ENGINEER - Vol. XXXVIII, No. 03, pp. 62-69, 2005© The Institution of Engineers, Sri Lanka

Geotechnical Design of Large Diameter Impact DrivenPipe Pile Foundations: New East Span

San Francisco-Oakland Bay BridgeS. Mohan1, R. E Stevens2 and B. H. Maroney3

Abstract: The San Francisco-Oakland Bay Bridge is one of the most heavily travelled bridges in theworld. The east span of the bridge will be replaced due to seismic safety concerns. The new bridge willbe founded mostly on large 1.8 to 2.5-metre diameter, approximately 60- to 100-metre long pile foundations.Pile foundations will experience tension loads of up to approximately 90 MN and compression loads ofapproximately up to 140 MN during the design earthquake. This paper discusses the geotechnical designapproach and techniques followed to evaluate the soil-pile-setup with time and the axial capacity oflarge diameter impact driven pipe piles for the project.

Keywords: Impact hammer, pile driving, bay mud, soil-pile setup, and acceptance criteria

1. Background

The San Francisco-Oakland Bay Bridge (SFOBB)is the primary link between the cities of SanFrancisco and Oakland and carries 10 lanes oftraffic across San Francisco Bay in California,USA. Yerba Buena Island (YBI) bisects the bridge,longitudinally, with the west span extending fromSan Francisco to YBI and the east span extendingfrom YBI to Oakland, Figure 1. The existing eastspan bridge is a 3.5-km-long, double-deckedstructure that was constructed in the 1930s. Theeast span of the SFOBB, which is one of the mostheavily travelled bridges in the world, will bereplaced due to seismic safety concerns. Theproposed east span bridge will be constructedalong a parallel alignment to the north of theexisting bridge.

The new bridge will consist of 5 segments: (1) anapproximately 460-metre-long transitionstructure extending from the YBI Tunnel to theeastern tip of YBI, (2) main-span signaturestructure extending offshore from the tip of YBIwith an approximately 625-metre-long, single-tower that rises 164-metre above the water andasymmetrical, self-anchored suspension (SAS)cable, (3) an approximately 2.1-kilometre-long,four-frame Skyway structure extending from thesignature structure eastward to the Oakland ShoreApproach, (4) an Oakland Shore Approachstructure extending about 700 metres from theSkyway structure to the north side of the OaklandMole, and (5) an earthen fill transition from theOakland Shore Approach structure to theroadways leading to and from the existing bridge.

SANFRANCISCO

SAN FRANCISCOBAY

Figure 1. Site Vicinity Map

'Eng. S. Mohan, P. Eng., C. Eng., MIE (SL), M.ASCE, B.Eng. (Hans)(UK), M. Sc. (USA), Presently. Senior Project Engineer/Associate CommaManager, Department of Transportation, California, USA, & CEO/Chairman, New-Frontiers (Pvt) Ltd, Developers & Civil EngineeringConsultants, Colombo, Sri Lanka.

2(Dr.) R. Stevens, P. Eng., F.ASCE, B.Eng. (USA), M. Sc. (USA), PhD(USA), Presently, Senior Technical Manager, Fugro-McClelland MarineGeosciences, Inc., Ventura, USA.

3(Dr.) S. Maroney, P. Eng., M.ASCE. B.Eng. (USA), M. Sc. (USA). PhD(USA), Presently, Principle Engineer/Toll Bridge Program Manager,Department of Transportation, California, USA, & Professor, Departmentof Civil & Environmental Engineering, University of California, Davis.USA

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riven

[ges in the•ridge willundations.m loads ofical designapacity of

2nts: (1) an:ransitionmel to thesignaturetip of YBI

ng, single-water andsion (SAS)netre-long,\g from thedand ShoreApproachs from thele Oaklandm from theire to theting bridge.

, B.Eng. (Hons)satiate ContractUSA, & CEO/

vil Engineering

Sc. (USA), PhDClelland Marine

Sc. (USA), PhDtgram Manager,fsor, Departmentalifomia, Davis.

It SUf-Specific Geology

nnutions that underlie the skyway.... . . . . I n * ! ' i n i l i i ' ln l l invingsequence: Young Bay

, M*'i nil I \ >sey-San Antonio formations, OldMtii l , t l|i|H-r Alamcda Marine sediment and

i i , , I i .wr i Alnmcda Alluvial sediment. Table 1i fig 2 provide key engineering soil properties

•ml an llltiili'iition of the subsurface conditionsthe new bridge alignment.

The Self-Anchored-Suspension (SAS) signatureportion of the bridge is supported by three piersupports. The West Pier-W2 and main Towerfoundation supports are situated in the area ofshallow, steeply sloping bedrock on and adjacentto YBI. West Pier W2 consists of two supports,one for eastbound and the other for thewestbound direction. The two W2 foundationsare 19 m x 19 m mass gravity foundations with athickness of 10 m and will be constructed in rock.

Table 1. Engineering Soil Properties

< . IM.I.IKI.H! rorm.Uion

YVM: vwy toft to firm ClayMI'SA 1 i r i i -M- ID Vrry Dense Sand with Stiff toVary Stiff City Layers

( >HM/U AM Sediments: Very Stiff to Hard Clay

I .A A Sriiimi-nts: Dense to Very Dense Sandmid I lard Clay

Friction'Angle,

Degrees-

35-42

-

40 +

UndrainedShear

Strength, kPa10-65

60 -175

100 - 250

225 - 400

SubmergedUnit Weight,

', kN/m3

4-7

6-11

6-9

9-12

>|«.i'ii.7iifflTte»[rt *|tj ffMAMCIttGAN roMMATIOM f*"

STRATK3RAPHYAL.QNG BR1DOE

ALIGNMENT

in

Structure Im»ge«J In

LITMOl̂ CMSV,•MMAM W*V|g

VKL.OCITV AMO «

,;'

, i

35;-

I

Figure 2. Interpreted Stratigraphic Section Along New Bridge Route

I, Proposed Foundations

1 1 ii proposed bridge foundation types are largelyi i .cil on the Geotechnical and geological• >,mill ions at the site. Along the new bridgealignment , the bedrock surface is relatively(hallow and dips steeply near the shoreline of YBII l i f i i Htceping to the east down to elevation of -1 1 H ) nu-lres at about 350 metres offshore from YBI.Hit- k-drock contact then slopes gently toward

At the Oakland end, the bedrockare buried by about 135 to 140 metres of

npiliments.

Excavation of the rock is currently underway. Thewestbound footing includes four 2.5 m diameter,10 m long cast-in-drilled-hole (CIDH) shaftssocketed in slightly weathered to fresh rock whilethe eastbound footing is placed directly on rock.The westbound footing is embedded in steepbedrock slopes and the CIDH shafts providelateral shear capacity.

In contrast, the SAS-East Piers and the Skywayand Oakland Shore approach portions of thebridge piers are underlain by over 85 to 135 metresof marine and alluvial sediments. SAS-East Piers

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2E and 2W foundations consist of (12) 2.5-mdiameter steel pipe piles up to 100 m long.Skyway eastbound and westbound Piers 3through 16 consist of (160) 2.5-m diameter pilesup to 100 metre long. Each pile will experiencetension loads of up to approximately 80 to 90 MNand compression loads of up to approximately120 to 140 MN during the design earthquake.Each of these piers will be supported by anoctagonal or retangular pilecap with 4 to 6 piles,at 1:8 batters.

The Oakland Shore approach eastbound andwestbound Piers 17 through 22 consist of 12 pierswith 104 steel pipe piles and 1.8 m in diameter.Each pile will experience tension loads of up toapproximately 9 to 30 MN and compression loadsof up to approximately 18 to 42 MN during thedesign earthquake. Each of these piers will besupported by a retangular or square pilecap with8 to 9 piles. Prior to pile driving, a cofferdam willbe constructed at each pier location (except Piers2 through 6), the pilecap footing box will be placedinside and the piles will be driven through thefooting box with the use of a 1700 kilo-jouleMenck MHU-1700 hydraulic impact hammer.

4. Axial Pile Capacity

4.1 Skin Friction

Since, the selected pipe pile diameters wereoutside the normal range of onshore/off shorepiles and to reflect the site-specific soil conditions,the modified American Petroleum Institute (API)method was used for calculating axial capacities.The API design procedure (1993) for calculatingaxial pile capacity in clay soils was adapted froman Offshore Technology Conference paper byRandolph and Murphy (1985) in which the unitskin friction transfer is calculated using therelationship:

/C V/2 /S«-(?) fc V2

Where: a-adhesion factor, Su-undrained shearstrength at any depth, a'-effective overburdenstress at any depth, c-undrained shear strengthand p'-vertical effective overburden pressure

The API adopted a lower-bound value of 0.25 forthe c/p' ratio for Gulf of Mexico clays, resultingin a default value of 0.5 for the (c/p')1/2 term Since,

the large-diami'ter ovrrwiilri pile Inundationsderive much of the skin I rii'lion i ,i|i,idty from theclay layers of the Old Bay Mud and the UpperAlameda-Marinc formations, tin.1 applicability ofthis assumption to the conditions at the baybridge site was researched in more detail.

The results of static pile load tests performed onpiles supported in Young Bay Mud wereevaluated. In those tests the ultimate side sheartransfer in the pile load tests were estimated tobe equal to the undrained shear strength.Hindcast analysis to match the observed staticload-settlement behavior of the pile resulted inan estimate of the undrained strength ratio (c/p') of 0.31. A number of, Ko Consolidated-Undrained Triaxial Tests and Direct Simple Sheartests were performed on soil samples as a part ofthe marine site characterization for the project.The results of those tests also suggested a c/p'value on the order of 0.31.

Based on available site-specific data, the designmethods presented in API (1993a,b) weremodified by increasing the value of the implicitc/p' ratio in API from 0.25 to 0.31. The ultimateunit shear transfer in the clay strata was thencalculated as:

(s V1/2 q^j for -^ < 1.0

(s \-v* s-£ \ for -g > 1.0

/

These formulations result in axial skin frictioncapacities in clay that are approximately 11percent greater than those given by the standardAPI formulation.

4.2 End-Bearing

In order to develop estimates of end-bearingcapacity, a statistical evaluation was performedto evaluate the probable presence of clay layersat the pile tip. To account for the potentialpresence of clay layers, values for the end-bearingcapacity were developed using weightedaverages based on the relative amounts andthickness of the interbedded clay layers. Thevalue of the unit end-bearing pressure (q ) foreach frame was calculated as:

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3.24 11. 97 X

100

Where: X-percentage of clay layers and Y-percentage of sand layers

4.3 Soil-Pile-Setup and Axial Capacity

In order to evaluate the soil-pile-setup and theaxial capacity of a large diameter impact drivenpipe pile with time, a series of CAPWAP analysiswere done and compared with upper- and lower-bound soil-pile-setup predictions based onSoderberg (1962) method and API predicted axialcapacity. A combined "best estimate" skin frictiondistribution was obtained by summing the largestmobilized skin friction increment values along thelength of the pile during initial driving orsubsequent re-strike(s). It is important torecognize that the largest mobilized skinresistance increment may come from theCAPWAP analysis performed at the end of re-strike [8]. This is due to the fact that subsequenthammer blows will breakdown the setup in theupper portion of the pile and mobilize skinresistance in the lower portion of the pile that wasnot mobilized at the beginning of the restrike.

A series of restrikes were conducted on the pileinstallation demonstration project (PIDP) to betterunderstand the magnitude and distribution of soilresistance along the pile. Three restrikes wereconducted on Pile-1, three re-strikes wereconducted on Pile-2, and two re-strikes wereconducted on Pile-3. Based on CAPWAP analyses,the trend of increase in total skin resistance withtime is illustrated in Fig. 3. For comparison, totalskin resistance profiles computed, based on APIdesign method (1993), is shown in Fig. 3.

Fig. 4 presents the total skin resistance at each re-strike versus time. It appears that after 33-daysof set up, Pile-1 has the static skin resistancecapacity of 70 MN, which is approximately 88percent of design skin resistance capacity. For Pile2, the total skin resistance capacity wasapproximately 67 MN after 22 days of set-up. Itappears that after 23 to 24 months of set up, allthree piles have the static skin resistancecapacities of 78 to 88 MN, which areapproximately 98 to higher than 100 percent ofdesign skin resistance capacity. However, due tothe presence of predominantly clayey soils at tin-site, it is anticipated that soil pile setup willcontinue for several months. Also, it is anticipated

End of Initial Driving, PDA 1 (Nov. 9, 2000)

End of Restrike 1, PDA 2 (Nov. 11, 2000)

Beginning of Restrike 2, PDA 2 (Nov. 20, 2000)

Beginning of Restrike 3, PDA 2 (Dec. 12, 2000)

Combined "Best Estimate"

Calculated Unit Skin Friction Profile, API (1993a,b)

Note: The CAPWAP analysis for Restrike 1was performed at the end of the restrike dueto relatively poor hammer performance/energytransfer at the beginning of the restrike.

0 4 10 20 30 40 50 60 70 80

TOTAL SKIN FRICTION (MN)

Fig. 3. Totiil Skin Resistanee, Combined CAPWAP Analysis

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that the skin friction capacity will exceed thecapacity of 80 MN predicted by the API designmethod.

5. Soil Resistance to Driving

Stevens et al. (1982) recommended that lower- andupper-bound values of soil resistance to driving(SRD) be computed for the coring pile conditionespecially for larger-diameter pipe piles. When apile cores, relative movement between pile andsoil occurs both on outside and inside of the pilewall. The lower bound was computed assumingthat the skin friction developed on the inside ofthe pile is negligible. An upper bound iscomputed assuming the internal skin friction isequal to 50 percent [3] of the external skin friction.For a plugged pile, a lower bound is computedusing unadjusted values of unit friction and unitend bearing. An upper bound plugged case forgranular soils is computed by increasing the unitskin friction by 30 percent and the unit end

bearing by 50 percent. For cohesive soils the unitskin friction is not increased and the unit endbearing is computed using a bearing capacityfactor of 15, which is an increase of 67 percent.For sandy soils, the unit skin friction and unit endbearing values that were used to predict the SRDwere the same as those used to compute staticpile capacity. For clayey soils, the SRD wascomputed using two methods.

5.1 Method-I

The unit skin friction was computed from thestress history approach presented by Semple andGemeinhardt (1981). The unit skin friction andunit end bearing for static loading is firstcomputed by using the method recommended bythe API (1986). The SRD is then calculated byreducing the unit skin friction values overincrements of depth by multiplying by a pilecapacity factor, determined empirically.

Fp = 0.5 x (OCR) °-3

110

Pile setup predictions are based on the Soderberg (1962) method.he predicted range of pile setup assumes:) Average clay sensitivity of 2.5,

2) 80% of skin friction capacity from clay layers, and!) Ultimate skin friction capacities ranging from 80 to 110 MN.

Estimated Total CAPWAPSkin Friction Capacity

Maximum Construction Load,Jack Span 4, Pier E8 (45 MN)

• file ^A Pile 2* Pile3

Allowable Pile CapacityFS = 1.5 (Skin Friction)

D Prlel

A Pile 2O Pile3

Estimated Pile Construction Load TimeHistory, Pier E8 (TY Lin/M&N, 2000)

0 100 200 300 400 500 600 700 800 900 1000Time (days)

Fig. 4. Estimates of PILE setup From PIDP

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The over-consolidation ratio (OCR) wascalculated with the following equations.

Su/Uunc = (CO?)0-85

Uunc = a1 vo x (0.11 - 0.0037 x pi}

where: Uunc- undrained shear strength-normallyconsolidated clay, c?'vo - effective overburdenpressure and Pi - plasticity index

The OCR can also be estimated from CPT tipresistances with the use of following equation.

= & p/o1 vo

a1 p = 0.33 x (qc - o1 vo)

where: qc = cone tip resistance and o' =preconsolidation stress

5.2 Method-II

Based on the "Sensitivity Method"; the unit skinfriction for static loading is first computed byusing the method recommended by API (1983,1993). The SRD is then calculated byincrementally reducing the unit skin frictionvalues by measured clay sensitivities, which isthe ratio of the undisturbed to remolded clayshear strengths.

The computed lower- and upper-bound SRDwere used to perform the wave equation analysesto predict the lower- and upper-bound acceptancecriteria. Wave equation analyses were performedusing GRLWEAP (1997-2) program for thecontinuous driving case. The soil quake anddamping parameters recommended by Roussel(1979) were used. The shaft and toe quakes wereassumed to be 0.25 centimetres for all soil types.A shaft damping value of 0.19 to 0.36 seconds permetre was assumed for clayey soil. The shaftdamping value in clayey soil decreases withincreasing shear strength [2]. The toe dampingvalue of 0.49 seconds per metre was assumed forall soil types.

Fig. 5 presents results of predicted and PDAobserved SRD profiles for Pile 2. Interestingly, thepredicted and observed blow count profiles, asexpected, correlate well with the predicted andPDA observed SRD. PDA observed SRD wascomputed based on maximum Case Method anddamping coefficient (J) of 0.5. The observed blowcounts and SRD spikes at penetration depths of

about 45 and 70m were as a result of soil pile set-up that occurred during driving delays such assplicing and welding of pile sections. Therefore,it demonstrates very well that wave equationanalyses can be used to reasonably predict theblow counts with the penetration depth providedthat similar hammer energies are applied in themodel.

Fig. 6 presents the results predicted by Case-Gobleformulation and wave equation analysesperformed based on the PDA measured drivingsystem performance data and observed field blowcounts for Pile 2. During continuous drivingobserved blow counts below a penetration of 35to 40 metres, tend to follow the lower bound ofpredicted blow counts based on Method-I and aregenerally bound by upper and lower boundsbased on the "Sensitivity Method" (Method-II).The sensitivity method seems to better predict theobserved blow counts in the soft Young Bay Mudsediments even in the upper 35 to 40 metres forPile 2. The observed blow counts spikes at certaindepths was as a result of soil-pile setup thatoccurred during driving delays such as splicing/welding of pile sections.

5.3 Acceptance Criteria

Piles driven to the required design tip elevationsand if meeting the coring case lower boundacceptance criteria for the given range of hammerenergy (hammer efficiency and field blow counts)will be accepted by either method. However, ifthe piles do not meet the lower-bound acceptancecriteria for the PDA-measured range of hammerenergy then additional restrikes and CAPWAPanalyses will be performed to evaluate thecapacities prior to accepting the piles.

If early pile refusal is met (generally 5 tolO mabove the specified design pile tip elevation,depending on pier location) but the designelevation for lateral capacity is reached, the pilecan be accepted if it met the following conditions.

• The specified primary hammer is operatingat full rated energy according to themanufacturer's specifications.

• Pile driving resistance exceeds either 250blows per 250 mm over a penetration of1500 mm or 670 blows for 250 mm ofpenetration.

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• If pile-driving operation is interrupted formore than one hour, the above definitionof refusal shall not apply until the pile hasbeen driven at least 250 mm following theresumption of pile driving.

• At any time, 670 blows in 125 mm shall betaken as pile driving refusal.

5. Conclusions

Load tests in San Francisco Bay soils have shownthat the unit side shear resistance on a pile canequal the undrained shear strength of thesupporting soil, whereas the average skin frictionvalues used in the static estimates generated usingthe modified API (1993) methodology can be onthe order of 70 percent of the undrained shear

strength of the surrounding soils. These load teststherefore indicate that the available side shearresistance may be as much as 40 percent higherthan that, which would be used for design. Thedata from the PIDP also suggests that skin frictioncapacity may exceed those predicted using theAPI procedures.

The combined CAPWAP analysis can be used toestimate the capacity of driven piles with timeand to proof-test the piles even if the hammerdoes not have sufficient energy to drive ormobilize the pile at their ultimate capacity. Thecombined CAPWAP analysis can also be used toestablish the soil-pile setup with time for theclayey soils. It will be valuable data during stagedconstruction in order to establish waiting periodsprior to loading the piles.

10

20

30

40

co'•a 50

0)CL

60

70

80

90

100

0 20 40 60 80 100

Soil Resistance to Driving (MM)

Fig. 5. PDA Measured and Predicted SRD

10

20

30

40

co'ra 5015c01Q.

60

70

80

90

100

-Observed Blow Counts

- - - Predicted Coring Case BlowCounts, Stevens et al. (1982)

-Predicted Coring Case BlowCounts, Sensitivity-BasedMethod

\

20 40 60 80Blows Per 0.25 Metre

100

Note: Blow counts were predicted using awave equation model with the hamm eroperated at the PDA-measured transferredenergies.

Fig. 6. Observed and Predicted Blow Counts

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Wave equation analyses can be used to predict

the blow counts with the penetration depth usingGRLWEAP program for the continuous drivingcase. Provided that piles are driven to the requireddesign tip elevations, piles can be accepted (byeither method for clayey soils) based on the coringcase-lower bound acceptance criteria for the givenrange of hammer energy.

References

1. American Petroleum Institute, 1986, 1993."Recommended Practice for Planning, Design, andConstructing Fixed Offshore Platforms". API RP-2A, API, Washington, B.C.

2. Coyle, H.M. and Gibson, G.C, 1970. "Soil DampingConstants Related to Common Soil Properties inSand and Clay". Texas Transportation Institute,Research Report 125-1, Texas A&M University.

3. 3. Mohan, S., Buell, R., Stevens, R. R, Howard, R.,and Dover, A. R., 2002. "Deepest Ever LargeDiameter Pipe Pile Installation demonstrationProject New East Span San Francisco-Oakland BayBridge", 9th International Conference on Piling andDeep Foundations, Nice, France.

4. Randolph, M.F. and Murphy, B.S., 1985. "ShaftCapacity of Driven Piles in Clay, Proceedings", 17thAnnual Offshore Technology Conference, Houston,Texas, Vol. 1, pp. 371-378.

5. Roussel, H.J, 1979. "Pile Driving Analysis of Large-Diameter, High Capacity Offshore Pipe Piles. PhDThesis", Dept. of Civil Engineering, TulaneUniversity, New Orleans, Louisiana.

6. Semple, R.M. and Gemeinhardt, J.P, 1981. "StressHistory Approach to Analysis of Soil Resistance toPile Driving. Proceedings", 13th OffshoreTechnology Conference, Houston, Texas, Vol. 1. pp.165-172

7. Soderberg, L.O, 1962. "Consolidation TheoryApplied to Foundation Pile Time Effects".Geotechnique, Vol.12, No.3, 465-481.

8. Stevens, R.F, 2000. "Pile Acceptance Based onCombined CAPWAP Analyses. Proceedings SixthInternational Conference on the application ofStress Wave Theory to Piles", Sao Paulo, Brazil.

9. Stevens, R.F., Wiltsie, E.A. and Turton, T.H, 1982."Evaluating Pile Drivability for Hard Clay, VeryDense Sand, and Rock. Fourteenth Annual OffshoreTechnology Conference", Houston, Texas, Vol.1,465-481.

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