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Chapter 1INTRODUCTION
1.1 General
In many countries, it is a common practice to construct RC frame building with open
ground storey (i.e. unlike other stories, no or scanty infill walls are provided in the
ground storey) in order to generate parking space, gardening space, and other utility
spaces for various purposes. Providing parking spaces in multi-storey buildings is an
essential requirement. Architect finds an easy solution by keeping ground storey open.
Also, the local municipal/building bylaw at many places supports/directs the same for
solving the parking problem. This is leading to a large number of open ground storey
building construction.
Generally, in open ground storey buildings, unreinforced brick masonry infills are
present in all floors except the ground story. This leads to severe stiffness and strength
irregularity and even sometimes leads to torsion irregularity. Buildings with these
irregularities has consistently shown poor performance during past earthquakes like
1999 Turkey, 1999 Taiwan and 2001 Bhuj, 2003 Algeria earthquakes and many
others. Normally, infill walls are considered as non-structural member; however,
practically it provides significant stiffness under lateral load. If special provisions
have not been followed in design, absence of infill at ground storey will lead to
formation of soft ground storey. Under lateral loading, lack of infill stiffness will lead
to larger inter-storey drift concentrated to ground storey leading to an early formation
of plastic hinges, further impending collapse of structure.
As per IS1893 (Part I): 2002 storey is considered as soft if its lateral stiffness is less
than 70% of that in the storey immediately above or less than 80% of the combine
stiffness of the three stories above. Also, an extreme soft storey is one in which the
lateral stiffness is less than 60% of that in the storey above or less than 70% of the
average stiffness of the three stories above.
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1.2 Literature Review on Behavior of RC Frame Building during Past
Earthquake
Past earthquakes have revealed the weaknesses of RC framed buildings and indicated
the points where improvements are required. In the following sub-sections some
damages to RC frame buildings have been highlighted including the damage and
behavior of open ground storey buildings and infill walls. Past earthquakes has
revealed the weaknesses of RC framed buildings and indicated the points where
improvements are required.
1.2.1 Seismic behavior of improperly designed, detailed and constructed RC frame
buildings
Due to ease of material handling and economy, the construction of RC frame
buildings in urban areas of the world is increasing day by day from last 4 decade.
Earlier, the main emphasis of design was for gravity loads, but the poor performance
of these gravity load designed buildings during earthquake has indicated the
importance of consideration of earthquake forces. In India, seismic codes are present
from last half decade, but due to lack of stringent guidelines and penalty, its
implementation in real construction is still not fully achieved. Gravity load designed
buildings suffers various types of damages during earthquake, following are few
commonly observed damages.
1) Brittle failure of RC column (Figure 1.1)
2) Short column failure (Figure 1.2)
3) Soft storey formation (Figure 1.3)
4) Weak column strong beam design (Figure 1.4)
5) Lap-splice failure (Figure 1.5 (a) and (b))
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Figure 1.2. Typical Earthquake Damage-Short Column Failure (1999 Athens Earthquake)
Figure 1.1 Brittle failure Photo from: Housner & Jennings, Earthquake Design Criteria, EERI, USA
Figure 1.3. Earthquake Damage- Dislodged Column due to Soft ground Floor effect (1999 Athens Earthquake)
Figure 1.4.The three-storyprimary school in Gedikbulak village after collapse (photos: Erdil) Turkey Earthquake
Figure1.5 (a).Lap splice joint fail Figure 1.5 (b).Closer view of lap splice failure of Figure 5 (a)
Takashi K. et al. (2000) reported that during Turkey earthquake of 17 August 1999, in
Kocaeli city which was also the epicenter of the earthquake, about 35839 residential
and 5478 workplaces RC buildings were heavily damaged / collapsed. Moderate
damage were observed in 41100 residences and 5861 work places, and slightly
damage in 45111 residences and 6122 workplaces all being constructed by reinforced
concrete. In another city Sakarya which is about 50 km from Kocaeli, number of
heavily damaged / collapsed RC buildings were 29844, moderately damaged were
22170 and slightly damaged were 26772.
After 12 January 2010 Haiti earthquake a survey 107 RC frame buildings conducted
by Eberhard et al. (2010) in Port-au-Prince downtown indicated that 28% had
collapsed and another 33% were damaged enough to require repairs. Another survey
conducted in Leogane city by the same author with 52 buildings, found that 62% had
collapsed and another 31% required repairs.
Issue of cumulative residual damage in buildings arises when the building is subjected
to high intensity multiple seismic loading during its designed life. The effect of
cumulative loading has been observed by Kam and Pampanin (2011) during two
consecutive earthquakes, first one on 4 September 2010 and second on 22nd February
2011. It has been observed that at Christchurch, the damage due to first earthquake is
confined to a moderate level, however, during the second earthquake many buildings
were severely damaged and about 135 buildings collapsed.
In India, the shortcomings of planning, design, detailing, and construction was
revealed during Bhuj earthquake. Bhuj earthquake of January 2001 was the first
earthquake in India which has affected urban area. Several thousand poorly designed
and constructed buildings were damaged and many of them collapsed. It was also
observed that the collapses of buildings were not limited only to epicentral region but
seen at Ahmadabad too, which is about 250 km away from the epicenter. In
Ahmadabad 75 RC frame buildings collapsed and several thousand were damaged in
and around the city, clearly demonstrating the seismic vulnerability of poorly
designed building (Jaiswal, et al.2003). In recent past, similar behavior of RC frame
buildings was also observed during Sikkim earthquake and Andaman earthquake.
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1.2.2 Seismic behavior of open ground storey buildings
Open ground storey buildings are those in which ground storey is kept open (free
from infill walls) to provide parking and other utility spaces in the building. In India
this is one of the most prevailing multi-storey construction practices. If open ground
storey buildings are improperly designed, it will have severe stiffness and strength
deficiencies. It has been observed during past earthquakes (1999 Turkey, 1999
Taiwan and 2001 Bhuj earthquake, 2003 Algeria earthquakes ) the damage in these
type of buildings are confined to the ground storey.
Earlier, this type of construction practices was advocated by various building by-laws.
The Taiwan government enacted a law to encourage contractor to construct buildings
with open ground storey (Tung and George, 2003). It was also instructed in the law
that ground storey height should be kept at least 5 m and in return the owners were
awarded with extra floor area.
Some common features buildings designed in Taiwan after the aforementioned law
was enacted are:
1. Generally ground storey was kept open with double height i.e. with a net
height approximately 7.6 meters (Figure 1.6 a).
2. Upper stories were having dense partition walls.
3. Due to presence of staircases and elevator shaft on the edge of the building
plan, torsional irregularity was also present.
4. Peculiar shapes of building plan were conceived by architect to provide good
outside view from every part of the building, natural lighting and ventilation.
5. To maximize parking space very few columns were designed into these
buildings at the basement. The primary load resisting system is reinforced
concrete moment resisting frame on a mat foundation.
The consequences of this law was revealed during 1999 Chi-Chi earthquake, where a
huge damage was observed in open ground storey buildings (Figure 1.6 b).
In India open ground construction is quite prevalent from last 25 years and its adverse
effect was observed during Bhuj earthquake. In Ahmadabad alone about 25,000, five-
storey buildings and about 1,500 eleven-storey buildings were damaged and about
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100 open ground storey buildings collapsed (Murthy 2005). Figure 1.7a and 1.7b
shows similar damage of open ground storey buildings during 2003 Boumerdes
earthquake and 2001 Bhuj earthquake, respectively. Further, in India, a large number
of similar open ground storey buildings exist in the various towns and cities situated
in moderate to severe seismic zones namely.
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Figure 1.6a. Double storey heighted open ground storey building in Taiwan (photo from World Housing Encyclopaedia Report)
Figure 1.6b. Collapsed Double storey heighted open ground storey building in Taiwan during Chi Chi earthquake 1999 (photo from World Housing Encyclopedia Report)
Figure 1.7a. Building collapse due to soft-story mechanism in the 2003 Boumerdes Earthquake (WHE Report 103, Algeria)
Figure 1.7b. Damaged RC frame building during Bhuj earthquake with damage confined to open ground storey (photo from EQ tips)
1.2.3 Seismic behavior of URM infilled frame
Unreinforced masonry (URM) infill walls are generally considered as non-structural
element. However, it has been observed that the behavior of infilled frame
significantly vary in comparison to bare frame under lateral loading. Modelling of
“Frame-Wall interaction” has remained a difficult task due to various reasons, such as
opening in wall, gap between wall and frame, and variation of material strength along
with significant increase in computational effort. A simplest modelling technique is
based on equivalent strut model. Masonry infill walls generally acts as a compression
strut when subjected to lateral loading as shown in figure 1.8. This model, initially
proposed by Poliakov consists of assuming that the effect of the infill panels can be
represented by introducing diagonal bars under compression. The existence of infill
walls can change the structural behavior from flexural action into axial action. Typical
failure mode of Infill wall and frame is shown in Figure 1.9. Failure modes include
corner crushing, frame damage, shear slip of wall, toe crushing, diagonal tension, etc.
The Advantages (Tabeshpour, et al., 2011) in the conversion of flexural action to axial
action are:
1) Reduce contribution of frame in lateral resisting
2) Reducing the lateral deformations
The Disadvantages (Tabeshpour, et al.2011) of converting the flexural action to axial
action:
1) Increase of the axial load in the column and foundation,
2) Creation of the concentrated shears at top and bottom of the column,
3) Creation concentrated shears at beginning and end of the beam
4) Creation of huge shears on the foundation
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Figure 1.8 Behaviour of infill wall subjected to lateral load. (photo from Klingner 1976)
Figure 1.9 Typical failure mode of URM infill wall and frame. (photo from Klingner 1976)
Figure 1.10 Failure of masonry walls during Turkey earthquake, 1999: (a) out-of-plane failure, (b) in-plane failure and (c) combined in- and out-of-plane failures (photo from Klingner 1976)
Figure 1.10 (a), (b), and (c) shows poor performance of infills during Turkey
earthquake. In Indian design practice the effect of infill in design is ignored, as a
result in Sikkim 14 Feb 2006 earthquake most RC buildings at Gangtok suffered
damages in some form or the other. The most common damage observed was cracks
in masonry infills, and separation between RC frame and infill. Not only private
society but also important structure like Government buildings including legislative
assembly building, Tashiling Secretariat, State Legislators’ Hostel, Geological Survey
of India (GSI) building at Deorali, suffered varying degree of damages. Among
Government buildings, GSI building was the worst affected. Fortunately, the ground
shaking was quite moderate and no RC building collapsed. And also in eastern and
southern Sikkim in private and government building infill and RC frame damage was
observed. In Sichey there was a government school a part of recently constructed
three-storey was found to be damaged. One masonry infill wall in the first storey
tilted out of plane (Figure 1.11 a) along with cracks in several other infills. (Kaushik,
et al.2006)
Figure 1.11. Government secondary school building at Sichey suffered moderate damages(Sikkim earthquake):(a) out of plane tilting of masonry infill wall, (b) inadequate shear reinforcement in columns, and spalling of cover concrete in columns at several locations due to corrosion
1.3 National and International Code provision for Infill Walls
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1.3.1 Indian Standard IS 1893 (Part-1): 2002
In IS1893:2002 Criteria for earthquake resistance design of structures Part 1 general
provisions and buildings gives only two formulae for frame without infilled (Eqn. 1.1)
and for frame with infilled (Eqn. 1.2). On modelling aspect of infill code is silent.
Ta=0.075 h0.75For RC frame without infill (1.1)
Ta=0 .09 h
√d For RC Frame with infill
(1.2)
1.3.2 FEMA-356 / ASCE-41
Federal Emergency and Management Authority or American Society for civil
engineering-41 provides guideline to generate infilled walls models in RC frame for
analysis. The elastic in plane stiffness of a solid unreinforced masonry infill panel
prior to cracking shall be presented with an equivalent diagonal compression strut of a
width ‘a’ given by equation below. The equivalent strut shall have the same thickness
and modulus of elasticity as the infill panel it represents (Figure 1.12)
a= 0.175 (λ1 hcol ) -0.04 rinf (1.3)
where: λ1=[ Eme t inf sin 2 θ
4 E fe I col hinf ]14 (1.4)
hcol= Column height between centerlines of
beams, in.
hinf= Height of infill panel, in.
Efe= Expected modulus of elasticity of frame
material, ksi
Eme= Expected modulus of elasticity of infill
material, ksi
Icol= Moment of inertia of column, in4.
Linf= Length of infill panel, in.
rinf= Diagonal length of infill panel, in.
tinf= Thickness of infill panel and equivalent strut, in.
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Figure 1.12 Showing compression strut analogy and parameters (Photo from FEMA356)
1.3.3 Eurocode 8
In Eurocode 8 which is for Design of structures for earthquake resistance gives some
clause about infills walls as follows:
4.3.6 Additional measures for masonry infilled frames
4.3.6.3 Irregularities due to masonry infills
4.3.6.4 Damage limitation of infills
5.9 Local effects due to masonry or concrete infills
6.10.3 Moment resisting frames with infills
1.4 Literature review on retrofitting techniques
In recent years there has been a substantial advancement in research on repair and
seismic retrofitting of existing building as evident from increasing number of the
growing number of research papers published in this area.
Repair: The process to regain original strength of a damage or deteriorated structure is
called as Repair.
Seismic Retrofitting: The process to enhance original strength of a deficient or
damaged structure and enabling it to satisfactorily can perform its intended
performance in future seismic event is called retrofitting.
1.4.1 Objective of retrofitting
Objective of seismic retrofitting is only to improve seismic performance of building.
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1.4.2 Retrofitting strategies
Retrofitting strategy is the basic overall approach to enhance the probable seismic
performance of the building or to otherwise reduce the existing risk to an acceptable
level [ATC-40]. Retrofitting strategies can be categorized as: (1) Completion of load
path and removal of structural irregularity (2) Strengthening of structure (3)
Enhancing deformation capacity of structure and (4) Reducing earthquake demand.
A structure, which is deficient in original design, can be retrofitted by strengthening
of the structure. This strengthening can be achieved either by adding new lateral load
resisting members or by strengthening the existing members. Large number of
techniques based on conventional strengthening method, such as addition of new
members (shear walls, bracings), RC jacketing, steel jacketing, as well as, based on
advance material such as FRP have been developed. Strengthening is the most
suitable and commonly used method of retrofitting for URM infilled RC frames.
Deformation capacity of URM infilled RC frame buildings can also be increased to
some extent by improving the ductility of beams, columns and infills. However, this
approach has limited scope due to brittle behaviour of URM infills. Supplemental
energy dissipation does not have much utility in URM infilled RC frame buildings
due to low inter-story drifts. Base isolation is a promising approach useful for low rise
buildings. In the present review, focus will be on various seismic strengthening
techniques and effectiveness of stiffness and ductility enhancement. There are four
strategies of retrofitting which are as follows:
1) Stiffness increase
2) Strength increase
3) Ductility increase
4) Mass reduction
But we are mainly focusing on stiffness and strength increase of structure due to
retrofitting. While remaining two strategies, ductility and mass reduction is not
primary attention.
1.4.3 Retrofitting techniques
Retrofitting techniques are the specific methods used to implement the overall retrofit
strategy. Under a given retrofitting strategy a number of retrofitting techniques are
available. Typical load-displacement relationships for different strengthening
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techniques have been shown in Figure1.13. The figure shows that for monolithic wall,
strength of structure is very high but ductility is very low and for unstrengthened
frame column ductility is very high but strength is less. So it can be understood from
this Figure that both ductility and strength cannot be achieved at the same time.
Figure 1.13 Typical load-displacement relationships for different strengthening
Techniques [Rodriguez et. al. (1991)]
1.4.3.1 Addition of shear wall
Addition of shear wall (Figure 1.14 a) into an existing building is most common
approach of seismic retrofitting. It has been used with frame, since long time. It is an
effective method of increasing building strength and stiffness. When shear walls are
situated at proper positions in a building, they can form an efficient lateral-force
resisting system, while simultaneously fulfilling functional requirements. Addition of
shear wall improves buildings strength and stiffness, and also it is economically
feasible and readily compatible with most of existing concrete buildings. It also gives
good aesthetic view. Many buildings all over the world have been retrofitted using
shear walls. In Japan, from a period of 1933 to 1975 about 85% case of retrofitting
was executed using shear walls [Rodriguez et al. (1991)]. Similarly, in countries like
United States, India, Turkey, and other places use of shear wall for seismic retrofitting
of existing buildings is well accepted [Pincheira (1993), Holmes (2000), Moehle
(2000)].
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Various analytical models have been proposed in literature to simulate the behaviour
of RC shear wall. These include Equivalent beam model, three vertical line element
model, inelastic analytical macroscopic model, Macro-finite-element model.
In equivalent beam model the shear wall member is replaced at its centroidal axis by a
line element and connected by rigid link to the frame beams. The main limitation of
this model lies in the assumption that rotation occurs around points belonging to the
centroidal axis of the wall so it does not accounts for migration of the neutral axis of
the wall cross-section, rocking of the wall etc. In three vertical line element model, a
generic wall member is idealized as three vertical line element with infinitely rigid
beam at the top and bottom floor levels. Two outside truss elements represent the
axial stiffness of the boundary columns, while the central element was one-component
model with vertical, horizontal and rotational springs concentrated at the base. This
model is capable of describing flexural and shears deformation including the
migration of the neutral axis of the wall cross-section, rocking of wall etc., but the
deformation due to the fixed end rotation and web splitting-crushing mode of failure
is not accounted [Vulcano and Bertero 1987]. Inelastic analytical macroscopic model
uses eight inelastic axial springs connected by two rigid beams to account plastic
bending deformation of wall, and three shear springs which expresses the shear
behaviour of panel and two boundary columns [Fu, et al., 1992]. Macro-Finite-
Element model consists of a number of vertical elements. These vertical elements
consist of vertical and horizontal springs at the centre of each vertical element. The
axial stiffness of each vertical spring is represented by two parallel components
representing mechanical behaviour of the concrete and steel. Horizontal spring
represents the shear springs. The stiffness of each shear spring is determined by the
different state of vertical spring. In this model the axial springs first reach the
nonlinear state and then the shear spring, thus the effect of axial stiffness on shear
stiffness was neglected.
As reported by Mike Griffith in his JRC Scientific and Technical Reports a recent
experimental study by Altin et al (1992) tested fourteen 2-storey by 2-bay concrete
frames that were strengthened with concrete infill walls cast-in place. The
effectiveness of various degrees of inter-connection between the infill wall
reinforcement and the surrounding concrete frame were assessed and all results were
compared to the hysteretic behaviour of the bare concrete frame. It was observed that
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while the peak strength (reached at about 0.4% drift) was not sensitive to the degree
of frame wall connectivity, the hysteretic behaviour was best for the most integrally
connected infill walls (maximum displacements corresponding to drifts of
approximately 1 to 1.5%). Finally, in the recent European Conference on Earthquake
Engineering, 3 papers were presented on this topic. The first of these presented the
results of analyses on the seismic response of concrete frame school buildings in
Taiwan retrofitted with concrete infill walls [Sheu, et al., 1998]. The paper by Pop et
al (1998) presented the experimental results of tests on bonded anchors for use
between concrete infill walls, which were added to pre-existing concrete frames. It
was determined that an embedment length of 8 bar diameters into the concrete frame
was required to achieve optimal force transfer and interaction. The third paper
[Ozcebe, et al., 1998] presented the results of an experimental investigation into the
effectiveness of cast-in-place concrete walls as a seismic retrofit strategy for concrete
frame buildings damaged in the 1995 Dinar, Turkey earthquake. The tests showed that
the once the walls had been added the frame had little apparent effect on the strength
of the rehabilitated structure. However, lap splices in the columns prevented the infill
from achieving full effect. Steel plate jackets were recommended for the column
splice zones to solve this problem.
There are some adverse effects of addition of shear wall for retrofitting also we should
aware of this. If large number of shear wall added then it result in increase in mass of
the building and therefore increase in seismic forces also demand i.e., requirement of
strength increases. Shear walls can effect into architectural impact through the loss of
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(b)Figure 1.14 Shear wall (a) shear wall in building (b) Proper anchoring of
vertical reinforcement into foundation (Photo from Murthy CVR EQ tips)
(a)
windows. It also requires special foundation (Figure 1.14 b) work which highly
expensive as it produces large overturning forces at their base (ATC-40).
1.4.3.2 Addition of bracing
Additions of bracing to RC frames are another common method of seismic retrofitting
which increase stiffness, strength and ductility. But it provide less stiffness and
strength compare to shear walls. It can construct with less disruption in building with
very small loss of lights and have smaller effect of traffic patterns within building.
While strengthening a frame using bracing, the main unknown is how the new bracing
system will interact and behave when attached to an existing structure. Generally, it
has been found that bracing is an efficient technique and can be used in combination
with other techniques such as interior shear wall or column jacketing. Experimental
studies on strengthening of existing frames using bracing shows a significant
improvement in strength, stiffness and even ductility [Baboux, et al., 1990, Masri et
al., 1996]. It’s very difficult to attach braces with frame in seismic retrofitting. In
some studies, the bracings are directly fitted within concrete frame, whereas in some
cases steel frame is used to attach the bracing within concrete frame. Currently, there
is a lack of rational provision for analysing, designing and detailing these systems.
For simplicity, many times the direct summation of the strength and stiffness of
concrete frame and steel frame are done; this will often underestimate the total
strength. The total strength of the retrofit system should be determined as the
composite strength of the steel/concrete system [Moehle, 2000]. Ranges of bracing
system have been proposed in the literature for upgrading existing concrete frame.
This include concentric bracing (diagonal and X-bracing), Eccentric bracing, post-
tensioned bracing and buckling restrained bracing.
The main steps in retrofitting with steel bracing are outlined in the flowchart [Badoux,
M. and Jirsa, J.O. 1990] of (Figure 1.15) Retrofitting decision is based on an
inadequate evaluation of structure (step 1), If the structure is found inadequate (step 2)
and a retrofitting option is chosen (step 3), the engineer must determine the aim of the
retrofitting operation (step 4)—the structural and Architectural requirements. The
selection of the best retrofitting scheme (step 5) depends on the existing structure. If a
bracing scheme is selected, the first step of the design is the choice of the layout of the
system (step 6).Next, the members and connections are detailed (step 7). The
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foundations of the braced frame may require strengthening because greater foundation
forces are typically generated in the retrofitted frames under lateral loads. Finally, the
designer must try to anticipate particular construction problems that are likely to come
up when working on an existing structure (step 9). Allowance should be made for
fitting to clearances and possible in situ modification.
Figure 1.15 Flowchart for Retrofitting with Bracing System
1.4.3.2.1 Concentric Bracing
This type of bracing (Figure 1.16 a) has been widely used in strengthening of existing
frames [Nateghi; 1995]. Concentric bracings essentially consist of one or more
diagonal members. Experimental tests on strengthening existing frames using
concentric bracing showed its adequacy for lateral load resistance. Test on one
diagonal brace shows 2.5 times increase in shear strength, whereas X-bracing showed
4 times increase [Maheri, 1997]. In the same experiment, behaviour of X-braced
frame under the loading showed that the tension braces carries a large portion of load
and the failure of bracing starts with failure of tension braces followed by buckling
failure of compression braces. Other tests also showed that, 60% to 70% of the total
load applied was carried by braces before buckling [Bush et al., 1991a, Bush et al.,
1991b] and in the same experiment it has been observed that The maximum lateral
load applied to the strengthened frame was 2.24 times the predicted design capacity
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and also The initial stiffness of the braced frame (precracked) was 1.5 times that of
the uncracked original frame.
1.4.3.2.2 Eccentric Bracing
An eccentric brace (Figure 1.16 b) differs from a conventional brace in a way that, the
centerlines of eccentric braces do not intersect at the center of the beam column joint,
but rather are offset horizontally from the joint. Eccentrically braced frames (EBFs)
are now gaining acceptance in seismic strengthening of RC frames [Ghobarah et al.,
2001, Perera et al., 2004]. EBFs have been recognized as efficient technique for
enhancing seismic resistance because in addition to strength and stiffness it also
provides ductility. One major advantage of eccentric bracing is protection from
buckling under the higher loads generated by a major earthquake. One more benefit of
using this type of bracing system is the considerable flexibility regarding the
placement of braces. In this type of bracing system, the axial forces induced in the
braces are transferred either to a column or to another brace through shear and
bending in a segment of the beam. The critical beam segment is called an “active
link” or simply link. Thus in this type of system yielding and inelastic energy
absorption occurs at the links created by brace. Stiffness of EBFs ranges between
stiffness of moment resisting frames to concentric braced frame.
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Figure 1.16 Bracing(a) Concentric bracing Lorant G.
<http://www.wbdg.org/resources/seismic_design.php>
Figure 1.16 Bracing(b) Eccentric bracing <http://web.iku.edu.tr/courses/insaat/ce007/>
Therefore these types of bracing system are a hybrid deriving its stiffness from truss
action and its ductility by inelastic deformation of the link [Popov, et al., 1987]. Based
on behaviour of links, EBFS are categorized as eccentric shear braces and eccentric
flexural brace. If ductility of the link beam relies primarily on shear deformation, it is
usually referred to as an eccentric shear brace (shear capacity of a short link is usually
less than its flexural capacity). If the deformation of the link beam is dominated by
flexure, it is described as an eccentric flexural brace. From experimental investigation
conducted by Malley (1984), it was found that, active link which yield primarily in
shear are more effective energy dissipater than which yield primarily in bending.
1.4.3.2.3 Post-tensioned steel bracing
The use of post-tensioned steel braces in seismic rehabilitation is relatively new
technique that can be applied efficiently to existing low and middle-rise frame
buildings. This type of bracing consists of high-slenderness steel strands which can
take only tension and are subjected to a prestressing force. Different levels of
prestress, based on building property, soil condition and desired building behaviour
can be applied to post-tensioned braces [Tena-Colunga, 1996] Teran-Glimore, et al.,
1996]. An initial prestress of 75% of their yield strength was applied during
strengthening of a low rise building in US [Pincheria, 1993] to determine its
adequacy. Such a level of initial brace prestress will cause braces to yield and allow
energy dissipation at relatively small lateral. Behaviours of the post-tensioned braces
was idealized using a bi-linear model that considered yielding in tension and elastic
buckling in.
1.4.3.2.4 Buckling restrained bracings
Even though conventional concentric bracing systems are efficient, but they suffer
buckling due to high slenderness ratio. The problem of buckling has lead to
development of buckling restrained braces (BRBs) or unbounded braces (UBs). BRBs
or UBs basically consists of three components, a steel core member, buckling
restraining part and the unbonding material as shown in Figure 1.17(a) & (b).
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(a) (b)
Figure 1.17 (a) Schematic of BRBs or UBs, (b) Typical types of BRBs [Tsai, et al., 2004]
In these types of braces, a core steel cross-shapes or flat bar member is encased into a
steel tube and confined by infill concrete. The steel core member is designed to resist
the axial force with full tension or compression yield capacity without the local or
global flexural buckling failure. When the brace is subjected to compression, the
unbonding material placed between the core member and the infill reduces the
friction. Figure 3(b) illustrates the typical cross sections of BRBs.
1.4.3.3 Jacketing
Jacketing adds both strength and stiffness to structure. There are various types of
jacketing generally observed. Such as column jacketing, beam jacketing, infill
jacketing, column-beam joint jacketing, etc.
1.4.3.3.1 Column jacketing
In 1970’s earthquake many of the structural failures due to inadequate shear strength
and/or improper spacing in confinement in concrete columns. So, to increase column
cross section column strengthening procedures i.e., jacketing of column (Figure 1.18)
is used. The main problem with this approach is that it often unacceptably increases
the dimension of the column, rendering the retrofit impractical. The use of thin carbon
fibre composite sheets avoids this problem and has consequently gained acceptance
over the past 10 years. Nevertheless, concrete jacketing of concrete columns has been
shown to be very effective in improving strength and ductility and converting strong-
beam weak-column buildings into buildings with a strong-column weak-beam
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mechanism (Choudhuri, et al., 1992; Rodriguez and Park, 1994; Bracci, et al., 1997;
Bush, et al., 1990).
1.4.3.3.2 Beam column joint jacketing
As we saw previously in earthquake damages survey one of main reason of RC frame
damages are beam column joint failure. And is due to inadequate joint reinforcement
and/or improper anchorage of longitudinal beam reinforcement or improper provision
of development length. As reported by Mike Griffith Ghobarah, et al., (1996a,b)
presented this type of retrofitting method of jacketing to RC beam-column joints with
corrugated steel sheeting. Column beam joint jacketing was shown to not suffer from
the outward swelling problems of steel jacketing with flat steel sheets as reported by
Priestley (1994a). The corrugations act to stiffen the jacket in the needed “hoop”, or
tie, direction and so maintain good confinement of the concrete. Four concrete test
specimens were tested, one which represented existing structures, one represented
current seismic design standards and two rehabilitated connections. Results indicated
good cyclic performance at high load levels and significant increases in shear capacity
(V kN kN u = 400 → 450 ) and displacement ductility ( = 2→ 4 Δ μ ) (As reported by
Griffith M.).
And also infill jacketing is also use for retrofitting. In which shotcreting,
prefabrication of reinforced concrete panels attached with dowels through masonry
walls, steel plates apply on masonry walls.
21
Figure 1.18 Jacketing of column (Photo from Famer group
<http://www.famergroup.com/earthquake.html>)
1.4.3.4 Friction dampers
Retrofitting the structure with friction dampers provide the strength control required
to obtain the optimum structural response. Adding friction dampers increase stiffness
of the frame until a certain shear level is reached, at which the dampers can be set to
slip. For earthquake loading a appropriate slip level can be selected to give optimum
response. The energy dissipated by the friction dampers reduces the energy demand
on the structure and damps out the structural response. (Pall A.S. and Pall R. 1991)
The Friction brake is widely used to extract kinetic energy from a moving body as it is
the most effective mean to dissipate energy. More than centuries, mechanical
engineers had successfully used this concept and also currently using this concept to
control motion of machinery and automobiles. This principle inspired to develop
friction dampers. Patented Pall friction dampers are available for tension cross
bracing, single diagonal bracing, cladding connections and friction base isolators.
Before shifted to site of building it is well tested to ensure proper slip load. It is very
inexpensive i.e., economical in cost and easy to handle. Its construction is also easy
and easy in install. It’s made up of specially treated series of steel plates to develop
friction as shown in Figure 1.19. They are clamped together with high strength steel
bolts and these plates are allowed to move on each other with a pre define load.
Figure 1.19 Pall Friction Damper (Photo from Golafshani. A. A and Gholizad.A 2009)
It is an effective way to control seismic response of structure and non-structural
damage. It does not impact the foundation design, features that make them
22
particularly attractive for upgrading existing buildings. They also seem to offer
savings in the initial cost of new structures and retrofitting of the existing ones
compared to conventional solutions. However, it is very difficult to maintain its
properties for long time intervals. It also has other disadvantages like they are
effective only for flexible structure, they encumber the design procedure and make it
more expensive and the selection of the appropriate slip load is a critical issue in the
performance of a structure with friction dampers.
These dampers have successfully gone through rigorous testing on shake table in
Canada and US. As reported (Filistrault 1986) in 1985 a 3 storey frame equipped with
friction-dampers was tested on a shake table. And it was found that frame using
friction dampers gives result much superior to that of moment- resisting frame and
moment resisting braced frame. An earthquake record with peak acceleration of 0.9g
did not cause any damage to friction- damped braced frame. While other two cause
permanent deformation. Also in 1987, a 9 storey three bay frame, equipped with
friction dampers, was tested on a shake table at Earthquake Engineering Research
Centre of the university of California at Berkeley. All members of the friction-
damped frame remained elastic for 0.84g acceleration which is maximum capacity of
shake table, while the moment- resisting frame would have yield at about 0.3g
acceleration (Aiken, et al., 1988).
1.5 Objective of the study1) To compare period & base shear of different frame like bare frame, infill
frame, and open ground storey frame by IS code & SAP model.
2) Performance of evaluation of these building by non-linear static procedure.
3) Comparison of performance enhancement of these building with different
retrofitting techniques.
4) Identification of most suitable retrofitting techniques.
1.6 Methodology
1) Generic plan of RC frame building selected.
2) Building will be model for different height i.e.,G+3, G+7, G+15
3) Selection of suitable modeling techniques in SAP
4) Modeling, design and comparison of base shear and period of vibration of
these building.
23
5) Performance of evaluation of these building by non-linear static procedure.
6) Selection of retrofitting techniques & corresponding modeling techniques in
SAP.
7) Base on result will identify most suitable retrofitting techniques.
1.7 Scope of the work
Work will be limited to one type of limited plan and three different heights.
24
Chapter 2
SELECTION OF GENERIC PLAN
2.1 Introduction
In this study, a generic plan of a typical college building has been selected. The plan
in one direction has large number of bays i.e., nine and in other direction it has three
bays. This type of plan has been selected to simulate different behaviour in
longitudinal and transverse direction. In longitudinal direction there are nine bays
which indicate sufficient redundancy. Where as in transverse direction there are only
three bays which represent comparatively lesser redundancy. In addition to this the
effect of infill walls will be prominent in longitudinal direction than in transverse
direction.
Figure 2.1 Showing plan and elevation of G+3 model
Figure 2.2 Showing Plan of G+3 bare frame model in SAP2000
25
Figure 2.3 Showing elevation of G+3 bare frame model in SAP2000
To calculate time period and acceleration by IS 1893:2002 take a generic plan of plan
dimension 41.4m X 14.7m shown in figure 2.1& 2.2 and in elevation (Figure 2.1 &
2.3 ) taking three building of G+3, G+7and G+ 15.
After design of G+3 building dimension of column, beam, slab, infill, and also load
on frame sections are
Section properties:
Column 1 (depth, width) =0.60 ×0.6 in meter
Column 2 (depth, width) =0.40 ×0.6 in meter
Beam 1 = 0.23x0.35 in meter
Beam 2 = 0.23x0.40 in meter
Beam 3 = 0.23x0.50 in meter
Slab (thickness) = 0.14 in meter
Outer infill wall (thickness) = 0.23 m
Inner infill wall (thickness) = 0.115 m
Load
Live load Roof floor =3.5 KN/m
Live load intermediate floors= 3.5 KN/m
2.2 Comparisons study
2.2.1 Time period Comparisons
In the present study, time period has been calculated as per IS1893:2002 using codal
provision method. It can be observed from Table 2.1 that while considering infill wall
26
effect time period reduces drastically as compared to the model without infill wall
effect.
Table 2.1 Comparison of time period with and without infilled wall model.
Time period (with infill wall)in Sec.
Time period (without infill wall) in Sec.
Storey X-Direction Y-Direction X and Y-Direction
G+3 IS-1893 0.233 0.373 0.597
G+7 IS-1893 0.440 0.711 0.968
G+15 IS-1893 0.865 1.387 1.598
2.2.2 Modal mass participation factors
It can be observed from Table 2.2 that the modal mass participation factors are well
distributed in bare frame and bare frame with infill load only. When model as bare
frame in SAP2000 mode shape are observed as 1st step in X direction 1st mode and 2nd
step in Y direction of a 1st mode. But when we model infill as a strut in SAP2000 then
x-direction first mode observed in step no.3 and Y-direction first mode in step 1 and
second step number is torsion mode it means the stiffness of model is changed due to
addition of infill wall. In Y direction stiffness is less compare to X direction as there
are only three bays in Y direction while 9 bays in X direction (cell filled with dark
colour).
Table 2.2 Comparison of Modal mass Participation factor by RSA in SAP 2000
Types of Building
Direction Mode No. 1 2 3 Steps no.
Bare frameX Modal Load
participation factor
0.83228 0.0919 0.02403 1'4'7'
Y 0.81587 0.10453 0.03005 2'5'9'
Bare frame with infill load only
X Modal Load participatio
n factor
0.84611 0.06364 0.02152 2'5'9'
Y 0.84029 0.09692 0.02432 1'4'7'
27
42
22
32
52
12
1
Full infilledX Modal Load
participation factor
0.8828 0.07065 0.01303 3'7'11'
Y 0.84107 0.09706 0.02433 1'4'6'
Open Ground storey
X Modal Load participatio
n factor
0.92287 0.02662 0.0042 3'6'10'
Y 0.84133 0.09694 0.02422 1'4'7'
2.2.3 Shear force and bending moment
Figure 2.4 Showing X-Z view of G+3 building in SAP 2000
In the present study, in Figure 2.4 different members of a building which have been
made bold are selected for calculating bending moment and shear force. Then these
moment and force with different models like bare frame, bare frame with infilled
load, full infilled and open ground storey building have been compared. From result it
has been concluded that if full infilled wall model take then both bending moment and
shear force increase (cell dark in table 2.3, 2.4, 2.5) compare to other frame model
i.e., bare frame and full infilled wall. And when we take member no.1 (table 2.3)
which is column of ground storey, shear force and bending moment are increase
compare to member 2 (table 2.4) which is column of first storey and increment is 2.14
times in shear force and 1.53 times in bending moment and also column no. 1 bending
moment and shear force increases compare to column element no. 4 (table 2.6). In
case of beam no. 3 compare to beam no. 5 bending moment and shear force increases
(table 2.5 & 2.7).
Table 2.3 Comparison of Shear force and Bending moment element No.1 shown in Figure 2.4 by RSA in SAP 2000
28
Type of Building
Location 1
CombinationsShear Force
in kNCombinations
Bending moment
kN m
Bare frame
Left side Corner ground storey
column of 1st frame
1.5(DL+EQx) 41 1.5(DL+EQx) 83
Bare frame with infill load
only1.5(DL+EQx) 77 1.5(DL+EQx) 196
Full infilled 1.5(DL+EQx) 56 1.5(DL+EQx) 112
Open Ground storey
1.5(DL+EQx) 176 1.5(DL+EQx) 321
Table 2.4 Comparison of Shear force and Bending moment element No.2 shown in Figure 2.4 by RSA in SAP 2000
Type of Building
Location 2
CombinationsShear Force
in kNCombinations
Bending moment
kN.m
Bare frame
Left side Corner 1st
storey column of 1st frame
1.5(DL+EQx) 30 1.5(DL+EQx) 61
Bare frame with infill load
only1.5(DL+EQx) 60 1.5(DL+EQx) 122
Full infilled 1.5(DL+EQx) 33 1.5(DL+EQx) 83
Open Ground storey
1.5(DL+EQx) 56 1.5(DL+EQx) 211
Table 2.5 Comparison of Shear force and Bending moment element No.3 shown in Figure 2.4 by RSA in SAP 2000
Type of BuildingLocation
3Combinations
Shear Force in kN
CombinationsBending moment
kN.m
Bare frame
Left side Corner 1st
storey slab Beam of 1st frame
1.5(DL+EQx) 59 1.5(DL+EQx) 110
Bare frame with infill load
only1.5(DL+EQx) 126 1.5(DL+EQx) 173
Full infilled 1.5(DL+EQx) 105 1.5(DL+EQx) 124
Open Ground storey
1.5(DL+EQX) 129 1.5(DL+EQX) 180
29
Table 2.6 Comparison of Shear force and Bending moment element No.4 shown in Figure 2.4 by RSA in SAP 2000
Type of Building
Location CombinationsShear Force
in kNCombinations
Bending moment
kN.m
Bare frame
Left side Corner 3rd
storey column of 1st frame
1.5(DL+EQx) 15 1.5(DL+EQx) 38
Bare frame with infill load
only1.5(DL+EQx) 19 1.5(DL+EQx) 64
Full infilled 1.5(DL+EQX) 20 1.5(DL+EQX) 60
Open Ground storey
1.5(DL+EQX) 20 1.5(DL+EQX) 57
Type of Building
Location CombinationsShear
Force in KN
CombinationsBending moment KN.M.
Bare frame
Left side Corner 3rd
storey slab beam of 1st
frame
1.2(DL+LL+EQx) 32 1.2(DL+LL+EQx) 41
Bare frame with infill load
only1.2(DL+LL+EQx) 39 1.5(DL+EQx) 55
Full infilled 1.5(DL+LL) 33 1.5(DL+LL) 30
Open Ground storey
1.2(DL+LL+EQx) 28 1.2(DL+LL+EQx) 27
Table 2.7 Comparison of Shear force and Bending moment element No.5 shown in Figure 2.4 by RSA in SAP 2000
Chapter 3
30
MODELLING AND ANALYSIS
3.1 Introduction
Performance based seismic engineering is the modern approach to earthquake
resistance design. Rather than being based on prescriptive mostly empirical code
formulation performance based design is an attempt to predict building with
predictable seismic performance. Therefore, performance objective such as life safety,
collapse prevention and immediate occupancy are used to define the state of building
following a design earthquake. This chapter provides what is nonlinear analysis and
pushover analysis with actual nonlinear analysis result and conclusion.
3.2 Non-linear static procedure
To model the complex behaviour of reinforced concrete structure analytically in its
non-linear zone is difficult. This has led engineers in the past to rely heavily on
empirical formulas which were derived from numerous experiments for the design of
reinforced concrete structures. For structural design and assessment of reinforced
concrete members, the non-linear analysis has become an important tool. The method
can be used to study the behaviour of reinforced concrete structures including force
redistribution. This analysis of the nonlinear response of RC structures to be carried
out in a routine fashion. It helps in the investigation of the behaviour of the structure
under different loading conditions, its load deflection behaviour and the cracks
pattern. Simplified nonlinear analysis procedures using pushover analysis methods,
such as capacity spectrum method requires determination of three primary elements:
capacity, demand (displacement) and performance.
Capacity: Capacity is representation of the structures ability to resist the seismic
demand. The overall capacity of a structure depends on the strength and deformation
capacities of the individual components of the structure. In order to determine
capacities beyond the elastic limits, some form of nonlinear analysis, such as
pushover procedure is required. This procedure uses a series of sequential elastic
analysis, superimposed to approximate a force-displacements capacity diagram of the
overall structure. The mathematical model of the structures is modified to account for
reduced resistance of yielding components. A lateral force distribution is again
31
applied until additional components yield. This process is continued until the structure
becomes unstable or until a predetermined limit is reached. The pushover capacity
curve as shown in Figure 3.1, approximates the structures behave after exceeding their
elastic limit.
Figure 3.1 Capacity curve
Demand (Displacement): Demand is representation of the earthquake ground motion.
During an earthquake Ground motion produce complex horizontal displacement
patterns in structures that may vary with time. Tracking this motion at every time
steps to determine structural design requirements is judged impractical. Traditional
linear analysis methods use lateral forces to represent a design condition. For
nonlinear methods it is easier and more direct to use a set of lateral displacement
demand is an estimate of the maximum expected response of the building during the
ground motion. Figure 3.2 shows elastic response spectrum also called demand curve.
Figure 3.2.Demand curve
Performance: Once a capacity curve and demand curve is defined, a performance
check can be done. In other words, the structure must have a capacity to resist
32
Roof displacement
Bas
e sh
ear
Capacity curve
Elastic response spectrum spectrum
earthquake demand such that the performance of the structure is compatible with the
objective of the design. A performance check verifies that structural and non-
structural components are not damaged beyond the acceptable limits of the
performance objective for the forces and displacements implied by the displacement
demand. Figure 3.3 shows intersection of demand curve and capacity curve which is
called as performance point.
Figure 3.3 Performance point
3.2.1 Pushover analysis
ATC 40 and FEMA 273, FEMA 356 and FEMA 440 have described the Push Over
analysis procedure, modelling of different components and acceptable limits. In
general, this method develops a damage curve (capacity curve) for a given direction.
To get the desired performance level from this curve, two methods, namely, Capacity
Spectrum Method and Displacement Coefficient Method have been introduced in
FEMA 440. This analysis procedure considers only first mode shape of the equivalent
single degree of freedom system and predefined vertical distribution of the load along
height in one direction at a time. These are the limitations of this method. Still it is
very efficient analysis procedure because it gives full insight of the nonlinear behavior
of the structure.
Pushover procedure is used to determine capacity curve. It determines capacity
beyond elastic limit. It is basically a step by step plastic analysis for which the lateral
loads of constant relative magnitude are applied to a given structure and progressively
increased until a target displacement is reached, while gravity loads are kept constant.
33
Performance point
Pushover analysis consists of a series of sequential elastic analysis, superimposed to
approximate a force- displacement curve of the overall structure. A two or three-
dimensional model which includes bilinear or trilinear load-deformation diagrams of
all lateral force resisting elements is first created and gravity loads are applied
initially. A predefined lateral load pattern, which is distributed along the building
height, is then applied. The lateral forces are increased until some members yield. The
structural model is modified to account for the reduced stiffness of yielded members
and lateral forces are again increased until additional members yield. The process is
continued until a control displacement at the top of building reaches a certain level of
deformation or structure becomes unstable. The roof displacement is plotted with base
shear to get the global capacity curve. Figure 3.4 shows flow chart of pushover
analysis by Capacity spectrum method (ATC-40)
Figure 3.4 Flow chart of capacity spectrum method (ATC-40)
Pushover analysis can be performed as force-controlled or displacement controlled. In
force-control pushover procedure, full load combination is applied as specified, i.e.,
34
force-controlled procedure should be used when the load is known (such as gravity
loading). Also, in force controlled pushover procedure some numerical problems that
affect the accuracy of results occur since target displacement may be associated with a
very small positive or even a negative lateral stiffness because of the development of
mechanisms and P-delta effects.
Generally, pushover analysis is performed as displacement-controlled proposed to
overcome these problems. In displacement controlled procedure, specified drifts are
sought (as in seismic loading) where the magnitude of applied load is not known in
advance. The magnitude of load combination is increased or decreased as necessary
until the control displacement reaches a specified value. Generally, roof displacement
at the centre of mass of structure is choosing as the control displacement.
The practical difficulties associated with the non-linear direct numerical integration of
the equations of motion leads to the use of non-linear static pushover analysis of
structures. Pushover analysis is getting popular due to its simplicity. High frequency
modes and nonlinear effects may play an important role in stiff and irregular
structures. The contribution of higher modes in pushover analysis is not fully
developed.
3.2.2 Modelling and analysis procedure used in the present study
In present study, analysis performed using SAP2000 nonlinear version 14.2.4. A three
dimensional model of the structure has been define as a frame element. Define plastic
hinges at both ends of beams and column. The nonlinear properties of plastic hinges
are to be given as input to SAP2000.
3.2.2.1 Nonlinear modelling of Masonary infilled
Large number of research has been carried out in the past on analytical modeling of
masonry infills, Based on these studies, it was observed that masonry infills can be
conveniently modeled as compressive diagonal struts. In the present study, the elastic
in-plane stiffness of a solid unreinforced masonry infill panel prior to cracking shall
be represented with an equivalent diagonal compression strut of width. The equivalent
strut shall have the same thickness and modulus of elasticity as the infill panel it
represents shown in figure 1.12. It was reported that specimens with strong frames
35
and strong panels exhibited both better lateral load resistance and energy dissipation
capacity performance than those with weak frames and weak panels. The width of
compressive struts ‘a’ was considered as given in FEMA 356. The thickness of struts
was taken as the actual thickness of walls (Figure 1.12).
Determination of equivalent strut
Figure 3.6 Showing sectional view in YZ plane and infill model as strut in SAP2000
Strut width is depend on various factors such as span of bay, height of infill, thickness
of infill, quality of masonary, Size of infill, adjacent column width and depth etc., In
figure 3.6 beam size change in 1st and 2nd bay from 3rd floor to 2nd floor that’s why
strut width change S-J, S-K, S-M, etc., but in 3rd bay, beam size is similar in all floor
that’s why strut width is same and denoted by ‘S-L’. Some infills struts widths are
shown in below table 3.1 by FEMA 356.
Table 3.1 Calculated equivalent strut width for different infill
Strut Name
Equivalent strut width in m.
S-J 0.569
S-K 0.505
S-L 0.938
S-M 0.567
S-N 0.502
36
The analytical modeling of infilled frames is a complex issue because these structures
exhibit highly nonlinear inelastic behaviour resulting from the interaction of the
masonry infill panel and the surrounding frame. The modeling approaches for
masonry can be grouped into micro models and macro models. Micro models capture
the behavior of infill and its interaction with the frames in much detail, but these
models are computationally expensive and time consuming. On the other hand, macro
models try to capture the gross behavior of the infill, are approximate but
computationally efficient.
In this study, masonary act as a strut and model as a lumped plasticity model. If the
lateral force is applied on building then the force is developed in strut which acts like
compression. So that’s why axial hinge is provided at centre of strut. To calculate
axial hinge we require some nonlinear properties which are shown in figure 3.7 & 3.8
like yield force, displacement control parameter C, D, E and also acceptance criteria.
Figure 3.8 Nonlinear static procedure-simplified Force- Deflection Relations for Masonary
infill panels (photo from FEMA356)
37
Figure 3.7-b Manual hinge provision in SAP2000
Figure 3.7-a Manual hinge provision in SAP2000
To calculate yield force there are some formulae, given in FEMA 356 which are
F y= Vinecos θ
(N) (3.1)
V ine=A∋x f vie(N ) (3.2)
Where, Fy is maximum allowable yield force,
Vine is design shear force,
Ani is area of strut,
Fvie is expected shear strength of masonry infill,
And to calculate values of IO, LS & CP (figure3.9) there are some formulae given in FEMA356 and also from table 7-9 (figure 3.8) A strut=t inf X ESW (mm) (3.3)
d y=% X h inf X cosθ (3.4)
Dy= Py
Astrut x EmeLinf
(3.5)
PD=(d y−D y )
1000
(3.6)
d ls=(% for LS) X hinf X cosθ
(3.7)
Where,Astrut = Area of strut,
Eme = Expected elastic modulus of masonry in Compression,
Linf = Length of infill,
hinf = Height of infill,
θ = Angle between infill diagonal and horizontal axis,
tinf = thickness of infill,
ESW=Equivalent strut width.
38
Figure 3.9 Generalized Force-Deformation relations for masonry Elements or Components (Photo from FEMA356)
Table 3.2 Nonlinear properties of infill hinges
Infill name
Yield Force
Displacement control parameter
Acceptance Criteria
Fy (kN) B C D, E LS CP
S-J 261.80 0 0.083 0.083 0.063 0.083
S-K 231.66 0 0.089 0.089 0.069 0.089
S-L 421.04 0 0.026 0.026 0.016 0.026
S-M 260.84 0 0.083 0.083 0.063 0.083
S-N 230.58 0 0.089 0.089 0.069 0.089
Figure 3.10 Nonlinear properties of axial hinge to be filled in SAP
Table 3.2 shows the calculated nonlinear properties of some infills and figure 3.10
shows how to fill these properties in SAP2000.
39
3.2.2.2 Pushover analysis of G+3, G+7, G+15 stories
Pushover procedure is use d to determine capacity curve. It determines capacity
beyond elastic limit. It is basically a step by step plastic analysis for which the lateral
loads of constant relative magnitude are applied to a given structure (figure 3.11) and
progressively increased until a target displacement is reached, while gravity loads are
kept constant.
Figure 3.11 Static approximation used in the pushover analysis (photo from Mortezaei A.)
Figure 3.12 Load pattern in pushover analysis (photo from A. Mortezaei)
Pushover analysis consists of a series of sequential elastic analysis, superimposed to
approximate a force- displacement curve of the overall structure. A two or three-
dimensional model which includes bilinear or trilinear load-deformation diagrams of
all lateral force resisting elements is first created and gravity loads are applied
initially. A predefined lateral load pattern, which is distributed along the building
height, is then applied (figure 3.12). The lateral forces are increased until some
members yield. The structural model is modified to account for the reduced stiffness
of yielded members and lateral forces are again increased until additional members
yield. The process is continued until a control displacement at the top of building
reaches a certain level of deformation or structure becomes unstable. Figure 3.13
40
shows yielding sequences when lateral load applied. The roof displacement is plotted
with base shear to get the global capacity curve.
In present work, Pushover analysis of 4, 8 & 16 story building with bare frame, full
infill and open ground story model have been studied.
Figure 3.13 Yielding sequences through conventional pushover analysis (photo from A. Mortezaei)
Figure 3.15 shows Pushover curve of g+3 stories in which 3 curves of bare frame, full
infilled and open ground story model. Bare frame model shows a large deformation
and ductility. Due to formation of mechanism open ground story model gives very
less displacement and there is not deformation further and analysis is stops before
formation of collapse hinges.
In Y direction, G+3 stories pushover curve shows (Figure 3.16) large displacement
and high ductility in bare frame. When same frame modelled as a full infilled, curve
shows decrease in ultimate displacement and less ductility. Strength carrying capacity
of this full infilled model is increase because of high rigidity & stiffness due to infill
but ultimate displacement is decrease. In OGS Strength increase as compare to bare
frame but ductility and deformation decreases.
Conclusion: From pushover curve in Y direction we conclude that due to infill,
structure strength increase but ductility decrease as compare to bare frame.
Figure 3.17 shows pushover curve of G+7 stories. In bare fame X direction, as usual
curve shows high displacement due to high stiffness and also shows large ultimate
displacement. In case of full infill same as G+3, strength capacity of building
increases but ultimate displacement decrease and in OGS frame ductility is reduce
drastically and structure fails early as compare to G+3 building.
41
Conclusion: As we go towards high stories OGS construction is dangerous. So there is
requirement of some retrofitting.
In Y direction of G+7 stories building (figure 3.18), pushover curve of bare frame
shows large displacement and full infilled and OGS also shows displacement. There is
not so much difference in strength carrying capacity also.
Conclusion: As we see plan of building in figure 3.18, there is only four frame in Y
direction as compare to 8 stories so there is not that much difference in base shear and
displacement as we observe in X direction.
Figure 3.19 shows pushover curve of G+15 stories. In bare frame X direction, as usual
curve shows high deformation due to high stiffness and also shows large ultimate
displacement. In case of full infill same as G+3 & G+7, strength capacity of building
increases but ultimate displacement decrease and in OGS frame ductility is reduce
drastically and structure fails early.
For Y direction in figure 3.20 same as in X direction
Figure 3.14 Plan and elevation of G+3 RC frame Building
42
0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.40
2000
4000
6000
8000
10000
12000
14000
full in-fill
OGS
Bare frame
Displacement in m
Ba
se s
hea
r in
kN
B- IO- LS- CP- DBE- MCE-
Figure 3.15 Pushover curve of G+3 building in X - direction
0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.40
2000
4000
6000
8000
10000
Bare frame
Infill
OGS
Displacement in m
Ba
se s
hea
r in
kN
B- IO- LS- CP- DBE- MCE-
Figure 3.16 Pushover curve of G+3 building in Y- direction
0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.4 0.45 0.50
3000
6000
9000
12000
15000
18000
21000
Bare frame
Full infill
OGS
Displacement in m
Base
sh
ear
kN
B- IO- LS- CP- DBE- MCE-
Figure 3.17 Pushover curve of G+7 building in X- direction
43
0 0.1 0.2 0.3 0.4 0.5 0.60
2000
4000
6000
8000
10000
12000
Bare frame
Full in-fill
OGS
Displacement in m
Base
sh
ear
in k
N
B- IO- LS- CP- DBE- MCE-
Figure 3.18 Pushover curve of G+7 building in Y- direction
0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.80
2000
4000
6000
8000
10000
12000
14000
Full infill
OGS
Bare frame
Displacemnt in m
Bas
e sh
ear
in k
N
B- IO- LS- CP- DBE- MCE-
Figure 3.19 Pushover curve of G+15 building in X- direction
0 0.1 0.2 0.3 0.4 0.5 0.60
1000
2000
3000
4000
5000
6000
7000
Full infill
OGS
Bare frame
Displacementi in m
Ba
se s
hea
r in
kN
B- IO- LS- CP- DBE- MCE-
Figure 3.20 Pushover curve of G+15 building in Y- direction
3.2.2.3 Bi-Linearization of curve
44
Figure 3.21 shows the bi-linearization method of curve, which is used in present study
to linearize pushover curve of all three type building. After linearization, calculate
ductility demand which is calculated by ratio of ultimate displacement (Δu) to the
yield displacement (Δy) and Fy and Fu which is yield force and ultimate force
respectively which are calculated by bi-linearization method and shown in table 3.3
for x- direction and in table 3.4 for y- direction.
Figure 3.21 Bi-linearization of curve (FEMA356)
3.1.2.4 Determination of ductility
Table 3.3 Ductility, yield force, ultimate force in bare frame, full infilled and open ground
story model of different building in X- direction.
G+3 StoryDuctility=
Δu/ΔyG+7 Story
Ductility
=Δu/ΔyG+15 Story
Ductility=
Δu/Δy
Bare frame
Fy = 4300kN
5.214Bare frame
Fy = 9800kN
4.273Bare frame
Fy=11997 kN
3.020
Δy = 70mm
Δy =110mm
Δy=245mm
Fu = 4486 kN
Fu=10000kN
Fu=11997 kN
Δu = 365mm
Δu =470mm
Δu =740mm
Full infilled
Fy = 6556 kN
1.000Full
infilled
Fy=16000kN
5.571Full
infilled
Fy=17100 kN
2.400
Δy = 20mm
Δy = 70mm
Δy =150mm
Fu = 6556 kN
Fu=18388kN
Fu=19118 kN
Δu = 20mm
Δu =390mm
Δu =360mm
45
Open ground story
Fy = 5400 kN
1.600Open
ground story
Fy=15200kN
2.000Open
ground story
Fy=14000kN
2.000
Δy = 30mm
Δy = 80mm
Δy =125mm
Fu = 5811 kN
Fu=16633kN
Fu=17346 kN
Δu = 48mm
Δu =160mm
Δu =250mm
Table 3.4 Ductility, yield force, ultimate force in bare frame, full infilled and open ground
story model of different building in Y- direction.
G+3 StoryDuctility=
Δu/ΔyStory G+7
Ductility=
Δu/ΔyStory G+15
Ductility= Δu/Δy
Bare frame
Fy =4100kN
6.182Bare frame
Fy=7950 kN
5.611Bare frame
Fy=2050kN
9.800
Δy =55mm
Δy =90mm
Δy=100mm
Fu=4927 kN
Fu=8708 kN
Fu=2182kN
Δu =340mm
Δu=505mm
Δu=980mm
Full infilled
Fy=6000 kN
3.000Full
infilled
Fy=8200 kN
6.800Full
infilled
Fy=6200 kN
4.211
Δy = 25mm
Δy = 50mm
Δy =95mm
Fu=7975 kN
Fu=10319k
N
Fu=7136kN
Δu =75mm
Δu=340mm
Δu=400mm
Open ground story
Fy=5800 kN
4.000Open
ground story
Fy=7900 kN
5.727Open
ground story
Fy=5400kN
2.167
Δy =30mm
Δy = 55mm
Δy = 90mm
Fu=7552 kN
Fu= 9963 kN
Fu=6200 kN
Δu =120mm
Δu =315mm
Δu =195mm
3.2.2.5 Determination of Performance point
For calculating performance point Displacement Modification Method (DMM) has
been used to obtain the performance point (FEMA 356). In the DMM, the target
displacement ɗt roof level can be as
46
ɗt=C0C1C2SaxT e2
4 X π 2 xg
Where, Co is modification factor to relate spectral displacement of an equivalent
SDOF system to the roof displacement of the building. C1 is a modification factor to
relate expected maximum inelastic displacement to displacement calculated for linear
elastic response. For period greater than 1.0s the value of C1 is taken as 1. C2 is
modification factor to represent the effect of pinched hysteresis shape, cyclic stiffness
degradation on maximum displacement response. For period greater than 0.7s the
factor C2 is 1. Sa is the spectral acceleration, at effective fundamental period Te and the
damping ration of the building in the direction under consideration; g is acceleration
due to gravity.
Figure 3.15 to 3.20 shows pushover curve with different performance level like B-
operational level in pink colour, IO- Immediate occupancy in blue colour, LS-Life
safety in faint blue colour, CP-Collapse prevention in green colour. Also shows
design base earthquake (DBE) by triangle sign in faint green colour and maximum
consider earthquake by square sign (MCE).
In figure 3.15, when DBE earthquake comes pink type hinge form in bare frame
means structure is in operational level and when MCE type earthquake comes then
structure is in immediate occupancy level in bare frame.
In figure 3.16, pushover in Y direction when DBE type earthquake come structure is
in operational level in bare frame but in full infill and OGS structure is in immediate
occupancy level and when MCE type earthquake comes full infill & OGS structure is
nearly collapse.
In figure 3.17, pushover in X direction of G+7 stories when DBE type earthquake
comes structure is in operational level in bare frame and in other type of model
structure is in IO level. When MCE type earthquake comes OGS structure is nearly in
collapse level and infill is in operational level. In Y Direction figure 3.18, when
consider MCE earthquake bare frame is in IO level, full infill & OGS is in life safety
level.
In figure 3.19, pushover curve in X direction of G+15 building when DBE type
earthquake comes structure is in operational level in bare frame and IO level in OGS
and full infill frame. When MCE type earthquake comes OGS is just near to collapse
level. In case of figure 3.20 in Y direction of a structure same.
47
3.2.2.6 Comparison of inter story drift ratio
Inter story drift ratio is very essential parameters for designing the non-structural
components of the building. In present work, It has been studied inter story drift ratio
in X & Y direction i.e., longer & short direction using pushover analysis. The
variation of inter story drift ratio at MCE level along the height of building is shown
in figure 3.28 to 3.33 and variation of inter story drift ratio at ultimate level is shown
in figure 3.22 to 3.27 . It has been observed that story drift demand is more in middle
stories in bare frame & full infilled frame model. The reason is sudden reduction of
cross section area of the frame elements along the height of the structures. When we
are considering infilled walls model drift demand has got substantially reduced. But in
case of open ground story model drift demand is more in ground story and as we go
above it decreases. Because there is no infill walls as result stiffness is reduces
drastically and it became soft story or weak story in both at MCE level and ultimate
displacement. For tall buildings, infills substantially affect the drift behaviour of the
buildings.
0.00 0.50 1.00 1.50 2.00 2.50 3.000
1
2
3
4
5
Bare frame
Full in-filled
OGS
Inter Story Drift Ratio (%)
Sto
ry le
vel
Figure 3.22 Inter Story Drift ratio in X-direction of 4 story at ultimate displacement.
48
0.00 0.50 1.00 1.50 2.00 2.50 3.00 3.500
1
2
3
4
5
Bare frame
Full in-filled
OGS
Inter Story Drift Ratio (%)
Sto
ry l
evel
Figure 3.23 Inter Story Drift ratio in Y-direction of 4 story at ultimate displacement.
0 0.5 1 1.5 2 2.5 30
1
2
3
4
5
6
7
8
9
OGS
Bare frame
Full infilled
Inter story Drift Ratio (%)
Sto
ry
lev
el
Figure 3.24 Inter Story Drift ratio in X-direction of 8 story at ultimate displacement.
0 0.5 1 1.5 2 2.50
1
2
3
4
5
6
7
8
9
OGS
Bare frame
Full in-filled
Inter story Drift Ratio (%)
Sto
ry le
vel
Figure 3.25 Inter Story Drift ratio in Y-direction of 8 story at ultimate displacement.
49
0.00 0.50 1.00 1.50 2.00 2.50 3.000
2
4
6
8
10
12
14
16
18
Bare frame
Full infill
OGS
Inter story Drift Ratio (%)
Sto
ry
lev
el
Figure 3.26 Inter Story Drift ratio in X-direction of 16 story at ultimate displacement.
0.00 0.50 1.00 1.50 2.00 2.50 3.00 3.500
2
4
6
8
10
12
14
16
18
Bare frame
Full infill
OGS
Inter story Drift Ratio (%)
Sto
ry
lev
el
Figure 3.27 Inter Story Drift ratio in Y-direction of 16 story at ultimate displacement.
0.00 0.20 0.40 0.60 0.80 1.00 1.20 1.400
1
2
3
4
5
Bare frame
Full in-filled
OGS
Inter story drift Ratio (%)
Sto
ry l
evel
Figure 3.28 Inter Story Drift ratio in X-direction of 4 story at MCE.
50
0.00 0.20 0.40 0.60 0.80 1.00 1.200
1
2
3
4
5
Bare frame
Full in-filled
OGS
Inter story Drift Ratio (%)
Sto
ry le
vel
Figure 3.29 Inter Story Drift ratio in Y-direction of 4 story at MCE.
0.00 0.10 0.20 0.30 0.40 0.50 0.60 0.700
1
2
3
4
5
6
7
8
9
OGS
Bare frame
Full in-filled
Inter story drift Ratio (%)
Stor
y le
vel
Figure 3.30 Inter Story Drift ratio in X-direction of 8 story at MCE.
0 0.2 0.4 0.6 0.8 1 1.20
1
2
3
4
5
6
7
8
9
OGS
Bare frame
Full in-filled
Inter story drift Ratio (%)
Stor
y le
vel
Figure 3.31 Inter Story Drift ratio in Y-direction of 8 story at MCE.
51
0.00 0.50 1.00 1.50 2.00 2.50 3.000
2
4
6
8
10
12
14
16
18
Bare frame
Full infill
OGS
Inter story drift Ratio (%)
Sto
ry l
evel
Figure 3.32 Inter Story Drift ratio in X-direction of 16 story at MCE.
0.00 0.50 1.00 1.50 2.00 2.50 3.000
2
4
6
8
10
12
14
16
18
Bare frame
Full in-fill
OGS
Inter story drift Ratio (%)
Stor
y le
vel
Figure 3.33 Inter Story Drift ratio in Y-direction of 16 story at MCE.
52
CHAPTER 4
SEISMIC RETROFITTING TECHNIQUES
4.1 Retrofitting techniques
As in previously study in non linear analysis the adverse effects of open ground story
like increase story drift ratio in ground story, Capacity curves gave small
displacement, hinges formations in ground story columns i.e., formation of
mechanisms and ground story columns fails before beams, etc. So these buldings are
requires special arrangements to increase the lateral stiffness & strength of the soft
story. In Indian code (IS 1893:2002 Part 1) there is provision for soft story
strengthening in clause 7.10.3 (a) which states that “The columns and beams of the
soft storey are to be designed for 2.5 times the storey shears and moments calculated
under seismic loads”(figure 4.1). While 7.10.3 (b) which states that “Placing shear
walls symmetrically in both directions of the building as far away from the centre of
the building as possible and it is to be designed for 1.5 times the lateral storey shear
force”.
53
Figure 4.1 OGS member designed for higher forces using code specified factors
Also other international code like Eurocode 8 CEN 2003 recommends increasing the
design forces for the soft first-story columns by 1.5 to 4.68 times & it depends upon
several factors. Israeli seismic code SII 1995 also recommends increasing design
forces of open first story and also for one adjacent story above by 2.1–3.0 times the
actual design forces, depending on the ductility level of the building. According to the
Bulgarian Seismic Code 1987, the seismic design forces for soft story in masonry-
infilled RC frames are required to be increased by 2 times the corresponding design
forces for a regularly infilled frame, and by three times the design seismic forces for a
regular bare frame. (Kaushik H.B.)
4.1.1 2.5 times increasing design forces of column & beam in soft story
The effectiveness of these strengthening schemes, in which the lateral strength of the
columns and the beams of the open first story are required to be increased by
increasing story shear and moment, by multiplying factor of 2.5. Then design for new
forces and again define plastic hinge properties in new section of soft story. After
nonlinear static analysis the new capacity curves shown in figure 4.2 to 4.7.
54
4.1.2 2.5 times increasing design forces in column only in soft story
In the case of the OGS frame strengthened using 2.5 times column & beams, it was
observed that plastic hinges developed in only the first-story columns. Stronger beams
would further increase the seismic demands on the columns. Therefore, increasing the
strength of first-story beams may exert additional force demands on the first-story
columns, whose strengths were also increased by using a predetermined multiplying
factor. This has been observed by Fardis and Panagiotakos 1997. As well, after
studying a similar clause in an older version of Eurocode 8 for buildings with severe
vertical irregularities. The new version of Eurocode 8 CEN 2003 requires that lateral
strength of only the first-story columns be increased. Also in Proposed Draft
Provisions and Commentary on Indian Seismic Code IS 1893 (Part 1) gives comment
on the clause for soft story that only column to be retrofitted by 2.5 times increasing
story shear and moment.
Figure 4.2 & 4.3 shows pushover curve i.e. capacity curve of G+3 building in X & Y
direction, when building is OGS then there is a formation of mechanism. It fails
before giving sufficient displacement or ductility. After retrofitting its stiffness,
strength increase as well as ductility. All form of hinges form and also energy
dissipated all over the frame model.
Figure 4.4 & 4.6 shows curve for G+7 & G+15 building in X-direction respectively.
After retrofitting it gives large deformation but it yields before as compare to OGS
model. Figure 4.5 & 4.7 shows curve in Y direction of G+7 & G+15 building. After
retrofitting strength, stiffness & roof displacement i.e., directly ductility increases.
0 0.02 0.04 0.06 0.08 0.1 0.12 0.14 0.16 0.180
2000
4000
6000
8000
10000
12000
14000
full infill
OGS
2.5 column
2.5 column & beam
Displacement (m)
Ba
se s
hea
r (
kN
)
B- IO- LS- CP- DBE- MCE-
55
Figure 4.2 Pushover curve of G+3 building retrofitting with 2.5 column & beam in X-
direction
0 0.05 0.1 0.15 0.2 0.25 0.30
2000
4000
6000
8000
10000
12000
Full infill
OGS
2.5 column only
2.5 column & beam
Displacement (m)
Ba
se s
hea
r (k
N)
B- IO- LS- CP- DBE- MCE-
Figure 4.3 Pushover curve of G+3 building retrofitting with 2.5 column & beam in Y-
direction
0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.4 0.450
3000
6000
9000
12000
15000
18000
21000
Full in-fill
OGS
2.5 RET of Col in X
2.5 col & beam in X
Displacement (m)
Bas
e sh
ear
(kN
)
B- IO- LS- CP- DBE- MCE-
Figure 4.4 Pushover curve of G+7 building retrofitting with 2.5 column & beam in X-
direction
56
0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.40
2000
4000
6000
8000
10000
12000
14000
Full infill
OGS
2.5 Ret col
2.5 Ret col & beam
Displacement (m)
Bas
e sh
ear
( kN
)
B- IO- LS- CP- DBE- MCE-
Figure 4.5 Pushover curve of G+7 building retrofitting with 2.5 column & beam in Y-
direction
0 0.1 0.2 0.3 0.4 0.5 0.6 0.70
2000
4000
6000
8000
10000
12000
14000
16000
18000
20000
Full infill
OGS
2.5 Ret of col only
2.5 Ret of col & beam
Displacemnt (m)
Ba
se s
hea
r (k
N)
B- IO- LS- CP- DBE- MCE-
Figure 4.6 Pushover curve of G+15 building retrofitting with 2.5 column & beam in X-
direction
57
0 0.1 0.2 0.3 0.4 0.5 0.60
1000
2000
3000
4000
5000
6000
7000
8000
9000
Full infill
OGS
2.5 Ret column
2.5 Ret column & beam
Displacementi (m)
Ba
se s
hea
r (k
N)
B- IO- LS- CP- DBE- MCE-
Figure 4.7 Pushover curve of G+15 building retrofitting with 2.5 column & beam in Y-
direction
4.1.3 Friction Damper
Retrofitting the structure with friction dampers provide the strength control required
to obtain the optimum structural response. Adding friction dampers increase stiffness
of the frame until a certain shear level is reached, at which the dampers can be set to
slip. For earthquake loading an appropriate slip level can be selected to give optimum
response. The energy dissipated by the friction dampers reduces the energy demand
on the structure and damps out the structural response. (Pall A.S. and Pall R. 1991)
Friction dampers are those that dissipate energy through the sliding of surfaces with
high coefficient of friction. Friction dampers are designed to have moving parts that
will slide over each other during a strong earthquake. When the parts slide over each
other, they create friction which uses some of the energy from the earthquake that
goes into the building. The earliest suggestion for the use of friction dampers is made
by W.O. Keightley in 1977. Keightley’s friction dampers are formed with two steel
plates slotted and clamped together with bolds and Belleville washers and lubricated
to prevent locking. The required normal force is provided by the tension in the bolts.
The Belleville washers are used to prevent loss of this tension. Energy is dissipated
after the tension or compression force applied to the plates along longitudinal
direction exceeds the friction force between the two plates and the plates slide one
58
with respect to the other. Application of cyclic load with a magnitude greater than the
slip force leads to rectangular force-deformation hysteresis loop.
Figure 4.8 Friction Damper in single diagonal (Pall A.S. and Pall R. 1996)
It is an effective way to control seismic response of structure and non-structural
damage. It does not impact the foundation design, features that make them
particularly attractive for upgrading existing buildings. They also seem to offer
savings in the initial cost of new structures and retrofitting of the existing ones
compared to conventional solutions. However, it is very difficult to maintain its
properties for long time intervals. It also has other disadvantages like they are
effective only for flexible structure, they encumber the design procedure and make it
more expensive and the selection of the appropriate slip load is a critical issue in the
performance of a structure with friction dampers.
In present work, friction dampers are defining as a link support element in which
Plastic (Wen) type damper are using. It is based on plastic wen theory which gives
elasto-plastic behaviour. It requires weight & some non-linear properties like stiffness,
yield strength, post yield stiffness ratio & yielding exponent.
Open ground story is specially constructed for parking. If we have to use friction
dampers in ground story then it disturbs vehicles to park. So by using trial & error
method best position of dampers such that it will not disturb parking such position&
also such yield strength were calculated. Only one direction i.e., in X direction & in
two internal frame of building friction dampers were used as it gives ease in parking
59
shown in figure 4.9. Figure 4.10 to 4.12 shows the capacity curve in X direction only
of G+3, G+7 & G+15 building using friction dampers.
After using friction dampers, capacity curve gives large displacement as compare to
open ground story means ductility increase and also increase in stiffness & strength
shown in figure 4.10 to 4.12. Table 4.1 shows comparisons of ductility in OGS &
retrofitting with friction dampers model. In G+3 model it increase from 1.6 to 4.24 &
in G+7 it increase from 2 to 5.86 & in G+15 it increase from 2 to 3.268.
Figure 4.9 G+3 building retrofitting with friction dampers showing Elevation & 3D view
0 0.02 0.04 0.06 0.08 0.1 0.12 0.14 0.160
2000
4000
6000
8000
10000
12000
14000
full infill
OGS
Ret with Fric damp.
Displacement (m)
Ba
se s
hea
r (
kN
)
B- IO- LS- CP- DBE- MCE-
Figure 4.10 Pushover curve of G+3 building retrofitting with Friction dampers in X-direction
60
0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.4 0.45 0.50
3000
6000
9000
12000
15000
18000
21000
Full in-fill
OGS
Ret friction
Displacement (m)
Ba
se s
hea
r (k
N)
B- IO- LS- CP- DBE- MCE-
Figure 4.11 Pushover curve of G+7 building retrofitting with Friction dampers in X-direction
0 0.1 0.2 0.3 0.4 0.5 0.60
3000
6000
9000
12000
15000
18000
21000
Full infill
OGS
Ret Frict dampers
Displacemnt (m)
Ba
se s
hea
r (
kN
)
B- IO- LS- CP- DBE- MCE-
Figure 4.12Pushover curve of G+15 building retrofitting with Friction dampers in X-
direction
Table 4.1 Ductility, yield force, ultimate force comparison in open ground story & retrofitting
with friction dampers model of different building in X- direction.
Type of building G+3 Ductility=
Δu/Δy G+7 Ductility= Δu/Δy G+15 Ductility=
Δu/Δy
Retrofitting with friction dampers
Fy = 9000kN
4.242
Fy = 15800kN
5.867
Fy=17100kN
3.268
Δy = 33mm
Δy =75mm
Δy=153mm
Fu = 10000kN
Fu = 18000kN
Fu =18800kN
Δu=140mm
Δu =440mm
Δu=500mm
Open Fy = 5400 1.600 Fy = 16633 2.000 Fy = 5400 2.000
61
ground story
kN kN kNΔy =30
mmΔy =80
mmΔy=125
mmFu = 5811
kNFu = 15200
kNFu = 5811
kNΔu =48
mmΔu=160
mmΔu=250
mm
4.1.4 Shear wall
Shear wall systems are one of the most commonly used lateral load resisting systems
in high rise buildings. Shear walls have very high in plane stiffness and strength,
which can be used to simultaneously resist large horizontal loads and support gravity
loads, making them quite advantageous in many structural engineering applications.
As a result of the large stiffness of walls, good story drift control can be achieved
during an Earthquake, as shown in Figure 4.13.
There are lots of literatures available to design and analyse of a shear wall. The
practice before 1960s has been to design buildings primarily for gravity loads and
check the adequacy of the structure for safety against lateral loads. It is established
that the design of a multi-storey building is governed by lateral loads and it should be
the prime concern of designer to provide adequate safety to structure against lateral
loads.
Figure 4.13 (a) Lateral load pattern; Lateral deformation pattern for (b) Frame element (c)
Wall element (d) Frame-wall building (Photo from Brama R.S. &Dasgupta K.)
Further, the old buildings were having substantial non-structural masonry walls,
partitions and connected staircases, which provided a significant safety margin against
lateral loads. The modern buildings are having light curtain walls, lightweight flexible
62
partitions along with high strength concrete and steel reinforcement, which reduce the
safety margin provided by non-structural components. Many existing RC frame
buildings located in seismic zones are deficient to withstand moderate to severe
earthquakes. Insufficient lateral resistances and poor detailing of reinforcement are the
main reasons for inadequate seismic performance of these structures. One of the most
popular methods of strengthening of seismically deficient structure is to provide shear
walls. “Shear walls” are defined as vertically oriented wide beams that carry
earthquake loads to the foundation. These are slender vertical cantilever RC walls
resisting the lateral load with or without frames. RC walls are often introduced into
multi-storey buildings at certain locations to resist lateral forces when frame systems
alone are insufficient. The term “shear wall” covers elevator shafts, stairwells and
central core units, in addition to plane walls. Shear walls acting with frames increase
the rigidity for lateral load resistance. When walls are situated in advantageous
position in a building, they can be very efficient in resisting lateral loads originating
from wind or earthquake.
Analysis for lateral loads of buildings containing shear walls is generally carried out
by assigning all lateral loads to the shear walls, since it is felt that the very big
difference in stiffness between the shear walls and the frame would cause the shear
walls to attract the total lateral loads. Shear walls in high seismic regions require
special detailing. However, in past earthquakes, even buildings with sufficient amount
of walls that were not specially detailed for seismic performance (but had enough
well-distributed reinforcement) were saved from collapse. Shear walls are easy to
construct because reinforcement detailing of walls is relatively straight-forward and
therefore easily implemented at site. The general method of providing a shear wall is
to fill the gaps between columns of the moment resisting frame with partial or
complete reinforced concrete wall is termed as Internal Shear Wall (Buttress Wall). A
new concept of providing walls on the outside is now being adopted in practice
(Yasaret al. 2008) known as External Shear Walls, Placing of reinforced concrete wall
along the external face of column with or without coupling beams is termed as
External Shear Wall. These types of shear walls are said to be advantageous
considering the fact that the building usage is not disturbed during retrofit. A study
has been carried out by M. Ashraf on optimum location of shear wall in a multi-storey
building. Few investigations have been carried out to evaluate the efficiency of
63
external shear walls in comparison to internal shear walls and to find the advantages
imparted by external shear walls to bare frame when subjected to lateral (Earthquake)
load. Present paper focuses on the influence of shear wall compared with that of
without shear wall on the resulting forces of the Moment Resisting Bare Frame under
consideration with different orientation and location of shear wall when subjected to
same type of load. A total of 5 models subjected to lateral load that may arise due to
earthquake on a fifteen storied building in zone II of the Seismic zones of India were
considered for the study. The top storey sway, support reaction, Column forces,
bending moment between the structural systems were compared with each other.
4.1.4.1 Addition of shear wall
Addition of shear wall into an existing building is most common approach of seismic
retrofitting. It has been used with frame, since long time. It is an effective method of
increasing building strength and stiffness. When shear walls are situated at proper
positions in a building, they can form an efficient lateral-force resisting system, while
simultaneously fulfilling functional requirements. Addition of shear wall improves
buildings strength and stiffness, and also it is economically feasible and readily
compatible with most of existing concrete buildings. It also gives good aesthetic view.
4.1.4.2 Modelling of shear wall
The analytical model for a solid wall element should represent the strength, stiffness
and deformation capacity of the wall for inplane loading. Out of plane behaviour need
not be consider, except where the wall acts as a flange for an intersecting wall
element. Solid walls may be considered “slender” if their aspect ratio (height/length)
is equal to or exceed 4 (hw/tw>= 4). Solid wall may be considered “squat” if their
aspect ratio is less than or equal to 2 (hw/tw =< 2). Slender walls usually are controlled
by flexural behavior, although shear strength may be a limiting factor in some cases.
Squat walls usually are controlled by shear behavior, although flexure sometimes may
be a limiting factor. The response of walls with immediate aspect ratios usually is
influenced by both flexure and shear. Potential failure of anchorages and splices may
require modelling of these aspects as well. Interaction with other elements should be
64
represented. Except for squat one and two story walls, sliding along construction
joints need not be modelled (ATC-40).
Various analytical models have been proposed in literature to simulate the behaviour
of RC shear wall. These include Equivalent beam model, wide column model, thin
shell model, three vertical line element model, inelastic analytical macroscopic model,
Macro-finite-element model.
In equivalent beam model the shear wall member is replaced at its centroidal axis by a
line element and connected by rigid link to the frame beams. The main limitation of
this model lies in the assumption that rotation occurs around points belonging to the
centroidal axis of the wall so it does not accounts for migration of the neutral axis of
the wall cross-section, rocking of the wall etc. In three vertical line element model, a
generic wall member is idealized as three vertical line element with infinitely rigid
beam at the top and bottom floor levels. Two outside truss elements represent the axial
stiffness of the boundary columns, while the central element was one-component
model with vertical, horizontal and rotational springs concentrated at the base. This
model is capable of describing flexural and shears deformation including the
migration of the neutral axis of the wall cross-section, rocking of wall etc., but the
deformation due to the fixed end rotation and web splitting-crushing mode of failure
is not accounted [Vulcano and Bertero 1987]. Inelastic analytical macroscopic model
uses eight inelastic axial springs connected by two rigid beams to account plastic
bending deformation of wall, and three shear springs which expresses the shear
behaviour of panel and two boundary columns [Fu,et al., 1992]. Macro-Finite-
Element model consists of a number of vertical elements. These vertical elements
consist of vertical and horizontal springs at the center of each vertical element. The
axial stiffness of each vertical spring is represented by two parallel components
representing mechanical behaviour of the concrete and steel. Horizontal spring
represents the shear springs. The stiffness of each shear spring is determined by the
different state of vertical spring. In this model the axial springs first reach the
nonlinear state and then the shear spring, thus the effect of axial stiffness on shear
stiffness was neglected.
In present work, shear wall has been model using equivalent wide column modelling
method. The shear wall member is replaced at its centroidal axis by a line element and
connected by rigid link to the frame beams. Line element is defining as column whose
65
width and length is shear wall’s width and length. Column width and length has been
taken as 0.23 m and 4.6 m respectively. Height to length (hw/tw) ratio is 0.78 means
shear wall is squat. Squat walls usually are controlled by shear behaviour. The wide-
column analogy was originally developed for planar wall structures such as walls with
openings (e.g.Clough et al., 1964) and was later extended to non-planar structures
(e.g. MacLeod and Hosny, 1977). Despite being widely applied for seismic analysis of
structures, only very little literature on wide-column models within elastic properties
has been found. Most of the research carried out in the past concentrated on the
behavior of wide-column models with elastic properties and on eliminating any
disadvantages related to the discretization of the walls into beam elements.
4.1.4.3 Analysis and design
In present work, shear wall has been model as wide column modelling as in non-
linear static analysis we requires non-linear properties and also define hinges of walls.
So it’s easy to give auto hinges to column as P-M2-M3. To check whether wide
column modelling is correct or not also thin shell model of shear wall has been
modelled. After design in both cases it gives approximately same percentage of steel.
Figure 4.15 and 4.16 shows percentage of steel in wide column modelling and thin
shell modelling in G+3 building respectively.
66
For optimization of best position of shear wall trial and error method has been used
and calculate best position. Then Shear wall is assigned in alternate position in x
direction i.e., longitudinal direction as shown in figure 4.17. Position is such that it
gives large displacement in capacity curve.
Figure 4.15. Thin shell model showing percentage of steel after addition of shear wall.
67
Figure 4.16 G+3 building retrofitting with shear wall showing 3D view of position of shear
wall
Figure 4.17 Pushover curve of G+3 building retrofitting with shear wall in X-direction
68
0 0.05 0.1 0.15 0.20
2000
4000
6000
8000
10000
12000
14000
full infill
OGS
Shear wall
Displacement in m
Ba
se s
hea
r in
kN
B- IO- LS CP - DBE- MCE-
Figure 4.18 Pushover curve of G+3 building retrofitting with shear wall in Y-direction
Figure 4.19 Pushover curve of G+7 building retrofitting with shear wall in X-direction
Figure 4.20 Pushover curve of G+7 building retrofitting with shear wall in Y-direction
Figure 4.21 Pushover curve of G+15 building retrofitting with shear wall in X-direction
69
0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.40
3000
6000
9000
12000
15000
18000
21000
Full infill
OGS
Ret Shear wall
Displacement in m
Ba
se s
hea
r k
N
0 0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.80
2000
4000
6000
8000
10000
12000
14000
16000
18000
20000
Full infill
OGS
Ret. Shear wall in-X
Displacemnt in m
Bas
e sh
ear
in k
N
0 0.05 0.1 0.15 0.20
1000
2000
3000
4000
5000
6000
7000
8000
9000
10000
Full infill
OGS
Shear wall in y
Displacement in m
Ba
se s
hea
r in
kN
0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.40
2000
4000
6000
8000
10000
12000
Full infill
OGS
Ret. Shear wall in Y
Displacement in m
Ba
se s
hea
r i
n k
N
Figure 4.22 Pushover curve of G+15 building retrofitting with shear wall in Y-direction
Figure 4.17 shows pushover curve of G+3 building in x direction. In which
comparison of capacity curve of full infilled, open ground story and retrofitted model
of shear wall. When retrofitting has been done it shows large deformation (figure
4.17). Initial stiffness also increased. Figure 4.18 shows pushover curve in Y direction
of G+3 building as there is no requirement of retrofitting in Y direction but due to
addition of walls in X direction it also affects in Y direction. Its initial stiffness
changes, load carrying capacity changes. But its analysis stops before due to
formation of mechanism and ductility reduced.
Figure 4.19 also shows pushover curve of G+7 story model in X direction with 3
different model but in this case ultimate load carrying capacity of retrofitted model is
decreases as compare to open ground story but ultimate displacement is drastically
increase. And figure 4.20 shows curves in Y direction same condition as described for
figure 4.18.
70
0 0.1 0.2 0.3 0.40
1000
2000
3000
4000
5000
6000
7000
Full infill
OGS
Ret. Shear wall in-Y
Displacementi in m
Bas
e sh
ear
in k
N
Figure 4.21 shows pushover curve of G+15 building in X direction with 3 different
models. In this case load carrying capacity, ultimate displacement, and initial stiffness
has been increased. Also in Y direction in figure 4.22 all three properties i.e., initial
stiffness, load carrying capacity, and ultimate displacement increase compare to open
ground story model.
Figure 4.24 G+3 building retrofitted with shear wall in X-direction showing hinges
formation.
Figure 4.23 shows different hinges formation in G+3 building retrofitted with shear
wall. There is no deformation in ground story compare to above stories. And analysis
stops due to formation of mechanism. In first story, in all columns hinges formed in
lower level.
After using shear wall, capacity curve gives large displacement as compare to open
ground story means ductility increase and also increase in stiffness & strength shown
in figure 4.17. Table 4.2 shows comparisons of ductility in OGS & retrofitting with
shear wall model. In G+3 model it increase from 1.6 to 7.118& in G+7 it increase
from 2 to 7.174& in G+15 it increase from 2 to 7.07.
Table 4.2 Comparison of ductility, yield force & ultimate force in open ground story &
retrofitting with shear wall model of different building in X- direction.
Type of building G+3 Ductility=
Δy/Δu G+7 Ductility= Δy/Δu G+15 Ductility=
Δy/Δu
Retrofitting Fy = 11105 7.118 Fy = 14047 7.174 Fy=17389 7.071
71
with Shear wall
Δy =17mm Δy = 46mm Δy=113mm
Fu = 12647 Fu = 15403 Fu =18964
Δu=121mm Δu =330mm Δu=799mm
Open ground story
Fy = 5400
1.600
Fy = 16633
2.000
Fy = 5400
2.000Δy =30mm Δy = 80mm Δy=125mm
Fu = 5811 Fu = 15200 Fu = 5811
Δu = 48mm Δu =160mm Δu=250mm
Chapter 5
CONCLUSIONS AND SCOPE FOR FUTURE WORK
5.1 Conclusions drawn from work
In India, most of the existing as well as new infilled RC frame buildings has been
designed and are being designed without considering strength and stiffness of Infills
(bare frame modeling). Due to inclusion of infills, behavior and failure modes of
buildings changes. This leads to serious concern about seismic safety of existing
buildings. In the present study, various strengthening techniques for open ground story
72
have been discussed. These techniques can be broadly categorized in two groups; 1
Strengthening of existing members, 2 Addition of new members.
1) In the first group, there are two methods; reinforced concrete jacketing and
steel jacketing. In the present work, jacketing is done by
a) Redesign beam and column of open ground story by increasing 2.5 times
seismic design forces. In present work, due to jacketing it gives good
ductility and increase up to 3 times compare to open ground story and also
increase in strength carrying capacity and initial stiffness of open ground
story.
b) Redesign only column of open ground story by increasing 2.5 times
seismic design forces. Also in this method it gives good ductility it
increase up to 3.5 times compare to open ground story and also increase in
strength carrying capacity and initial stiffness of open ground story.
2) In the second group, there are two popular methods; addition of shear wall and
addition of friction dampers.
a) Addition of friction damper is attractive and easy to construct but needs
sophisticated method for proper fixation with existing frames. In present
work, due to retrofitting with friction damper, there are change in ductility
it increase up to 2.5 times compare to open ground story and also increase
in strength carrying capacity and initial stiffness of open ground story. For
Indian RC framed buildings, it is observed that beam column joints are
most vulnerable to seismic failure, sometimes; addition of dampers
without proper strengthening of joints may lead to catastrophic failure of
building.
b) Addition of new shear wall can efficiently be used for buildings with only
local interventions. In present work, due to addition of shear wall
ductility which is increase up to 3.5 times compare to open ground story
and also increase in strength carrying capacity and initial stiffness of open
ground story. On the other hand, addition of shear wall needs laying of
new foundation, which in itself a difficult task.
In many cases, it could be difficult to achieve a single retrofitting technique for
attaining the desired performance of buildings. Combination of some of the above
mentioned techniques may be required. However, to determine the performance of
73
these techniques both experimental as well as analytical verifications in conjunction
with Indian construction practice are needed.
The RC frame with open-ground-storey exhibited very poor lateral strength stiffness
and energy dissipation capacity due to formation of shear hinges in ground-storey
columns under lateral load resulting uncontrolled excessive deformation in the
ground-storey in a nonlinear static analysis of a typical building.
5.2 Future work
1. Single strut model for infills can accurately predict the lateral stiffness and
strength of masonary infilled RC frame. However, use of single strut can only
take into account its compressive failure; it can’t predict local failure in frame
member. Single strut models underestimate the force resultants in frame
member.
2. In the present study, openings were not considered in infills. Presence of
opening in infills significantly reduces the stiffness and strength of the infilled
frames. Suitability of the proposed strengthening schemes must be verified for
masonary-infilled frames with openings with walls.
3. Also for future work, non linear dynamic analysis (time history analysis) is a
best method for analyzing the strengthening methods like friction dampers.
4. The experimental work should be carried out on a reduced scale three story
with first story without infilled wall under gradually increased cyclic lateral
displacements to further verify the effectiveness of proposed strengthening
schemes.
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