post-earthquake fire and seismic performance of welded steel–concrete composite beam-to-column...

Journal of Constructional Steel Research 67 (2011) 1358–1375

Contents lists available at ScienceDirect

Journal of Constructional Steel Research

journal homepage: www.elsevier.com/locate/jcsr

Post-earthquake fire and seismic performance of welded steel–concretecomposite beam-to-column joints

R. Pucinotti a,∗, O.S. Bursi b, J.F. Demonceau c

a Department of Mechanics and Materials, Mediterranean University of Reggio Calabria, Italyb Department of Structural Engineering, Trento University, Trento, Italyc Department ArGEnCo, University of Liège, Belgium

a r t i c l e i n f o

Article history:Received 24 October 2010Accepted 6 March 2011

Keywords:Seismic designFire designPost-earthquake fireSteel–concrete composite jointNumerical simulationsExperimental data

a b s t r a c t

The performance of steel–concrete composite full strength joints endowed with concrete filled tubes,designed with a multi-objective methodology dealing with seismic actions followed by fire is presentedin this paper. In detail, instead of a traditional single-objective design where fire safety and seismic safetyare independently achieved and the sequence of seismic and fire loading are not taken into account, theproposed design approach guarantees: (i) both seismic safety and fire safety with regard to accidentalactions; (ii) fire safety for at least 15 min fire exposure on a joint characterised by stiffness deteriorationand strength degradation due to seismic loading.

In order to achieve themulti-objective design, full strength beam-to-composite tubular column jointswere designed bymeans of the component method of Eurocode 4 Part 1-1 and Eurocode 3 Part 1-8, whileEurocode 4 Part 1-2 was considered for fire design. Moreover, to face a seismic-induced fire, they wereenhanced with specific joint components which will be detailed.

Both the experimental programme and the results provided by seismic tests, pre-damaged tests andfire tests carried out on beam-to-column joints are presented and discussed. The results demonstratetheir adequacy in terms of design and performance. Moreover, non-linear numerical simulations clearlyshow that these joints can be deemed adequate for moment resisting frames of medium ductility classcharacterised by a behaviour factor of about 4.

© 2011 Elsevier Ltd. All rights reserved.

1. Introduction

Steel–concrete composite structures are becoming increasinglypopular around the world due to the favourable stiffness, strengthand ductility performance of composite systems under seismicloading, and also due to the velocity and ease of erection.Moreover,such a structural typology exhibits better fire resistance character-istics compared to a bare steel structure, if one also considers thehigh probability of fire after a seismic event. Generally, they are en-dowed with columns made with concrete-filled tubes or partiallyencased profiles, because they allow one to satisfymore easily driftlimits and to increase fire resistance.

In some structural typologies, beam-to-column joints directlycontribute to lateral stiffness, and therefore, they play a vital role inbuilding survival during and after a seismic event. As a result, theyrequire a design procedure that incorporates a strength hierarchyin all of its components and mechanisms to ensure yielding

∗ Corresponding author. Tel.: +39 0965875223; fax: +39 0965875201.E-mail address: [email protected] (R. Pucinotti).

0143-974X/$ – see front matter© 2011 Elsevier Ltd. All rights reserved.doi:10.1016/j.jcsr.2011.03.006

at predefined locations, without excessive strength degradation.Even though one fulfills the capacity design criterion for non-ductile components, the joint between concrete filled circulartubes and beams can be critical. In fact, Beutel et al. [1] showedthat joints designed to yield at the column’s face would not besuitable for seismic applications. Conversely, the joint performancesignificantly improved when it was rigid and full strength andyielding shifted in framing beams.

Cheng and Chung [2] investigated the axial load influence onthe shear transfer behaviour in the panel zone of beam-to-columnjoints. Test results showed that all specimens failed by fracture ofwelding between diaphragms and tube while entering in the non-linear regime. Moreover, it was found that the higher the appliedaxial load the better the ductility achieved by joints.

Liew et al. [3] investigated the seismic performance of four steelbeam-to-Concrete Filled steel Tube (CFT) column joints with floorslabs and evaluated both the composite effect of the beam and theseismic behaviour of new connection details, such as taper flangesor larger shear tabs in beam ends. Test results showed that taperedbeam flanges and lengthened shear tabs stiffened at the beam endseffectively moved plastic hinges away from the column face toprevent premature brittle failure of the welds in the beam flange.

http://dx.doi.org/10.1016/j.jcsr.2011.03.006

http://www.elsevier.com/locate/jcsr

http://www.elsevier.com/locate/jcsr

mailto:[email protected]

http://dx.doi.org/10.1016/j.jcsr.2011.03.006

R. Pucinotti et al. / Journal of Constructional Steel Research 67 (2011) 1358–1375 1359

Fig. 1. Probability of exceeding fire phases vs. p.g.a. (cm/s2) after Sekizawa et al. [9].

An innovative type of joint for CFT column-to-steel beamcomposite structures was conceived in [4] in order to ensure asatisfactory joint seismic behaviour. The joint was made of anextended end plate bolted to a CFT column using high-strengthsteel rods on site. The experimental investigation indicated thatthe presence of a floor slab significantly contributed to the jointstrength, and reduced beam sections were effective in shifting thebuckling zone away from the welds close to CFT columns.

On the other hand Azizinamini and Schneider [5] studiedthrough beam joints. External diaphragm and continuous webdetails exhibited a favourable inelastic behaviour, but the flexuralstrength of these joints severely deteriorated under seismicloading. By continuing the flanges through the composite columnshowed adequate strength; however due to excessive slippage,these joints did not dissipate significant inelastic energy. Otherstudies conducted by Elremaily and Azizinamini [6] suggested thatjoints which transfer load from the slab to the column concretecore potentially offer a better seismic performance than joints tothe steel tube alone. In fact, joints to the steel tube alone mayexhibit a large distortion of the tube wall around the connectionregion. Besides, components transferring slab forces into theconcrete core exhibit better strength and stiffness characteristicsthan a simple joint to the tube face. As a result, joints in whichthe beam extends through the composite column generally offerthe most effective method in developing the ideal rigid slab jointbehaviour. Numerical results after Varma et al. [7] suggest thatjoints to the steel tube alone might lead to fracture of the tubewall or to the beam flanges thus preventing full development ofthe plastic bending hinge in the wide flange of the slab.

Different from the non-seismic case, where effective and cheaprigid and/or semi-rigid composite joints can be realised, e.g. [8],the studies presented above indicate that it is difficult to connect acomposite beam to a circular CFT in an economical way. Moreover,it also appears cumbersome to ensure a ductile failure mode ofthese joints with an adequate rotational capacity. Therefore it ismuch more convenient to conceive a rigid full strength joint andshift potential plastic hinges in adjacent beams. This is one of theobjectives pursued for the proposed beam-to-column joints.

As far as the design of steel–concrete composite buildings inEurope is concerned, both the seismic safety and the fire safetyare separately considered and the possible sequence of a fireafter an earthquake is not taken into account. Nonetheless, therisk of loss of lives increases if a seismic-induced fire occursin a building. In this respect, post-earthquake fire is a scenariowith high probability of occurrence as highlighted by recentearthquakes in Northridge 1994 and Kobe 1995. This scenario isa low probability event but with high potential consequences.For instance, following the Kobe earthquake, Sekizawa et al. [9]clearly showed, see Fig. 1, that the probability of exceeding more

and more severe fire phases grows as peak ground acceleration(p.g.a.) increases. In a greater detail from Phase 1 to Phase 5:(i) fire occurs and is growing in Phase 1; (ii) fire is growing andcannot be extinguished by fire security staff; (iii) fire is growingand people can still stay in a room; (iv) fire is fully developed andconfined in a room; (v) fire spreads out to adjacent rooms in Phase5. Moreover, studies of future scenario large-scale earthquakesin San Francisco and Tokyo areas indicated that seismic-inducedfire is an important factor in the subsequent damage to propertyand loss of lives [10]. Thus, it is evident that post-earthquakefire is a design scenario that should be properly addressed inany performance-based engineering design approach, especially inearthquake-prone zones.

To summarise, it appears that few researches were beenconducted on the scenario ‘‘fire following earthquake’’, highlightedas a high potential consequence event. Also, it was highlighted thatthe performance analysis of steel–concrete composite full strengthjoints with CFT under fire remains largely unexplored. It is thetopic covered by this paper describing the design of full strengthbeam-to-column joints to CFT columns able to guarantee: (i) anadequate seismic performance for Medium Ductile frames [11],with a rotation capacity not less than 25 mrad and withoutdegradation of strength and stiffness greater than 20%; (ii) asatisfactory fire resistance of at least 15 min of fire exposureafter an earthquake. These objectives were achieved in this studythrough a balanced combination of numerical and experimentalwork. Joint configurations with prefabricated slabs and steelsheeting slabs were considered later.

In summary, this paper focuses on the seismic activitiesrelevant to the proposed beam-to-column joints together withnumerical simulation performed on moment resisting compositeframes; details on the related research on fire resistance can befound in [12].

To give an overview, Section 2 introduces the concepts adoptedto design both the reference frames and joints under seismic aswell as fire loading. The subsequent experimental programmes onbeam-to-column joints is discussed in Section 3, whilst Section 4presents experimental results. In greater detail, this section alsoreports briefly the results of pre-damage and fire tests, in order toappreciate the post-earthquake fire performance of the examinedjoints. Section 5 presents the calibration of a FE beam-to-columnjoint model on test data and reports the outcome of non-lineardynamic time history analyses carried out on moment resistingreference frames. Conclusions and future research needs aresummarised in Section 6.

2. Design of reference frames under earthquake and fire

This section introduces the main steps used to design bothframes and joints.

1360 R. Pucinotti et al. / Journal of Constructional Steel Research 67 (2011) 1358–1375

1

5

9

13

17

2

6

10

14

18

4

8

12

16

20

a

c

b

Fig. 2. Geometric layout of the reference structures: (a) structure with slabs endowed with prefabricated lattice girders; (b) structure with slabs endowed with profiledsteel sheeting; (c) main frame elevation.

2.1. Frame design

The actions considered for the design of the proposed jointswere derived from the analyses of two moment resisting frameshaving the same structural typology but different slab systems:a slab with prefabricated elements and a composite slab withprofiled steel sheeting. The composite steel–concrete officebuilding from which the studied frames were extracted wasendowed with 5 floors with 3.5 m storey height as shown inFig. 2. It was made up by three moment resisting frames spacedof 7.5 m while it was braced in the transverse direction. Differentdistances between the secondary beams were adopted for the twosolutions to take into account the different load bearing capacitiesof the two slab systems as well as the need to avoid proppingsystems during the construction phase. All slabs were arrangedin parallel with the main frames. The seismic performance of theframes was evaluated by means of non-linear static and dynamicanalyses. The corresponding fire design was made and the fireperformance of the complete frames was evaluated by means ofthe SAFIR program [13], with five different fire scenarios. The fireload followed the ISO 834 Standard Fire Curve (SFC) [14]. Moreinformation on fire analysis and design can be found in [15].

The effect of the seismic loading on frames prior to fire loadingwas taken into account by imposing one loading–unloadingcycle through identical horizontal forces applied to each floor.This loading cycle induced some inelasticity in the beam ends.Subsequently, analyses showed that the impact of the earthquakeon the fire resistance of the analysed frames appeared to be not sosignificant [15].

Slab properties of reference frames were used to define the slabproperties of the tested composite joints, as depicted in Fig. 3.

The properties of the slabs in the vicinity of the beam-to-composite tubular column joints are as follows. The first one is aconcrete slab, 150mm thick, with a prefabricated lattice girder, seeFig. 3(a), with slab reinforcements made of 2 + 2φ12 longitudinalsteel bars and by 5 + 5φ12@100 mm plus 8 + 8φ16@200 mmtransversal steel bars. Two additional longitudinal rebars (1 +

1φ12) are added to support seismic actions. A mesh φ6@200 ×

200 mm completes slab reinforcements. The second one, depictedin Fig. 3(b), is a composite slab, 150 mm thick, with a profiledsteel sheeting andwith the same slab reinforcements. The concreteclass was C30/37 while the steel grade S450C is adopted for thereinforcing steel bars.

All connections between steel beams and slabs were fulland made of Nelson 19 mm stud connectors, with an ultimatetensile strength fu = 450 MPa. In both cases, composite beamswere realised with S355 IPE400 steel profiles, while compositetubular columns were realised with 457 mm circular steel tubeswith 12 mm thickness; column reinforcements consisted of 8φ16longitudinal steel bars and stirrups φ8@150 mm as illustrated inFig. 4. Tables 1 and 2 report both nominal and actual values ofmechanical properties of both steel and concrete.

2.2. Joints design

In order to obtain a beam-to-column joint complying with thedesign performance objectives, and on the basis of some researchwork reported in the introduction, different solutions were


a

b

Fig. 3. Slab reinforcement of (a) a prefabricated lattice girder slab, (b) a composite slab with steel sheeting.

Fig. 4. Column stub and reinforcements capable of hosting a through-column web plate.

Table 1Steel properties.

Element Material Nominal value (MPa) Actual value (MPa)Yield strength Tensile strength Yield strength Ultimate strength

Steel of beam S355 J0 ≥355 510–680 415 560Steel of column S355 J0 ≥355 510–680 483 576Reinforcing steel bars B450C ≥450 ≥540 506 599Profiled steel sheeting Fe E 250G ≥250 / / /

analysed and compared in terms of structural performance andcost effectiveness. Besides, in order to understand the feasibility ofeach proposed solution, several steel shopswere asked to give their

technical advice. Four structural solutions, depicted in Fig. 5, wereanalysed mainly differing from the type of element penetratinginto the column’s concrete core. The parameters taken into


a

b

c

d

Fig. 5. Different solutions of beam-to-column joints: (a) Typology A; (b) Typology B; (c) Typology C; (d) Typology D.

Table 2Concrete properties.

Element Material Nominal value (MPa) Actual value (MPa)Cylindrical characteristic strength (fck,nom) Cylindrical mean strength (fc,act ) Cylindrical characteristic strength (fck,act )

Prefabricated concrete slab C30/37 30.00 41.3535.46Profiled steel sheeting slab C30/37 30.00 38.55

Filling column C30/37 30.00 50.51

account for the evaluation of the above-mentioned solutionswere:(i) the required time to make on shop operations; (ii) the requiredadditional time and cost effectiveness of in-situ operations; (iii)difficulty of erection due to the column’s self-weight; and (iv)the easiness to assemble the different joint parts. By takinginto account all previous considerations, the completely weldedsolution was adopted, see Figs. 5(b) and 6, also enhancing the jointperformance both in terms of strength and rigidity.

For the assembling of welded joint specimens, also thecomplexity of the erection sequence was considered at the

Laboratory of the Trento University. Therefore, the welding ofdifferent elements depicted in Fig. 7 were realistically simulated.

In order to consider the fact that sometime patented weldersare not available, in a parallel Italian study [16], a bolted solutionwas conceived to guarantee easiness of assembly [17]. As a resultthe beam-to-column composite joint was made of two horizontaldiaphragm plates and a vertical through-column plate attached tothe tube by groovewelds as illustrated in Fig. 8. Flanges andweb ofeach beamwere connected to the horizontal plates and the verticalplate respectively by two and three rows of bolts M27 10.9.


Fig. 6. Welded beam-to-column joint: (a) solutionwithout Nelson stud connectorsaround the column; (b) solution with Nelson stud connectors around the column.

In the aforementioned joint solutions, the seismic design wasconceived to provide both adequate overstrength and stiffnesswith respect to the connected beams, thus forcing plastic hingesin adjacent beams. Joints were detailed by using the componentmethod in agreement with Eurocode 3 Part 1–8 [18] and Eurocode8 Part 1 [11] as shown schematically in Fig. 9. Through thismethod,the joint was simulated by a series of different componentsand these components were modelled by an elastic springcharacterised by specific stiffness and strength. The appropriatecoupling in parallel and series of these springs provided the globalstiffness of the joint.

In particular, in the application of the component method thecomposite column was assumed to be infinitely rigid. Moreover,as Eurocode 3 Part 1.8 [18] does not provide formulas for somecomponents activated under sagging moments and joints withtubes, the concrete slab in compression and both the top andbottom collar plates (Fig. 8(b)) were characterised by means of FEmodels set with the ABAQUS software [19]; indeed, stiffness andstrength of these components were defined by means of refinedFinite Element (FE) models of the joints as depicted in Fig. 10.Details of the modelling of components ‘‘slab in compression’’ and‘‘bottom collar plate in tension’’ are illustrated in Figs. 11 and13, respectively. In the FE models, eight-node linear brick C3D8Relements were employed to model all joint components, whilefour-node linear tetrahedron C3D4 elements were employed tomodel the transition zones between different meshes. Reinforcingbars were modelled by linear beam elements B31. The consideredcomponents were different in sagging and hogging moment. Theyare summarised in the following sections.

2.2.1. Activated components under sagging bending momentThe following components were considered:

(i) concrete slab in compression: both the stiffness and strengthof this component were defined by means of refined FEmodels of the joint including friction between the slab andcolumn. In detail, the ‘‘Concrete damaged plasticity’’ materialmodel available in ABAQUS [19] was employed for concrete.Depending on the level of friction, the distribution of thecompression forces in the slab under sagging bendingmomentwas found to be different. It was localised in front of thecolumn for a friction coefficient equal to 0.35 as indicated inFig. 11(a), while it spread over a more extended portion ofthe slab owing to increasing values of the friction coefficient.In detail, the diffusion angle β becomes greater than 80°for a friction coefficient of about 1 as shown in Fig. 11(b).The results showed that, in order to activate the transfermechanisms proposed in Eurocode 8 Part 1 Annex C [11],i.e. the front Mechanism 1 and the strut and tie Mechanism 2both shown in Fig. 12(a) and (b), respectively, it was necessaryto increase the level of friction between the concrete slab andthe composite column. Accordingly, in order to better appraisethe activation of these transfer mechanisms in the slab, adesign solution with 19 mm Nelson stud connectors weldedaround the column was adopted as indicated in Figs. 6(b)and 7(b). The corresponding solution without Nelson Studs isillustrated in Figs. 6(a) and 7(a). The maximum value of theforce transmitted to the slab, via Mechanism 1 depicted inFig. 12(a), was taken as:

FRd1 = bb · deff · fcd (1)

where

deff is the overall depth of the slab in case of solidslabs or the thickness of the slab above the ribs ofthe profiled sheeting for composite slabs;bb = D · sin

β

2

is the bearing width of the

concrete of the slab on the column;D is the diameter of the column;fcd is the design strength of the concrete slab.

The compressed concrete struts inclined at 45° to the columnsides i.e. the Mechanism 2 shown in Fig. 12(b), was taken as:

FRd2 = 0.7 · D · deff · fcd. (2)

(ii) upper horizontal plate in compression: the effectivewidth bpf ofplates was approximately assumed to be half of the total platewidth equal a 660 mm, see Figs. 12(d) and 13(b); the strengthof this component was defined as:

Fpf ,Rd1 = Apf ,Rd1 · fyd (3)

where:• fyd is the design strength of steel.• For Apf ,Rd1 (see Fig. 13(b)):

– if the neutral axis is in the slab (x < hs), then Apf ,Rd1 = 0;– if x > hp, then Apf ,Rd1 = bpf · tpf

with tpf the collar plate depth.(iii) web plate in bending: the web plate was designed to resist

shear forces. However, it also provided a non-negligiblecontribution to the joint resisting moment. Moreover, thepossible onset of local instability mechanisms was consideredaccording to Eurocode 3 Part 1-1 in Section 5.6 [20]. Also, thecomplete yielding of the two areas of the web plate separatedby the neutral axis, i.e. the compression zone and the tensionzone, was considered.


a

b

Fig. 7. Beam-to-column connection: (a) solution without Nelson stud connectors around the column; (b) solution without Nelson stud connectors around the column.

The strengths of these components, depicted in Fig. 12(c),were respectively:

Fpw,Rdc = (x − a − hp) · tpw · fyd (4)

and

Fpw,Rdt = (hpw − x) · tyw · fyd, (5)

with twp vertical plate thickness.

In Eq. (4), Fpw,Rdc was limited to the buckling resistance ofthe plate according to the Eurocode 3 Part 1-1.

(iv) lower horizontal plate in tension:

Fpf ,Rd2 = Apf ,Rd2 · fyd (6)

where, in this particular case, Apf ,Rd2 = Apf ,Rd1.


Fig. 8. Bolted solution for the beam-to-column joint: (a) global view; (b) details of joint components.

Fig. 9. Mechanical model of a steel–concrete composite interior joint subjected to sagging moments.

2.2.2. Activated components under hogging bending momentFor hogging bending moment, the following components were

considered:

(i) reinforcing bars in tension: rebars in the slab were designedaccording to the ECCS Report N.109 [21] and Eurocode 4 Part1 [22]. The reinforcing bars considered in the design werethose contained in the effective width of the slab according toSection 5.4.1.2 of Eurocode, 4 Part 1 [22]. The strength of thiscomponent was taken as equal to:

Ftr,s,Rd = Ar,s · fsd. (7)

Ar,s is the cross-sectional area of the longitudinal reinforce-ment in row r within the effective width of the concrete flangedetermined for the cross-section at the joint;fsd is the design strength of rebar steel;

(ii) upper horizontal plate in tension, web plate in bending and lowerhorizontal plate in compression: the same approaches as theones followed for these components activated under saggingmoments were adopted.

2.2.3. Design criteriaBeam-to-column joints were designed to be rigid and full-

strength joints with respect to adjacent beams. Thus in agreementwith the Eurocode 8 Part 1, they satisfied the following criterion:

Mj,Rd ≥ 1.1 · γov · Mb,pl,Rd (8)

where Mj,Rd is the resisting moment of the beam-to-column jointand Mb,pl,Rd is the resisting moment of the adjacent compositebeam.


Fig. 10. Abaqus simulations of a joint endowed with (a) a slab with a steel sheeting; (b) a slab with a prefabricated lattice girder.

Fig. 11. Distribution of compression stresses in the slab for: (a) friction coefficient equal to 0.35; (b) friction coefficient equal to 1.

Moreover, full groove and fillet welds were used. In detail, theysatisfied the relationship:

Rd ≥ 1.1 · γov · Rfy (9)

with an overstrength factor γov = 1.25, where Rfy is the plasticresistance of the connected element.

2.2.4. Thermal analysesThe ABAQUS FE software analysis was also used for the thermal

analyses of the joints: the relevant meshes are shown again inFig. 10. In a greater detail, all components such as columns,beams, slabs and welds were modelled using eight-node linear

brick DC3D8 elements while four-node linear tetrahedron DC3D4elements were employed, to model the transition zone betweendifferent meshes. Fig. 10 indicates temperature distributions injoints endowed with a steel sheeting slab and a prefabricatedlattice slab, respectively.

All steel parts exposed to fire increased their temperaturevery quickly, reaching temperatures of about 600 °C after only15 min of exposure [15]. Conversely, both in concrete and in steelcomponents embedded or close to concrete, i.e. reinforcing bars,the collar plate close to the slab and the vertical plate passingthrough the column, the temperature did not increase so quickly,and remained close to ambient temperature. On the basis of this


a

d

c

b

emn

a

hphs

fy

hpwfy

fcd

g

Fpw,Rdc

Fpw,Rdt

x

Fig. 12. Strut and tie mechanisms assumed in the slab: (a) Mechanism 2; (b) Mechanism 1; (c) frontal view of the joint; (d) details of the plates.

FE result, in order to enhance the seismic-induced fire behaviour,the joint design included: (i) a welded top collar plate close tothe slab; (ii) a web-through plate partially embedded into theconcrete of column; (iii) two additional φ12 rebars (1 + 1φ12longitudinal steel bars) in order to take into account the seismic-induced damage [15].

3. Test programme

The experimental programme involved the execution often seismic tests and six fire tests on full-scale substructuresrepresenting interior and exterior welded beam-to-column joints.Seismic testswere carried out at the University of Trento and at theUniversity of Pisa, considering both cyclic andmonotonic loadings.Fire tests on pre-damaged joints were conducted at the BuildingResearch Establishment, UK, with asymmetric loading on joints tosimulate adjacent primary beams of different lengths. Additionalinformation on the full research programme can be found in [15].

3.1. Seismic tests

Both interior and exterior welded beam-to-column joints withconcrete filled tubeswere tested. Six experimental tests on interior

Table 3Experimental programme on interior composite joints.

Label Test protocol Type of specimen

S-IWJ-P1 Cyclic Prefabricated lattice girderS-IWJ-P2 Cyclic Prefabricated lattice girderS-IWJ-PM Monotonic Prefabricated lattice girderS-IWJ-S1 Cyclic Steel sheetingS-IWJ-S2 Cyclic Steel sheetingS-IWJ-SM Monotonic Steel sheetingIWJ-P = Interior welded joint with prefabricated slabIWJ-S = Interior welded joint with steel sheeting slab

joints (4 cyclic and 2 monotonic tests) were carried out at theUniversity of Trento while the remaining four experimental testson exterior joints (3 monotonic and 1 cyclic test) were conductedat the University of Pisa.

Within the present paper, only tests carried out on interiorjoints are discussed. In particular, Table 3 reports the experimentalwork-programme on interior composite joints, while the relevanttesting equipment is sketched in Fig. 14. Relevant instrumentationis detailed in [15]. Specimens were subjected to monotonic andcyclic loading up to collapse, according to the ECCS stepwiseincreasing amplitude loading protocol [23], modified with the SAC


Fig. 13. FE model of a plate in tension: (a) elastic stresses; (b) inelastic stresses and effective width.

procedure [24], by using a horizontal yielding displacement ey =

0.005h = 17.5 mm, where h represents the storey height. The testprogramme was conceived to analyse different design solutions.For consistency with other projects, no axial load was applied tothe column during testing.

3.2. Pre-damaged tests

The objective of this experimental programme carried out atthe Building Research Establishment (BRE), UK consisted in theevaluation of the fire resistance of joints partly damaged by anearthquake. Therefore in order to estimate damage, simulations onthe frames introduced in Section 2.1, were performed [25].

As at the BRE, it was not possible to perform cyclic testscapable of simulating damage caused in joints by an earthquake,monotonic forces were applied to four specimens listed in Table 4in order to produce the same damage of an high intensityearthquake. The frame analyses carried out to assess the damagesare described in Section 5.

Experimental data of joints were used to define both hystereticlaws in IDARC-2D [27] and the damage index value D — thatranges between 0 and 1 — according to the Chai and Romstadcriterion [29]. This criterion, thatmodifies thewell-known damage

Specimen

Fig. 14. Lateral view of the test set-up at the University of Trento.

model proposed by Park and Ang [30], based on a linearcombination of damage owing to excessive deformation andcumulative plastic strain energy, accounts for the energy Ehm


Table 4Summary of results for joints under pre-damage loading.

Specimen label Joint position Max. est. load (kN) Max. deflection (mm) Max. est. moment (kN m) Max. rotation (mrad)

D-IWJ-S1 Interior 424 32 887 10.0D-EWJ-S3 Exterior 258 58 541 7.4D-IWJ-P1 Interior 425 21 893 7.3D-EWJ-P3 Exterior 398 110 836 12.6

IWJ-P = Interior welded joint with prefabricated slabIWJ-S = Interior welded joint with steel sheeting slabEWJ-P = Exterior welded joint with prefabricated slabEWJ-S = Exterior welded joint with steel sheeting slab

Table 5Joint specimens under fire loading.

Specimen label Maximum atmosphere temperature (°C) Maximum steel temperature (°C) Test duration (min) Comments

FD-IWJ-S1 1024 747 40 Test terminated due to runawaydeflection. Full depth crackingand separation between steelsheet and slab

F-IWJ-S2 970 966 60 No permanent deformationFD-EWJ-S3 972 963 60 No permanent deformationFD-IWJ-P1 1196 810 34 Test terminated due to runaway

deflection. Local buckling of thelower flange

F-IWJ-P2 982 721 45 Test terminated due to runawaydeflection. Cracking and spallingof concrete

FD-EWJ-P3 944 726 56 Test terminated due to runawaydeflection. Local buckling of thelower flange

dissipated by the member at the design strength Pu during amonotonic loading process; only the surplus of cumulative energy(Eh − Ehm) is considered significant to damage, Eh being thehysteretic total energy at the design strength Pu.

In these conditions, the damage index reads:

D =∆M

∆um+

β∗(Eh − Ehm)

Py∆um(10)

where

β∗ is an empirical factor determined by experimentaldata;∆M is the maximum response displacement;∆um is the maximum response displacement under amonotonic loading;Py is the yield strength.

By imposing D = 1, the relation that defines the damage limitdomain is obtained as follows:

EhPy∆um

=1β

−

1β∗

∆M

∆um(11)

where 1β∗ =

1β

−Ehm

Py∆um.

With β the strength degradation parameter calibrated for themodified model.

In order to avoid initial negative values of D, a parameter αwhich weights the peak response term ∆M/∆um is introduced inthe original formulation proposed in [28]. Thereby, starting fromthe relation

D = α∆M

∆um+

βEhPy∆um

(12)

and imposing both D = 1 and Eh = Ehm for the case of amonotonictest, one gets:

α = 1 −βEhmPy∆um

. (13)

As a result, the damage index D reads:

D =∆M

∆um+

β

Py∆um

Eh − Ehm

∆M

∆um

. (14)

Eq. (14) is equivalent to Eq. (10) in terms of damage limit domain,but the damage evolution is slightly different.

3.3. Fire tests

A total of six fire tests were carried out. Relevant specimens arelisted in Table 5. As stated in the Introduction, the performancecriterion for the examined joints was to exhibit a 15 min fireresistance once damaged by earthquake effects without anyadditional fire protection. In this respect and for the sake ofcomparison with tests available in the literature, the fire loadfollowed the ISO 834 [14], rather than a natural fire or a parametriccurve [26].

4. Test results

In this paper only the seismic tests results, carried out at theuniversity of Trento, on interior joints, are presented. Pre-damagedtests and fire tests results are fully described in [15].

4.1. Seismic test results

Monotonic test results for the specimens S-IWJ-PM and S-IWJ-SM without Nelson connectors welded around the columns areplotted in Fig. 15(a) and (b), respectively. A favourable behaviour interm of strength of the specimen S-IWJ-PM endowed with latticegirders slab was evident owing to a better composite action inthe yielded section. Experimental tests carried out on interior andexterior joints showed that joints exhibited enough stiffness andstrength to be considered as rigid and full strength according toEurocode 3 Part 1-8 and Eurocode 4 Part 1-1 [18,22].


a b

Fig. 15. Monotonic test results for interior joints. Force vs. interstorey drift; (a) prefabricated lattice girder slab; (b) steel sheeting slab.

a b

c d

Fig. 16. Results of cyclic tests on interior joints. Force vs. interstorey drift (a), (b) prefabricated lattice girder slab; (c), (d) steel sheeting slab.

The force–interstorey drift relationships of S-IWJ-P1 and S-IWJ-P2 specimens with electro-welded lattice slabs and without/withNelson connectors around the column are illustrated in Fig. 16(a)and (b), respectively. Plastic hinges developed in beams adjacentto joint and progressive deterioration of strength and stiffnesswas associated with beam flange buckling. Failure was due tobeam flange cracking. In detail not so much difference was notedby comparing the global behaviour of these two specimens: thisimplies that the presence of the Nelson connectors around thecolumn did not influence the specimen response, the yieldingmechanism being localised in the composite beams. Similarresults were obtained for specimens S-IWJ-S1, see Fig. 16(c),and S-IWJ-S2, see Fig. 16(d), endowed with slabs with profiledsteel sheeting and without/with Nelson connectors around thecolumn, respectively. Differently from the two previous cases,i.e. the specimens endowed with a prefabricated concrete slab,in both beam and weak section of the joint, the neutral axis waslocated on the beam web for the sagging moment. This happenedbecause the steel sheeting slab was more damaged comparedto the prefabricated lattice slab. The overall force–interstoreydrift relationships relevant to plastic hinges formed in thecomposite beams exhibited a hysteretic behaviour with largeenergy dissipation without evident loss of resistance and stiffness.

Moreover, Fig. 17 shows the cyclic moment–rotation rela-tionships of both prefabricated lattice girder slab specimens andsteel sheeting slab specimens, respectively. The experimental testsshowed a remarkable and progressive deterioration of strength,stiffness and energy absorption capacity as a consequence of theformation of a plastic hinge associated with local buckling of beamflanges. Based on the experimental data one can state that the per-formance of joints was satisfactory from a seismic point of view,satisfying the requirements of medium ductility structures withplastic rotations greater than 25 mrads and without degradationof strength and stiffness greater than 20% as required by Eurocode8 Part 1 [11] for moment resisting steel frames.

As mentioned in Section 2.2, a bolted solution of beam-to-column joints, that did not require welding on-site was proposed.The relevant force–interstorey drift relationships of bolted jointsS-IBJ-P2 with prefabricated slabs and S-IBJ-S2 with steel sheetingslabs, are illustrated in Fig. 18(a) and (b) with the correspondingwelded solutions. From the responses, one can deduce that thedissipation capabilities of the beam-to-column solutions are notinfluenced by the limited performances of bolted joint owing toyielding away from joints.


a b

c d

Fig. 17. Results of cyclic tests on interior joints. Moment vs. rotation; (a), (b) prefabricated lattice girder slab; (c), (d) steel sheeting slab.

a b

Fig. 18. Force vs. interstorey drift response of an adjacent beam plastic hinge relevant to both bolted and welded solutions; (a) prefabricated lattice girder slab; (b) steelsheeting slab.

Table 6Damage index relevant to welded joints.

Joint position Joint with steel sheeting slab Joint with prefabricated slab

Interior 0.27 0.42Exterior 0.31 0.50

4.2. Results from pre-damaged tests

The pre-damaged tests were performed with the objectiveto simulate joint damage owing to strong seismic events. Theevaluation of the damage index is explained in Section 5 togetherwith relevant joint results.

Only one of each type of interior joints was subjected topre-damaging tests while all exterior joint specimens were pre-damaged [15].

Table 4 summarises significant results fromdamage testswhichare consistent with damage values reported in Table 6.

4.3. Results from fire tests

In agreement with the project objectives, both pre-damagedand undamaged specimens were subjected to fire loading.

Fig. 19(a) shows the test set-up used for fire tests. In detail,both the furnace zone and the forces applied to beams canbe seen in order to simulate the accidental load combinationon beams of unequal length. The temperature vs. time curveimposed to the specimens FD-IWJ-S1 - F-IWJ-S2 and FD-IWJ-P1 - F-IWJ-P2 is shown in Fig. 19(b) and (c), respectively.Specimens FD-IWJ-S1 and F-IWJ-S2 endowed with profiled steelsheeting slabs exhibited failure owing to an excessive rate ofdeflection at approximately 40 min. The test on specimen FD-IWJ-S1 terminated after approximately 34 min owing to runawaydeflection. Following the fire test, the profiled steel sheetingseparated from the slab; then the slab cracked both along thesurface and through the depth with extensive buckling at 40 minboth of the lower flange and theweb of the adjacent east beam. FD-IWJ-P1 and F-IWJ-P2 specimens endowedwith prefabricated slabsendured one hour of fire; however, in both cases specimens werevery close to failure as indicated, in Fig. 19(c), by an increasing rateof deflections towards the end of the test. However, at this stage,therewas no permanent deformation and no sign of any significantdamage from fire tests. In detail, it can observed that: (i) therewas no noticeable difference in the fire performance betweenpre-damaged and undamaged specimens both with precast andsteel sheeting slabs; this result was in agreement with damage


c

b

a

Fig. 19. (a) Load introduction in the specimen; (b) comparison between pre-damaged (FD-IWJ-S1) and undamaged (F-IWJ-S2) steel sheeting specimens;(c) comparison between pre-damaged (FD-IWJ-P1) and undamaged (F-IWJ-P2)precast specimens.

values reported in Table 6 and with the inherent design safety ofcomposite joints [15,11]; (ii) precast slabs performed better thanthe corresponding specimenswith steel sheeting at a fire exposurein excess of the 15min required also in fire tests; (iii) all specimensexhibited a favourable fire resistance. Relevant results of fire testsare summarised in Table 5 and more details are reported in [15].

5. Calibration and numerical analyses

Data obtained from monotonic and cyclic tests described inSection 4.1 were used to calibrate a joint model, in order to bothsimulate damage induced by earthquakes and to perform seismicsimulations on several moment resisting frames. The model usedto perform these analyses is depicted in Fig. 20(a) and relied on:(i) two parallel springs A at the end of each beam, see for detailsFig. 20(b), connected to a rigid panel-each spring A was used tomatch the properties of a composite beam under sagging andhogging bending moments, respectively-; (ii) a spring B used toconnect this panel to the shear panel of the column—the mainpurpose of this additional springwas to take into account the sheardeformation of the joint.

The calibration of the springs used in the joint model describedabove was done twice for each type of slab: (i) for the beam-to-column joints without Nelson stud connectors welded around thecolumn; (ii) for the joints equipped with Nelson stud connectorsaround the column. The IDARC-2D program [27] was used to carry

out simulations. A hysteretic lawwas used to take into account theseismic degradation of the joint according to a modification of theBouc–Wen model implemented by Silvaselvan and Reinhorn [28]in IDARC-2D. The actual measured properties of both concrete andsteel, reported in the Tables 1 and 2, were used in the modelto match as accurately as possible experimental data. The actualbehaviour was validated via the comparison between: (i) theamount of energy dissipated by the whole joint model and theenergy wiped out by the specimen during each laboratory test;(ii) the hysteretic behaviour exhibited via simulations and the oneshown by each specimen during testing. The quality of calibrationcan be assessed from Fig. 20(c), where with reference to thespecimen S-IWJ-S1, experimental and numerical data are verywell-correlated, with an error in the dissipated energy of about6%. As shown in Section 3, specimens were endowed with sensorsin order to measure the relative rotations of joint components.As a result, it was also possible to define damage limit domainsfor joints, see Eq. (11), and relevant β parameters for Eq. (14),i.e. β = 0.256 for sagging moment and β = 0.066 for hoggingmoment, respectively.

Once the hysteretic model of joints was calibrated, the twoframes depicted in Fig. 2, were simulated by the model illustratedin Fig. 20(d), bymeans of the IDARC-2D software [27]. In detail, fourframe simulations were performed — with electro-welded latticeslabs, with steel sheeting slabs, with and without the column’sNelson studs — in order to evaluate the seismic behaviour of theproposed joints as well as the relevant overall frame performance.

In agreement with the joint responses highlighted in Sec-tion 4.1, Incremental Dynamic Analyses (IDA) were carried out onthe four frames, in order to find some correlations between therequired plastic rotations for medium ductility moment frames,i.e. about 25 mrad specified in Eurocode 8, Part 1 [10], interstoreydrifts and the relevant demands in p.g.a. In this respect, three in-cremental time history analyses were performed with each frame.The accelerograms used to perform the simulationwere artificiallygenerated tomatch type 1 response spectrumand soils typeA, typeB and type D, respectively. In detail, Fig. 21(a) shows the responsespectra of the artificially generated accelerograms matching itscorresponding Eurocode 8 Part 1 spectrum [11]; whilst Fig. 21(b)highlights an artificial accelerogram that matches soil type A. Eachaccelerogram was used to perform twenty time history analyseswith different peak ground accelerations ranging from 0.1g to2.0g , with an interval of 0.1g .

From the IDA results, it was possible to estimate joint damageowing to a seismic event by using the model described inSection 3.2. Relevant values of damage indices are reported inTable 6 and they were not so high for both interior and exteriorjoints. Following the indications reported in [31] a damage indexof D < 0.66 would correspond to repairable damage, whereas0.66 < D < 1.0 would entail irreparable damage. These figuresindicate that damage in joints was limited and repairable.

Moreover, frame interstorey drifts ranged between 3.41 and5.80% to which corresponded p.g.a. ranging between 1.34 and1.92g. As an example, some results provided by these analysesfor joints endowed with Nelson studs around the columns aregathered in Table 7 for soil type B. It is evident, that therequirement of 25mrad of plastic rotations to the composite jointsunder examination, see Column 3 of Table 7, entails large p.g.a. andcorresponding significant interstorey drifts.

The aforementioned IDA frame analyses finally allowed be-haviour factors to be estimated. They are reported in Table 8 forthe case of Nelson stud connectors welded around the column. Be-haviour factor values were of about 4. These results confirm thatthe proposed joints can be exploited with moment frames of Duc-tility Class M.


0

17.50m

14.00m

10.50m

7.00m

3.50m

0.00m

Interior Joint

a

b

c

d

Fig. 20. (a) FE model of an interior composite joint; (b) details of a joint model; (c) comparison between experimental and simulated data for the S-IWJ-S1 specimen;(d) FE model of an entire moment resisting frame.

Table 7Seismic demands and drifts for plastic joint rotations from frame analyses.

Specimens endowed with electro-welded lattice slabs and with column’s Nelson studsTime history p.g.a. (g) Plastic joint rotation (mrad) Location in Fig. 2(c) Interstorey drift (%)

Accelerogram 1 1.92 26.1 10 5.80Accelerogram 2 1.80 26.2 6 4.55Accelerogram 3 1.50 26.7 2 4.19

Specimens endowed with steel sheeting slabs and with column’s Nelson studs

Time history p.g.a. (g) Plastic joint rotation (mrad) Location in Fig. 2(c) Interstorey drift (%)

Accelerogram 1 1.72 23.0 7 5.36Accelerogram 2 1.60 29.4 4 3.68Accelerogram 3 1.40 26.1 1 3.98


1.60

1.40

1.20

1.00

0.80

0.60

0.40

0.20

0.00

Sa/g

0.0 0.5 1.0 1.5 2.0 2.5 3.0 3.5 4.0

T [s]

0.50

0.40

0.20

0.30

0.10

0.00

-0.10

-0.20

-0.30

-0.40

-0.50

a/g

T [s]

0.0 2.0 4.0 6.0 8.0 10.0 12.0 14.0 16.0

a

b

Fig. 21. (a) Response spectra of accelerograms matching type 1 Eurocode 8 Part 1spectra; (b) Time history of an artificial accelerogram matching spectrum type B.

Table 8Behaviour factors from IDA analyses.

IDA A IDA B IDA D

Steel sheeting slab 4.24 4.10 4.00Prefabricated slab 3.75 3.94 5.03

6. Conclusions

This paper investigated the performance of steel–concretecomposite full strength joints endowed with concrete filled tubesdesigned with a multi-objective methodology, in order to considerthe scenario ‘‘seismic actions followed by fire’’. In fact to faceseismic-induced fire, beam-to-column joints were enhanced withwebs going through the columns, horizontal connectors weldedaround the column at the slab level and specific reinforcementlayouts.

Both the experimental programme and the results providedby seismic tests were presented and discussed. In detail, resultsshowed that joints were rigid and full strength; consequently,energy dissipating mechanisms in moment resisting framesendowed with these joints can entirely rely on the formation ofplastic hinges at adjacent beam ends. Moreover, results showedthat seismic-induced damage does not influence the fire resistanceof the examined joints; and that beam-to-column joints endowedwith prefabricated slabs exhibited a better performance thanjoints incorporating steel sheeting. Finally, non-linear numericalsimulations clearly showed that these joints can be deemedadequate for moment resisting frames of Medium ductility ClassM characterised by a behaviour factor of about 4.

Acknowledgements

The writers are grateful to the European Union for financialsupport under the project PRECIOUS-RFS-CR-03034 and to thepartners, ArcelorMittal, Building Research Establishment Ltd.,Ferriere Nord S.P.A., University of Navarra and University of Pisa.Nonetheless, conclusions of this paper are those of the authors anddo not necessarily reflect the view of the sponsor’s agency and ofpartners.

References

[1] Beutel J, Thambiratnam D, Perera N. Cyclic behaviour of concrete filled steeltubular column to steel beam connections. Engineering Structures 2002;24:29–38.

[2] Cheng C, Chung L. Seismic performance of steel beams to concrete-filled steeltubular column connections. Journal of Constructional Steel Research 2003;59:405–26.

[3] Liew JYR, Teo TH, Shanmugam NE. Composite joints subject to reversal ofloading—part 1: experimental study. Journal of Constructional Steel Research2004;60(2):221–46.

[4] Li X, Xiao Y, Wu YT. Seismic behavior of exterior connections with steel beamsbolted to CFT columns. Journal of Constructional Steel Research 2009;65:1438–46.

[5] Azizinamini A, Schneider SP. Moment connections to circular concrete-filledsteel tube columns. Journal of Structural Engineering, ASCE 2004;130(2):213–22.

[6] Elremaily A, Azizinamini A. Experimental behavior of steel beam to CFTcolumn connections. Journal of Constructional Steel Research 2001;57(9):1009–19.

[7] Varma AH, Ricles JM, Sause R, Lu LW. Experimental behavior of high strengthsquare concrete-filled steel tube columns. Journal of Structural Engineering,ASCE 2002;128(3):309–18.

[8] Schaumann P, Zhao B, Bahr O, Renaud C. Fire performance of external semi-rigid composite joints. In: Sixth international conference on structures in fire.SiF‘10. Michigan State University. MI, USA, 2010.

[9] Sekizawa A, Ebihara M, Notake H. Development of seismic-induced fire riskassessmentmethod for a building. In: Proceedings of seventh 7th internationalsymposium. IAFSS. 2003.

[10] Scawthorn C. Fires following the northridge and kobe earthquakes. In:Gaithersburg, MD, Beall, KA. editors, NISTIR 6030; US/Japan governmentcooperative programonnatural resources. UJNR. Fire research and safety. 13thjoint panel meeting. March 13–20 1996. vol. 2. 1997. pp. 325–35.

[11] EN 1998-1. Eurocode 8: design of structures for earthquake resistance—part1: general rules, seismic actions and rules for buildings. 2005.

[12] Bursi OS, Ferrario F, Pucinotti R. Seismic-induced fire analysis of steel–concretecomposite beam-to-column joints: welded solutions. In: Engineering confer-ences international. Composite construction VI. Copper mountain. 2008.

[13] Franssen JM. SAFIR. A thermal/structural programmodelling structures underfire. Engineering Journal, AISC 2005;42(3):143–58.

[14] ISO 834-1. Fire-resistance tests—elements of building construction— part 1:general requirements. 1999.

[15] Bursi OS, et al. editors. Final report PRECIOUS project contr. N. RFSCR-03034.Prefabricated composite beam-to-concrete filled tube or partially reinforced-concrete-encased column connections for severe seismic and fire loadings.2008.

[16] RELUIS—Line 5-responsibles: R. Zandonini and O. S. Bursi. Design bytesting based on the capacity design of structural elements and joints insteel–concretemoment-resisting frames and bridges. Research project fundedby Italian Civil Protection. 2005.

[17] Bursi OS, Ferrario F, Pucinotti R, Zandonini R. Seismic-induced fire anal-ysis of steel–concrete composite beam-to-column joints: bolted solutions.In: Engineering conferences international. Composite construction VI. Coppermountain. 2008.

[18] EN 1993-1-8. Eurocode 3: design of steel structures—part 1–8: design of joints.2005.

[19] Hibbitt, Karlsson and Sorensen 2000. ABAQUS user’s manuals. Pawtucket (RI):1080 Main Street. 02860.

[20] EN 1993-1-1. Eurocode 3: design of steel structures—part 1-1: general rulesand rules for buildings. 2005.

[21] ECCS. Design of composite joints for buildings European convention forconstructional steelwork. Brussels (Belgium): ECCS Publication No. 109; 1999.

[22] EN 1994-1. Eurocode 4: design of composite steel and concrete structures—part 1-1: general rules and rules for buildings. 2004.

[23] ECCS. Recommended testing procedures for assessing the behaviour ofstructural steel elements under cyclic loads. ECCS Publication No. 45. 1986.

[24] Karl F, Helmut K, Robert S. Protocol for fabrication, inspection, testing, anddocumentation of beam–column connection tests and other experimentalspecimens. Report no. SAC/BD-97/02. Sacramento (CA, USA): SAC JointVenture; 1977.

[25] Bursi OS, Cajot L-G, Ferrario F, Gracia J, Plumier A, Pucinotti R. et al. Seismicperformance ofwelded steel–concrete composite beam-to-column jointswithconcrete filled tubes. Beijing (China): 14 WCEE; 12–14 October. 2008.


[26] EN 1991-1-2. Eurocode 1: actions on structures—part 1–2: general actions—actions on structures exposed to fire. 2004.

[27] Valles RE, Reinhorn AM, Kunnath SK, Li C, Madan A. IDARC2D version 4.0: aprogram for the inelastic damage analysis of buildings. Tech. report NCEER-96-0010, 1–8. 1996.

[28] Silvaselvan MV, Reinhorn AM. Hysteretic model for cyclic behaviour ofdeteriorating inelastic structures. Technical report MCEER-99-0018. 1999.

[29] Chai YH, Romstad KM. Correlation between strain-based low-cycle fatigueand energy-based linear damage models. Earthquake Spectra 1997;13(2):191–209.

[30] Park YJ, Ang AHS. Mechanistic seismic damage model for reinforced concrete.Journal of Structural Engineering 1985;111(4).

[31] Williams MS, Sexsmith RG. Seismic damage indices for concrete structures: astate-of-the-art review. Earthquake Spectra 1995;11(2):319–49.

post-earthquake fire and seismic performance of welded steel–concrete composite beam-to-column...

Documents