the homogeneous bars [2]:

( ) IE

M

�

1

ybb

y

yy −==Φ (2.5)

where: �y is the radius of curvature of the member’s axis and (EbIb)y is the bending stiffness modulus of the cross-section, both being variables on the member’s height. The quadratic differential equation of the deflected axis is:

( ) IE

M

dy

ud

bb

y2y

2

yy

Φ=−= (2.6)

where: uy is the horizontal displacement of the actual cross-section. Integrating twice the equation (2.6) it follows:

��

�� !� �!��

127

( ) Φ⋅=⋅Φ⋅=��

��

�⋅Φ= �� ycy

0y

00

y

0y Ay-ydydydydyu

yyy

y (2.7)

where: Φ

yA = the area bounded by the portion of the function �y graph from the above the actual cross-section,

ycy = the distance from the centroid Cy of area ΦyA to Ox axis,

( ) Φ⋅ ycy Ay-y = the statical moment of the area ΦyA calculated to the actual cross-section.

Horizontal displacement of the cross-section of the top of the member produced by the bending moment M is further obtained customizing the formula (2.7):

( ) Ay-Hdydyu C

H

0

H

0ymaxM,M

Φ⋅=⋅Φ⋅==∆ � � (2.8)

where: yc = the coordinate of the centroid of the curvature � chart from the entire height of the

member, ΦA = � chart area from the height of the member in the considered loading stage (�b<�bu).

Thus results the maximum horizontal displacement due to the bending moment, calculated at

the top of the member for a given loading stage of its base cross-section (characterized by the value of the �b strain at the most compressed concrete fibre of the cross-section), is calculated as the statical moment of the chart of the curvatures (�1, �2,…. �i,…. �n) from the height H of the member to an axis parallel to Ox taken at the base cross-section.

1.2. Calculation of the Displacements due to the Shear Force (�T)

Given the linear variation of M on the height of the eccentrically compressed member (column, shear-wall, etc.), the shear force is constant on H and has the value:

HM

--ST n== (2.9)

at each considered loading stage (�b) (i.e. for any value given to �b at the base of the cross-section).

The average angle of shearing deformation on the length dy of the compressed member, located in the actual cross-section having the abscissa y, calculated with the equation for the beams made of homogeneous materials [2], is:

dydu

� Tm = (2.10)

where: ( )yuu TT = is the displacement due to the shear force Ty acting in actual cross-section.

Further on it will be customized m� from formula (2.10) in two different ways, initiating from the relations established in The Strength of Materials [2] for the beams made of homogeneous materials.

�

��

128

1.3.1. Customized relationship for calculating the average angle of shearing-deformation determined in The Strength of Materials

The relationship established in The Strength of Materials [2] for the average angle shearing deflection is:

AG

Tk� y

m ⋅⋅= (2.11)

where: Ty = the shear force in the cross-section with abscissa y, A = the homogeneous cross-section area, G = the modulus of elasticity in shear of the homogeneous material, k = the cross-section shape factor which takes into account the non-uniform shear stress �

distribution along the height of the cross-section. Given the particularities in structure and in loading of the eccentrically compressed member

with T-shaped cross-section, which is the object in this study, the formula (2.11) turns as follows: • The area of the gross cross-section Ab is replaced with the concrete ideal section area which

has the formula:

( ) ��==

⋅+−+=j

1iaiai

k

1iaiaibcbi AnA1nAA (2.12)

where:

Abc = the area of the compressed zone of the web = ,

b

aiai E

En

′′

= = the equivalence ratio between the steeel reinforcement bars of the row „i” and

the concrete, in which aiE′ and bE′ have been introduced at 1.1 from above, Aai = the area reinforcement bars of the row „i”, j = the number of the rows of tensiled reinforcement bars of the cross-section, k = the number of the rows of compressed reinforcement bars of the cross-section.

• it is considered ( ) ϕbb

b

E�12

EGG

′=

+′

== (2.13)

• Ty = T = constant along the height H of the member It follows:

bibbibm AE

Tk

AGT

k�⋅′

⋅⋅=⋅

⋅= ϕ (2.14)

where: Abi is given by (2.12). The horizontal displacement of the considered basic element, due to the shear force action in

the cross-section is [2]:

dyAE

Tkdy�du

bibmT ⋅

⋅′⋅⋅=⋅= ϕ (2.15)

��

�� !� �!��

129

But: - from formula (2.3) results: y

b�xΦ

=

- according to [6], k = 1 for T-shaped cross-sections Being given the (2.12) and making the substitutions in (2.15), by successive transformations

it is obtained:

( )

dy T� dy TAEAEE��b

� du y

j

1iaiai

k

1iaibaiyb

yT ⋅⋅=⋅⋅

��

��

�⋅′+′−′+⋅

⋅=

��==

ϕ (2.16)

where to the parameter yΦ , which varies along the height of the member, it can be assigned the significance of a corrected curvature. The horizontal displacement in the actual cross-section due to the shear force results by integrating (2.16):

�� ⋅⋅==y

0

y

y

0TyT, dy�Tduu (2.17)

The maximum horizontal displacement due to the shear force is obtained at the top of the

member, as follows:

� ⋅==H

0

ymaxT,T dy�Tu� (2.18)

Taking into account that H/MT max= where Mmax is the maximum value of the bending

moment from the height of the member, it can be also written:

� ⋅⋅==H

0

ymax

maxT,T dy�H

Mu� (2.19)

where: the integral has the geometric significance of the area ( )ΦA bounded by the function y� graph plotted on the height of the member. Practially, for the given value bi� of the strain in the

most compressed concrete fibre of the cross-section, the corresponding values for , ,M ii Φ �i iΦ are obtained. Giving bi� a set of "n" increasing values as many values for the bending moments Mi and

for iΦ corresponding to the cross-sections nii M/MHy ⋅= , are obtained. So, it can be plotted the

function yΦ chart from which then we determine ΦA from (2.18) and (2.19). 1.3.2. The use of the internal forces of the cross-section to determine the average angle

shearing-deformation From Hooke's law related to the shearing action for the bars made of homogeneous materials,

it results the average angle of shearing-deformation formula:

�

��

130

b

bm G

�� = (2.20)

Applying Juravski formula for homogeneous materials, we get [2] the expression of the shear stress b� :

b

yb Ib

ST�

⋅⋅

= (2.21)

where: b� has the value from the neutral axis of the cross-section if the statical moment is

calculated to the neutral axis (S = S0). Replacing (2.21) in (2.20) it results:

b

0

bm Ib

STG1

�⋅⋅

⋅= (2.22)

from which, through successive transformations, we get:

( )

( )b

0

bb

0

bb

0

bm Ib

STEIb

STE�12

IbST

�12E1

�⋅⋅

⋅′

=⋅⋅

⋅′+=

⋅⋅

⋅

+′= ϕ

(2.23)

We aproximate the lever arm (z) of the resultant force of the internal compressive forces to

the one of the tensile forces, as follows:

bacba CCM

CCCM

CM

z+

=′′−′+

=≅ (2.24)

From [2], for bars with homogeneous cross-section subjected to bending, the lever arm is

0

b

SI

z = . Given the (2.24) it comes the following approximate relationship:

,CC

MSI

ba0

b

+≅ from where: ( )

MI

CCS bba0 ⋅+≅ (2.25)

But: x�

Ex�

E�EIE

MIM b

bb

bbbbb

=′⋅=′⋅=′⋅⋅′

= (2.26)

It follows: ( )b

ba0 �

xCCS ⋅+≅ and replacing in (2.23) we get:

��

�� !� �!��

131

( ) ( )T

IE�bxCC

�IxCC

bT

EIS

bT

E bbb

ba

bb

ba

bb

0

bm ⋅

⋅′⋅⋅⋅+

⋅=⋅

⋅+⋅⋅

′≅⋅⋅

′= ϕϕϕγ (2.27)

As T = constant on H and the other parameters vary, the formula of the maximum horizontal

displacement produced at the top of the member is obtained, as follows:

( )�� ⋅′⋅⋅

⋅+⋅⋅=⋅==

H

0 bbb

baH

0mmaxT,T IEb

xCCTdyu�

σϕγ (2.28)

The integral equation (2.28) can be numerically solved by dividing the height of the member

into intervals having the length (Li – Li-1) on which the bending moment to be considered constant and equal to the maximum value Mi from the cross-section "i". CONCLUSIONS

Based on those shown in section 1 above, it can be underlined the following main findings: • it is possible to calculate the total horizontal displacement of a reinforced concrete member

eccentrically compressed, subjected to the combined action of a constant axial load with a horizontal force applied at its top, monotonically increasing up to the failure, with the separate assessment of the displacement components due to the bending moment and the shear action;

• in the calculation presented in this paper, there are considered accurately enough the mechanical properties of the compressed concrete and of the reinforcement steel, in tension and in compression, including those from the web of the T-shaped of the cross-section of the member;

• because the method allows the displacement calculation at each loading stage with the horizontal force, based on the results obtained for each stage, it can be plotted the force - top horizontal displacement curve, until the final stage (failure) occurs, that the push-over inelastic analysis of the structure to which the studied member belongs becomes possible.

REFERENCES 1. AGENT, R., POSTELNICU, T. (1982, 1983), Calculul structurilor cu diafragme din beton armat, vol.

I. elastic �i vol. II. postelastic (Analysis of Structures with Reinforced Concrete Shear-walls), Editura Tehnic�, Bucure�ti.

2. BIA, C., ILLE, V., SOARE, M. V. (1983), Rezisten�a materialelor �i teoria elasticit��ii, (Strength of Materials and Theory of Elasticity), Editura Didactic� �i Pedagogic�, Bucure�ti.

3. BRINZAN, I., TRIFA F. (1988), Cercet�ri �i experiment�ri privind zvelte�ea inimii diafragmelor din beton armat monolit cu leg�turi transversale pe contur. Referat cu concluzii. Completare prescrip�ii tehnice �i propuneri pentru proiectarea tip, (Researches and Experimentations regarding the Slenderness of the Reinforced Concrete Shear-wall Webs with Boundary Transverse Connections. Report with conclusions. Supplement of Design Provisions and Proposals for the Design of the Type Buildings) – INCERC – LSC, Bucure�ti.

4. CADAR, I., CLIPII, T., TUDOR, A. (2004), Beton armat, ediia a II-a, (Reinforced Concrete),Editura Orizonturi Universitare, Timi�oara.

5. KISS, Z., ONET , T., (1999), Beton armat, (Reinforced Concrete), u.t. Pres, Cluj-Napoca. 6. PARK, R., PAULAY, T., (1975), Reinforced Concrete Structures, New York Wiley Interscience. 7. TRIFA, FL. S., PRADA, M. (2000), Metod� de calcul a capacit��ii portante a sec�iunilor diafragmelor

din beton armat monolit, (Calculation Method of the Bearing Capacity of the Reinforced Concrete Shear-walls cross-sections), Analele Universit�ii din Oradea, TOM III., Fascicula Construcii �i

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Instalaii Hidroedilitare. 8. TRIFA, FL. S., (2003), Metod� de calcul a deplas�rilor postelastice ale diafragmelor pline din beton

armat, (Calculation Method of the Inelastic Displacements of the Reinforced Concrete Shear-walls), Analele Universit�ii din Oradea, Fascicula Construcii �i instalaii hidroedilitare, vol VI.

9. *** STAS 10107/0-90, Calculul �i alc�tuirea elementelor structurale din beton, beton armat �i beton precomprimat, (Calculus and Detailing of the Reinforced Concrete and Prestressed Concrete Structural Members).

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CLIMATE CHANGE INFLUENCE ON HYDROTECHNICAL STRUCTURES, EXISTING AND FUTURE

TROFIN Florin,

Military Technical Academy Bucharest, e-mail: [email protected]

A B S T R A C T Article approach a very topical issue and little studied: the influence of climate change on existing and future hydro facilities. Climate change may induce changes related to the stresses they are subjected to the resistance of hydraulic structures. This article is a bold attempt to address the issue of climate change impacts on existing infrastructure and design, implementation and operation of infrastructure to be designed in future.

Keywords: global warming, water resources, freeze thaw action, cracking

Received: January 2012 Accepted: January 2012 Revised: March 2012 Available online: May 2012

INTRODUCTION

Global warming, felt especially in the last decade of the twentieth century and early twenty-first century generated an increase of the extreme weather events. According to forecasts studies, it is estimated that over the next 100 years, global warming will lead to the extreme weather phenomenon, with implications to infrastructure engineering. Under these conditions forecasted of climate change, infrastructures rehabilitation engineering a topical issue giving that requires adequate design and impact assessment. Numerical models available coupled with the contemporary computational capabilities make it possible to engineers to forecast complex situations due to extreme events. Unfortunately, worldwide, are very few case studies to approach such issues. That is why this paper represents a bold to attempt the issue of climate change impact on the existing and future hydro technical infrastructure and design.

MATERIALS AND METHODS 1. The influence of climate change on water resources management

Climate changes are already happening. These affect natural water resources and also water users too. As a result, many countries, including Romania already have developed new strategies to adapt to these changes. Studies on influence of climate change, including those who carried out several river basins from Romania, have shown a reduction of the average annual flow, increasing flow in December - January period, maximum flow increases during summer months, reducing the thickness and length of layer snow due to a temperature increase during winter months [1].

The influence of climate change on water user is less studied; fact is the water requirements will certain increase in the near future. Adapting measures aimed to achieving these conditions, lead to a new balance between the existing water resources and user needs. This balance can be achieved by acting both on increasing water resource and water need, within the meaning of their diminution. It is obviously those measures are aimed to adopt a new water management approaches.

It’s known that socio – economic water resource is that part of natural water resources, which by engineering infrastructures, is available for consumer use. These infrastructures are represented by surface water intakes, dams, reservoirs, adductions, etc.

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2. The influence of climate change on existing hydro technical infrastructure It is estimated that over the next 50-100 years, global warming will have serious influence on

infrastructure engineering construction [2]. The issue is how we specialists can respond to these potential effects of global warming,

specific to hydro technical infrastructures? It is said that the main problem that arises is if engineers, can adapt to climate changes effects with responses studied, or will offer randomly responses to ongoing events. Some engineers might say is nothing but classic adaptation of risk management, only that their management is amplified by the uncertainties associated with climate changes.

It is known that one of the most significant impacts of climate change is the modification of hydrological cycle. Extreme events generated by climate change, such as floods or droughts, will become more intense on extensive areas. There will be a change in the probability of overcome the size of the verification flow for dams, dikes and other engineering infrastructures, for which they were designed. As a result, will be needed further more hydrological studies to establish new figures for flows and volumes of flood wave, which correlated with risk analysis, to lead to new solutions for a safe evacuation of these floods. Some dams would require additional arresters of large waters, and also changes in regulations to operating the reservoirs. The same issue arises in the case of flood protection embankments, to whose size should be reviewed.

Uncertainties in measuring the maximum flow process will also require changes in measurement technique. Changing hydrological cycle due to global warming will generate increased levels of seas and oceans, with estimated values from 20 to 50 cm, also more severe hurricanes and storms, with a higher influence on marine hydraulic infrastructures, as well as other infrastructures located on the coastal area such as coastal roads and railways, will be relocated.

Also, it has been estimated a reducing of intake and sediment transport in watercourses, in deepening of their minor riverbeds, with serious consequences on hydro technical works.

Increasing temperatures will lead to appearance and developed of cracks and fissures in the body of dam and dykes. Extreme events such as floods occur in small basins, will become more frequent. In conclusion, is necessary a prudent approach to climate change influence on the infrastructure engineering, so predictions of extreme events should be set in planning, design and rehabilitation of hydro technical structures.

3. Climate change influence to design, implementation and exploitation of new future hydro

technical structures Worldwide, it is estimated that climate change will be responsible for approximately 20%

water resources reduction. To reduce the effect of global weather changes, but also to satisfy the global water consumer needs, were developed scenarios such as the need to increase water volumes in reservoirs, with 9 to 30% of the current volumes [3].

These scenarios lead Romania to achieve over the next 40 to 50 years, hydro technical reservoirs with volumes between 0.7 to 2.4 km3. Worldwide, held a series of seminars on topics such as “Dams for a Changing World” or “The role dams in adaptation measures”. Unfortunately, these seminars have not tackled the issues of the kind referred to this article.

It is assumed that in the next 50 to 100 years will not be cause spectacular evolutions in the hydro technical structures design [9]. Significant developments could occur mainly to construction and operation of these structures.

Growth of air temperature and extreme flows are factors which should be considered when are designing hydro technical structures. For example, the arch dams will need in – depth studies regarding temperature variation effects, temperature distribution, both from the air and reservoir.

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The delimiting areas where arch dams can be placed should be characterized by average temperatures from 5 to 10° C upstream, and amplitude oscillations in the air between 12 to 15° C. Similar issues are also available to dams build from concrete piles and slabs.

Probably in future, the number of the inflatable dams and small earthen dams, placed locally on certain valleys or lowlands, for temporary accumulation of water from the rainfalls purpose, will extend. As a result, water supply systems will require new additional reservoirs for water storage, in order to compensate the increasing consumer needs, and water supply systems will be necessary new interconnections.

4. Other extreme events generated by climate change influence on hydro-technical construction 4.1 Aridity

Aridity may reduce the amount of water stored in reservoirs, degradation of water quality in reservoirs, temporary out of service of dams, fairway failure, less water for irrigation systems, soil erosion which led to an increasing amount of silt and clogging in reservoirs.

Dam cracking is just a symptom and not an actual degradation process, usually accentuated because of "stress" induced by thermal gradients[8]. Cracking may be given by the reactions of alkali or concrete contractions, but, in most of cases is also linked to design errors and initially contraction of concrete or special feature of the soil. It may also be a result of hydrostatic loading and thermal cycles. Real-time tracking of dam behavior revealed the emergence of cracks on the vertical buttresses, some of them having a complex (fig. 1).

Fig. 1. The emergence of cracks on the vertical buttresses

Generally, cracks are due to the contraction of the concrete and heat exchange, but also, local conditions or an inappropriate use of technology can increase intensity of the cracks. Seasonal temperature variation and levels of water retention can cause also slow process of deterioration generated by cracking (fig. 2).

Fig. 2. Process of deterioration generated by cracking

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However, these cracks may not affect the response of external loads applied to monolithic dam, but the cracks filling and restoration of concrete buttresses degraded can improve the stability of the structure [4]. In operation, the dam structure degraded by cracking, induce more sensitivity to prolonged or repeated efforts. Actually, cracking process does not affect directly the dam structure, unless there are difficulties in transmitting shear efforts. Laboratory experiments made on an excerpt from a haunch cracked, extracted by logging, to tracking the water pressure effects in the downstream direction, showed that upstream-downstream cracks propagation to such kind of external efforts, is rarely enough. Specialist’s trend is to overestimate the cracking effect. Usually, the shear efforts can have bad influences on the structure safety.

4.2 Rainfalls

Extreme amounts of rainfall may have the following effects on hydro technical structures: additional volumes of water entering the reservoirs, major damages to tipper waters and bottom drains, dam slopes and dyke damages, increase infiltration flow, sliding slopes or banks of lakes as a result of raising groundwater, damage or destruction of hydraulic groundwater capture, increase infiltration flows, increased risk of dykes and dam discharge, generating and amplifying phenomena of sliding slopes or banks of lakes, as a result of raising groundwater, damage or destruction of hydraulic groundwater capture, out of use of meteoric water discharge networks, damage of water collecting works and transportation, deterioration and damage of river beds regularization works, land degradation by erosion, landslides or water stagnation phenomena [10]. 4.2.1. Influence of water discharge

Exceeding the limit of safety thresholds in terms of precipitation may result in discharge water over the dam canopy. The stability problems are given by the increasing permeability and destructive action of currents and waves.

Fig. 3. Types of dam damages

4.2.2. Influence of water growth infiltrations

Dam built-up by local materials, strong infiltration downstream slope may be due to a high piezometric level. These can lead to deep slide (fig. 4).

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Fig. 4. a) Piezometric level inside the dam body; b) Gliding on to downstream facing

To increase the hydro technical constructions safety, are required:

- vertical drain to routing the infiltrations to filter based on slope; - filter on the old contour downstream embankment; - thickening downstream slope with gravel, to increase stability and filter protection

(fig. 5).

Fig. 5. Repairing of Downstream slope

4.2.3. Influence of hydro geological conditions changes

When using clays, hydro technical construction safety can be affected by changing hydro geological and geotechnical conditions. Such situation may occur to the central and downstream area of the dam. Foundation layer consists of marl, superimposed on a layer of clay altered (fig. 6).

Fig. 6. Initial cross section of the dam

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Dam consolidation can be done by putting back into operation the drainage network to the dam foot, from upstream and downstream, also putting into work the material and compacting with a roller tire, and execution of scarifications, with purpose to remove the structure defects (fig. 7).

Fig. 7. Strengthening the dam

In this case, it is necessary to use a measuring and control devices network, to tracking the

dam areas of weakness. Profile of the dam can be modified with purpose to increase the stability coefficient. 4.2.4. Frost phenomena

Frost phenomena usually generate ice-thaw, additional efforts on hydro technical structures, degradation to protection structures and improvement of a river channel, degradation of concrete structures (cracking) under freeze-thaw cycles. 4.2.5. Deterioration due to freeze-thaw action

In cold climates, dissolution of concrete constituents can combine effects occurred with those generated by freeze thaw action, with extremely rapid destruction of poor quality concrete [7]. Under the action of freeze – thaw cycles, concrete is deteriorating while the water content of structural defects (voids, crazing), exceeds the correspondent threshold of saturation, and ambient temperature is below 0 degrees C. This type of damage is manifested especially to hydro technical works located at high altitude to the old cold climate countries.

The main causes are given by multiplication of cycles freeze - thaw to concrete wetted, characteristic to cold climates. The frost effect is fast acting where the structures are more fragile. Crest of wave is also a subjected to freeze-thaw action, but that does not compromise the security of structure. To dams where used appropriate concrete and additives, resistance to this type of aggression has been increased considerably. Generally, frost action does not result in significant degradation of structure.

Motivation: - Submerged upstream of the dam is not subject to frost action; - Dams subjected to severe winter weather conditions may have a seasonal operation, so that

the effect of number of freeze-thaw cycles on the building is significantly reducing; - If the retention varies less, attacks on structures are localized to marginal areas. Freezing action of meteoric waters to downstream facing lead to a concrete exfoliation,

without significant effects to dam structure, still this action facilitates the appearance of vegetation and also degradation on the concrete surface. An inefficient drainage can cause dripping on the downstream facing, with significant negative effects. It can also cause damages to leaking valves or lower water area. In time, damages caused by freezing can increase, if accompanied by dissolution of concrete or alkali reactions. The most exposed hydro technical construction to freeze-thaw cycles are the arch dams build from concrete elements, with relatively small thickness compared to gravity

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dams [9]. Thermal regime extremely negative may lead to cracking of arches based (fig. 8 and Figure 9).

Fig. 8. Plan and elevation of a multiple arch

Fig. 9. Different types of cracks: 1 – parallel cracks; 2 - top-down oriented cracks;

3 - oblique cracks; 4 –concrete-rock contact damages Development of cracks can lead to an increased infiltration flow of drainage system of arcs.

4.2.4.1. Damage of dam protection masks and upstream Protection of upstream dams from local materials is achieved in most of cases, with masks of

concrete, bitumen concrete or membranes. Fragile elements of these sealing systems are joints between tiles and mask-spur connection. In time, or because changes in operational tasks, these may be affected by subsidence caused by:

- loads from ice-thaw pressure exerted; - inadequate compaction of breast body; - deformability of dam foundations; - earthquakes.

Strengthening the damaged masks can be done by: - restoration of dam joints with mastic asphalt or bitumen; - restoration of dam mask surface with mastic asphalt or bitumen; - building a new dam mask;

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- Execution of synthetic membranes. Upstream embankment filling causes mask damage and deterioration, but these incidents do

not directly affect dam foundation. We conclude that: a) Absence of the upstream protection to dams built from rock fill can produce superficial

erosions padding and fragmentation. Surface erosion is not dangerous, is easy to fix, but disorganization of filling and its protection due to the formation of ice lenses in the middle or near the level filling retention, can induce under certain conditions of diffusion, the effect of granulometry.

b) Protections upstream of the dams, which is acting as sealing, can be executed from rock fill or pitching. Rock fills can be weathered by waves, in terms of physics - chemical or geometrical of the riprap blanket. The pitching is more susceptible to such attacks, because the moderate down-grade breast can increase water stagnation and vegetation occurrence. Ice pressure may also generate harmful effects. The uplift pressure, in case of large waves, as well as damaged pitching, can foster bleeding in rock fill, being a harmful factor regarding its stability [5].

c) Facing defacement may affect stability of the breast during the time. These have a similar action as thin concrete dam. First undermining figure out to joints or to the improperly compacting concrete, due to implementation difficulties in work, of high hydraulic gradients and to concentrate efforts where excessive mound movement. Rock fill dam with moraine core are often built in the Nordic countries. Defects caused by the wave action on the upstream slope are presented to Figure 10.

Fig. 10. Rock fill arch dam upstream face gliding due waves action

The rehabilitation resource to restore dam tightness consist in execution of grout curtain on

the crest of wave [6]. The stability of the dam can be increased by thickening downstream mound. Repairs and reinforcements needed are presented to Figure 11.

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Fig. 11. Repairs and reinforcements to upstream face of a rock fill arch dam

CONCLUSIONS

Hydro technical construction safety is an important issue which should increase the designers and specialists attention. Any potential crash of these structures could have serious effects, similar as those caused by large natural disasters. Therefore, the risk assessment should be done with utmost responsibility and revised according to the new present - day weather climate.

Safety and risk concepts concerning hydro technical construction management are inseparable elements of the engineering design and operation processes. In conditions of climate change, the ratio between these two elements changes with the effect of decrease risk safety and increase operational. Therefore, is necessary to monitoring the climate change, induced by extreme events, such as:

- Changing geological conditions of the site; - Response of hydro technical structures to operating efforts; - Increase of infiltration flow through hydraulic structures; - Changing takeover flow conditions downstream of hydraulic structures (exceeding the

carrying capacity of the riverbed, the erosion of riverbeds, lower water deterioration of ecosystems);

- Degradation structures to cracking efforts; - Need to update data for design and operation assumptions; - Need to restore the operating regulations and regulations for the management of emergency

situations; - Need to review the warning systems and emergency situations alarm; - Decrease the safety and increase operating risk of hydro technical constructions.

REFERENCES 1. CR�CIUN I. GIURMA I., GIURMA-HANDLEY R. (2009), Quality Risk Evaluation of the

Groundwater Resources on the Moldavian Area, Environmental Engineering and Management Journal, Vol. 8/2009, no.3, 391-395, ISSN 1582-9596.

2. ANTOHI C.- M., TELI�C� M., GIURMA I., GIURMA-HANDLEY C.R. (2007), Climatic Changes – Research Hypothesis, International Conference „Monitoring of Disasters and Pollution”, IC.DMP03, 01 - 02 November, Iassy, Section II “Pollution”, pp.155-158.

3. SOLOMON S, D. QIN, M. MANNING, Y. CHEN, M. MARQUIS, K.B. AVERZL, M. TINGOR, and H.L. MILLER (2007), The Physical Science Basis, Contribution of Working Group I to the Fourth

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Assessment Report of the Intergovernmental Panel on Climate Change, Cambridge University, United Kingdom, and New York, USA, IPPC, 2007b, Climate Change.

4. DORIN POPA, IOAN IENCIU (2006), Studiu privind p�trunderea apei în betoanele folosite în construc�iile hidrotehnice (Study regarding the penetration of water into concrete used to hydrotechnical structures), RevCAD, r.6, 56-59, Alba Iulia.

5. ICHIM, I., R�DOANE M. (1988), Efectele barajelor în dinamica reliefului (Dam effects in the dynamics landscape), Editura Academiei.

6. IONESCU, �T. (2001), Impactul amenaj�rilor hidrotehnice asupra mediului (The environmental impact to hydro technical facilities), Editura H.G.A., Bucure�ti.

7. GIURMA I., CR�CIUN I.,GIURMA C.-R. (2006), Hidrologie (Hydrology), Ed. Politehnium, Ia�i. 8. *** ICOLD (1980), Dams and environment, Bulletin 35. 9. *** ICOLD (1989), Dams safety – guidelines, Bulletin 74. 10. *** ICOLD (1993), Dams and environmental – geophysical impacts, Bulletin.

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AUTHORS INDEX

C�t�rig Alexandru Ph.D.Prof.Eng., Technical University of Cluj-Napoca, Romania, e-mail: [email protected]

Cîrstolovean Ioan Lucian Ph.D. Lecturer Eng., University Transilvania Brasov, Faculty of Building Engineering, Romania, e-mail: [email protected]

Chira Alexandru Technical University of Cluj-Napoca, Romania

Constantinescu Horia Ph.D.Student.Eng., Technical University of Cluj Napoca, Department of Structures, Romania, e-mail: [email protected]

Dârmon Ruxandra Teaching Assistant Eng., Technical University of Cluj Napoca,Romania, e-mail: [email protected]

Domni�a Florin

Ph.D.Lecurer.Eng., Technical University of Cluj-Napoca, Building Services Faculty, Romania, e-mail: [email protected]

Dub�u C�lin Gavril

Ph.D. Lecturer Eng., University of Oradea, Faculty of Environmental Protection, Romania, e-mail: [email protected]

Ho�upan Anca

Ph.D. Teaching Assistant Eng., Technical University of Cluj-Napoca, Building Services Faculty, Romania, e-mail: [email protected]

Jumate Elena Ph.D. Student Eng., Technical University of Cluj Napoca,Romania, e-mail: [email protected]

Kopenetz Ludovic Gheorghe

Ph.D.Prof.Eng., Technical University of Cluj-Napoca, Romania, e-mail: [email protected]

Li�man Drago� Florin

Ph.D. Student Eng., Technical university of Cluj Napoca, Romania, e-mail: [email protected]

Lupan Lidia-Maria Technical University of Cluj-Napoca, Romania, e-mail: [email protected]

Manea Daniela Lucia Ph.D. Prof. Eng., Technical University of Cluj-Napoca, Faculty of Constructions, Romania, e-mail: [email protected]

M�gurean Cornelia Ph.D.Prof.Eng., Technical University of Cluj Napoca, Department of Structures, Romania, e-mail: [email protected]

Moga Ioan Ph.D. Prof.Eng., Technical University of Cluj-Napoca,Faculty of Civil Engineering, Romania, e-mail: [email protected]

Molnar Iulia

Ph.D. Student Teaching Assistant Eng., Technical University of Cluj-Napoca, Faculty of Civil Engineering, Romania, tel.:0040744 875 191 e-mail: [email protected], [email protected]

Mo�oarc� Marius Ph.D. Lecturer Eng., “Politehnica” University of Timisoara, Faculty of Architecture, Romania, e-mail: [email protected]

Peredi �tefan Technical University of Cluj-Napoca, Romania, e-mail: [email protected]

Pintea Augustin Technical University of Cluj-Napoca, Romania, e-mail: [email protected]

Popovici Tudor

Ph.D. Prof.Eng., Technical University of Cluj-Napoca, Building Services Faculty, Romania, e-mail: [email protected]

R�dulescu Adrian T.G. Technical University of Cluj-Napoca, Romania

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R�dulescu Corina Technical University of Cluj-Napoca, Romania

R�dulescu Gheorghe M.T. Technical University of Cluj-Napoca, Romania, e-mail: [email protected]

R�dulescu Virgil Mihai G.M. Technical University of Cluj-Napoca, Romania, e-mail: [email protected]

Stoian Valeriu Ph.D. Prof. Eng., “Politehnica” University of Timisoara, Civil Engineering and Equipment Department, Romania, e-mail: [email protected]

Trifa Florin Sabin

Ph.D. Student Lecturer Eng., University of Oradea, Faculty of Constructions and Architecture, Department of Constructions, Romania, e-mail: [email protected]

Trofin Florin Ph.D.Student, Military Technical Academy Bucharest,Romania, e-mail: [email protected]

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