cyclic behaviour of typical metal connectors for cross-laminated (clt) structures

ORIGINAL ARTICLE

Cyclic behaviour of typical metal connectorsfor cross-laminated (CLT) structures

Igor Gavric • Massimo Fragiacomo •

Ario Ceccotti

Received: 19 November 2013 / Accepted: 5 March 2014

� RILEM 2014

Abstract An extended experimental programme on

typical cross-laminated (CLT) connections was per-

formed at IVALSA Trees and Timber Institute. The

paper discusses the results of monotonic and cyclic

tests in shear and tension (pull-out) carried out on

hold-downs and steel angle brackets used to anchor the

wall panels to foundations or to connect wall panels to

floor panels. Mechanical properties such as strength,

stiffness, energy dissipation, impairment of strength

and ductility were evaluated and are critically dis-

cussed in the paper. Significant ductility and energy

dissipation was attained in most of the tests. Never-

theless, brittle failure modes were observed in some

tests, indicating the need for introduction of capacity

based design principles for CLT connections. The

overstrength factors, which are needed for capacity

based design, were also evaluated for the different

types of connection tested. A comparison between the

test results and the analytical formulas provided by

current codes of practice and new proposals is also

provided. The approach developed by Uibel and Blaß

gives slightly more accurate CLT metal strength

predictions compared to the existing formulas in

Eurocode 5. Both approaches lead to very conserva-

tive results. However, analytical models for the

prediction of CLT metal connectors’ stiffness signif-

icantly overestimate the experimental values. There-

fore, it is recommended that currently only

experimental strength and stiffness values of hold-

downs and angle brackets be used in seismic analyses.

Some proposals to improve the mechanical perfor-

mance of metal connectors in terms of strength and

stiffness are also given based on this experimental and

analytical study.

Keywords CLT metal connectors �Cyclic tests �Analytical models � Capacity based design �Overstrength ratio � Ductility � Timber

1 Introduction

Cross-laminated timber panels (CLT or Xlam) are

used more and more as construction material for

buildings, both as load-bearing elements (floor, wall

and roof panels, linear beam members, etc.) and non-

load-bearing elements. Applications include residen-

tial and non-residential buildings, reconstructions,

I. Gavric (&) � M. Fragiacomo

Department of Architecture, Design and Urban Planning,

University of Sassari, Palazzo del Pou Salit, Piazza

Duomo 6, 07041 Alghero, Italy

e-mail: [email protected]

M. Fragiacomo


A. Ceccotti

CNR IVALSA Trees and Timber Institute, S. Michele

all’Adige, Trento, Italy


Materials and Structures

DOI 10.1617/s11527-014-0278-7

extensions and upgrades, building retrofits, and multi-

storey mid-rise and high-rise construction.

Since such panels are fairly stiff in their plane,

particularly when obtained by gluing adjacent layers

of planks, they cannot dissipate significant amount of

energy during an earthquake. The mechanical con-

nections between adjacent panels and with the foun-

dations hence become the most important component

affecting the static and cyclic behaviour of CLT

buildings, as most of the building flexibility is

concentrated in the connections. In addition, mechan-

ical connections in CLT buildings play an important

role in maintaining the integrity of buildings and

providing the necessary strength, stiffness, stability

and ductility [17].

Metal angle brackets, hold-downs, plates and straps

are used to transfer forces from walls to floors, from

one level to another level, and to foundations. Hold-

downs are mainly used in the corners of wall segments

and close to door opening, to resist overturning forces

that result from an earthquake or wind. On the other

hand, the main role of L-shaped metal brackets is to

resist shear forces in wall panels caused by wind or by

a seismic event. Nails with specific surface features

such as grooves or helically threaded nails are mostly

used with perforated metal plates and brackets and

installed on the surface of the panel to connect the

metal hardware to the CLT panel.

The different orientations of adjacent layers in CLT

panels, their built-up nature, the gaps between the

longitudinal boards, the edge gluing, if any, and the

possible presence of grooves sawn into the boards to

relieve drying stresses, make the determination of the

fastening capacity in CLT relatively difficult to predict

compared to traditional sawn timber or engineered

wood products. For a reliable seismic analysis,

therefore, it is crucial to investigate the cyclic

behaviour of the connections between CLT elements

and with the foundations. Experimental test results are

required to feed advanced numerical models [23], for

the development of design methods, and for the

derivation of analytical calculation models of input

parameters such as strength and stiffness of connec-

tions, which are required in seismic analysis of CLT

structures [16].

So far, no relevant design codes for CLT construc-

tion were published in Europe. In recent years, several

research projects in Europe and in North America have

been launched, with the aim to better understand the

potential of CLT technology as a seismic resistant

construction system. A comprehensive investigation

was undertaken at University of Ljubljana, Slovenia,

to determine the seismic behaviour of CLT connec-

tions and wall panels [6]. An experimental research

programme on CLT metal connectors was also carried

out at the University of Trento, Italy [1, 26], FPInno-

vations, Canada [25] and Technical University of

Graz, Austria [15]. Experimental mechanical proper-

ties of CLT metal connectors under cyclic and

monotonic loading were determined and failure modes

were presented.

Within the SOFIE project, carried out by CNR-

IVALSA Trees and Timber Research Institute (San

Michele all’ Adige, Italy), seismic behaviour of multi-

storey CLT buildings was studied. An extensive

experimental programme was conducted, including

racking tests of wall panels and sub-assemblies [3],

pseudo-dynamic tests on a one-storey building, and a

series of shaking table tests on three- and seven-storey

CLT buildings [2, 4]. The tests provided excellent

outcomes, as the buildings were able to survive strong

recorded earthquakes, with minimal structural dam-

age, while at the same time demonstrating significant

energy dissipation.

Further research was still needed in order to better

define the seismic behaviour of typical CLT connec-

tions, as they govern the behaviour of CLT wall

systems and of entire CLT buildings. Thus, in 2010 an

extensive experimental programme aimed to fully

characterize the typical CLT connections initiated as a

research collaboration between CNR-IVALSA and

the University of Trieste, Italy [18–20]. Twenty

different configurations of CLT connections used in

the SOFIE buildings were tested under monotonic and

cyclic loading.

In this paper, the results of hold-down and angle

bracket connections are presented and critically dis-

cussed. Important quantities for seismic design such as

strength, stiffness, energy dissipation, impairment of

strength, and ductility of the connections are evalu-

ated. A comparison between experimental test results,

the analytical calculation method developed by Uibel

and Blaß [28, 29], and the analytical formulas of

Eurocode 5 [9] is also provided. Analytical models for

the determination of stiffness of CLT metal connectors

are also presented and compared with the experimen-

tal results. In addition, the overstrength factors are

calculated, which are needed in capacity-based


seismic design and are currently not provided by codes

of practice such as Eurocode 8 [10]. Lastly, some

suggestions to improve the behaviour of the CLT

metal connectors are given.

2 Experimental programme

2.1 Test configurations and setups

A reference point for the experimental programme on

typical CLT connections was the three-storey SOFIE

building, which was tested on shaking table in

Tsukuba, Japan, in 2006 [2]. All wall elements of

the building were 5-layered panels made of spruce and

with 85 mm thickness. The panel build-up was 17-17-

17-17-17 (in mm), where the numbers in bold

represent the thickness of layers with orientation of

grain perpendicular to the layers with thickness

provided by numbers in normal font. The floor panels

were also made of 5 orthogonally crossed layers with

doubled layer in the middle: 27-17-27-27-17-27 (in

mm). The total thickness of the floor panels was

142 mm. CLT wall panels and floor panels were

connected to each other with self-tapping screws,

while the wall elements were anchored to CLT floors

and foundations with metal connectors (hold-downs

and brackets). The building was built on a steel base

foundation as shown in Fig. 1.

Based on the connections that were used in the

reference building two main CLT connection test

groups were determined: (i) metal connectors (hold-

downs and angle brackets) with eight different test

configurations (Test #1–Test #8); and (ii) screwed

connections with twelve test configurations (Test #9–

Test #20). The types of connections and CLT panels

used in the tests were equal or similar to the ones used

in the reference building. The number of fasteners

(nails) used to connect the metal connectors to the

CLT panels and the fastener spacing was in accor-

dance with the reference building as well. The

reasoning behind this choice is the need to fully

characterise the cyclic behaviour of each connection in

the SOFIE CLT building, with the aim to provide input

data to advanced numerical model and design values

to practicing engineers. In this paper experimental and

analytical study of CLT metal connectors (hold-downs

and angle brackets) is presented, while more details on

experimental study of CLT screwed connections can

be found in Gavric et al. [20] and Gavric [18].

Two different types of hold-down and angle bracket

connections were tested under monotonic and cyclic

loading: CLT panel-steel base foundation and panel–

panel connections, representative of wall panels to

foundations and wall-to-floor panels connections in

the upper stories of CLT buildings, respectively. Both

Fig. 1 Three-storey SOFIE building

Fig. 2 Elevation (right) and cross-section (left) of test setup

used for wall-foundation hold-down connection loaded in

tension (Test configuration #1)


hold-down and angle bracket tests were performed

separately in two directions: shear and tension. The

test setups of hold-down tests in tension direction are

displayed in Fig. 2 (Test configuration #1—wall

panel-foundation) and Fig. 3 (Test configuration

#2—wall panel-floor panel).

Similarly, there were two cases of shear tests of

hold-downs: wall panel-foundation (Test configura-

tion #3, see Fig. 4) and wall panel-floor panel (Test

configuration #4).

The test setups for angle brackets configurations #5

(panel–foundation tension tests), #6 (panel–panel

tension tests), #7 (panel–foundation shear tests) and

#8 (panel–panel shear tests) were fairly similar to the

ones for the corresponding hold-downs tests, respec-

tively. The only differences lay in the choice of an

adequate wall panel size [18]. In Table 1 information

of test configurations with the number of cyclic tests

performed and number of repeated cycles at each

displacement rate are presented. In addition, maxi-

mum resistances (Fmax) obtained in the tests are

presented along with the maximum resistances after

the 3rd loading cycles (Fmax(3rd)). Test results will be

discussed in Sect. 3.

The hold-downs used in Test configurations #1 and

#3 were type WHT540 with twelve 4 9 60 mm Anker

annular ring nails. This relatively new type of hold-

down matches well the geometry and thickness of the

older type HTT22 of hold-down, originally used at the

ground storey of the three-storey SOFIE building. The

hold-downs used in test configurations #2 and #4 were

type WHT440 with nine nails (same type of nails as for

tests #1 and #3). This hold-down has the same

geometry as the hold-down HTT16, which was

originally used in the upper stories of the three-storey

SOFIE building. The only differences between hold-

downs WHT540 and WHT440 are the height of the

connector, and consequently the number of holes

available for fastening the hold-down to the wall

panel. Both types of hold-downs were anchored to a

steel base with a 16 mm diameter bolt.

The type of angle brackets used in Test configura-

tions #5 and #7 was BMF 90 9 116 9 48 9 3 mm

with eleven 4 9 60 mm Anker annular ring nails.

Each bracket was anchored to foundations with one

U12 bolt, the same as the brackets used at the ground

level of the three-storey SOFIE building. In Test

configurations #6 and #8, 100 9 100 9 90 9 3 mm

BMF angle brackets were used, representing the wall-

floor connection of upper stories in CLT buildings.

The brackets were fastened to the wall panel with eight

4 9 60 mm annular ring nails and to the floor with six

4 9 60 mm annular ring nails with two additional

HBS screws 4 9 60 mm.

2.2 Loading protocol

Eurocode 8 [10] states that the properties of dissipative

zones of timber structures should be determined by


used for wall-floor hold-down connection loaded in tension

(Test configuration #2)


used for wall-foundation hold-down connection loaded in shear



tests performed either on single joints, on whole

structures or on parts thereof in accordance with EN

12512 [11]. The standard procedure for cyclic testing

of joints made with mechanical fasteners prescribed

by EN 12512 was followed in all of the tests, with

input displacement rate varying from 0.2 to 0.8 mm/s

so that the duration of each test did not exceed the time

limit of 30 min. Monotonic tests were carried out by

displacement controlled ramp loading at a rate varying

from 0.05 to 0.2 mm/s. Specimens were stored and

tested under controlled environmental conditions with

50 % relative humidity and 20 �C temperature. After

each test, the moisture content and density of the CLT

specimens were measured.

All cyclic shear tests were conducted using a

reversed cyclic loading procedure with predefined

yield values, which were varied from configuration to

configuration, depending on experimental yield values

obtained from the monotonic tests. However, all

tension tests were subjected to a non-reversed modi-

fication of the procedure outlined in EN 12512 due to

restrained movement (timber–timber or timber-foun-

dation contact) in compression. Basically the displace-

ment was cycled from zero to a positive (tensile) value

without the excursions in the negative (compression)

values. For each of the eight different configurations, at

least one monotonic and six cyclic tests were per-

formed in order to obtain statistically representative

values.

3 Experimental test results and discussion

The values of mechanical properties for a single

specimen were analyzed by taking into account the

results from both sides of the hysteretic loops, except

for the cases where loading in tension only was carried

out. The mean values for a specific connection

configuration were obtained by calculating the average

value of all six cyclic test results within the specific

configuration. Based on a statistical analysis of the

experimental results, the coefficients of variation

(COV) were calculated. All graphs, results and anal-

yses of connection tests presented in this paper refer to

one connector (one hold-down or one angle bracket).

3.1 Cyclic tests of hold-downs

All cyclic tests of hold-downs in axial (tension)

direction revealed failure of nails due to vertical

displacement of the panel. After the end of the tests, the

steel part of hold-down was virtually undamaged,

although significant bending deformation could be

observed even in the elastic range. Failures occurred

due to pullout of the nails, either by tearing of the nail

cap or by a combination of nail withdrawal and

bending. Figure 5 shows the typical hysteresis loops

for hold-downs loaded in tension (Test configuration

#1).

In the shear direction, mainly deformation of the

steel parts of the hold-down was observed, leading to

failure of the steel due to local buckling or to a

combination of bending of steel and withdrawal of

nails (Fig. 6). Both failure mechanisms occurred at

very high (more than 100 mm) displacement levels. At

the ultimate slip value of 30 mm, according to EN

12512, the connections were still in the second linear

branch of the backbone curve (Fig. 7).

In the case of wall-floor hold-down connections,

compression deformation on the lower part of the floor

Table 1 Metal connectors test configurations and test information

Conf.

no.

Metal connector Nails

no.

Connection type Loading

direction

Cyclic

test no.

No. of

cycles

Fmax

(kN)

Fmax(3rd)

(kN)

1 Hold-down WHT540 12 CLT–foundation Tension 6 3 48.33 41.33

2 Hold-down WHT440 9 CLT–CLT Tension 6 3 36.21 31.27

3 Hold-down WHT540 12 CLT–foundation Shear 6 3 9.98 9.17

4 Hold-down WHT440 9 CLT–CLT Shear 6 3 7.88 6.85

5 Angle bracket BMF 90 9 116 9 48 9 3 11 CLT–foundation Tension 6 3 23.47 18.59

6 Angle bracket BMF 100 9 100 9 90 9 3 8 CLT–CLT Tension 6 3 12.57 10.96

7 Angle bracket BMF 90 9 116 9 48 9 3 11 CLT–foundation Shear 6 3 26.85 18.80

8 Angle bracket BMF 100 9 100 9 90 9 3 8 CLT–CLT Shear 6 3 19.91 14.26


panel was negligible during the loading in tension

direction (Test configuration #2). The failure mecha-

nisms of all the specimens in this test configuration

were the same as the ones in Test configuration #1. It

can be concluded that when twelve 4 9 60 mm Anker

annular ring nails or less are used in these types of

hold-downs, the failure mechanism in tension direc-

tion will be due to the yielding of nails, which is a

ductile failure mechanism. When wall-floor hold-

down connections were tested in shear direction (Test

configuration #4), the behaviour was very similar to

the configuration #2. Basically all shear deformation

was a result of the buckling of the hold-down, with

some minor nails extraction at higher displacement

levels. It can be concluded that in shear direction, nails

in hold-downs do not develop their full strength

capacity, as the failure occurs due to the plasticization

of steel part of the hold-down.

Table 2 lists the strength and deformation proper-

ties of hold-downs loaded in tension and in shear

according to EN 12512 [11]. Both mean value (xmean)

and COV are displayed for each mechanical property.

More specifically, kel and kpl represent the initial and

plastic stiffnesses; Fy and vy signify the yielding load

and displacement; Fmax and vmax denote the maximum

load and displacement; Fmax(3rd) is the maximum load

at the 3rd loading cycle; Fu and vu signify the ultimate

load and displacement; D signifies the ductility ratio

(ratio of the ultimate displacement to the yield

displacement). As one of the criteria in EN 12512 to

determine the ultimate value is also the strength at

30 mm displacement, the corresponding force F30 and

ductility ratio D30 are presented for the relevant tests.

The average strength degradation between the 1st and

the 3rd maximum load cycle is denoted with DF1–3.

The energy dissipation properties are measured by the

quantities meq(1st) and meq(3rd), which represent the

equivalent viscous damping ratios calculated at the 1st

and 3rd cycles of the hysteretic loop at the displace-

ment value where the maximum load Fmax was

reached (Eq. 1). The equivalent viscous damping ratio

meq,i expresses the hysteresis damping properties of the

connection and is defined as the ratio of the dissipated

energy in the cycle i (Ed,i) to the available potential

energy in that cycle (Ep,i) multiplied by 4p [5].

meq;i ¼Ed;i

4pEp;i: ð1Þ

Fig. 5 Typical hysteresis loops and monotonic curve for Test

configuration #1

Fig. 6 Wall-foundation hold-down connection loaded in shear



configuration #3


The available potential energy Ep,i can be deter-

mined as Ep,i = �Fi�ui, where Fi and ui are the

maximum force and maximum displacement attained

in cycle i, respectively [5]. In addition, the 5th and the

95th percentiles of experimental strength capacities

(F0.05 and F0.95, respectively), and the overstrength

factors cRd were derived (see Sect. 3.3 for details).

The pull-out (tensile) strength was almost five times

greater than the shear strength in both cases (wall-

foundation and wall-floor hold-down connections),

while the ductility behaviour was significantly better

in shear as was the dissipated energy, resulting in an

equivalent viscous damping meq ranging from 19.8 to

21.4 % for the 1st cycles and 14.6 to 16.1 % for the

3rd cycles, as opposed to 8.1–8.5 % for the 1st cycles

and 2.8–3.6 % for the 3rd cycles in the axial direction.

This difference is mostly due to bending of the steel

part of the hold-down under shear loading. The

difference between the maximum load at the 1st and

3rd loading cycles in Test #1 and Test #2 was less than

16 %. The yielding force Fy and the maximum force

Fmax in the case of Test #1 was on average 1.3 times

higher than in Test #2. The number of nails used in

Test #1 was twelve while in the case of Test #2 only

nine nails were used, which gives a ratio of 1.3. This

indicates a very similar behaviour of hold-downs when

used for wall–foundation and for wall–floor connec-

tions. It must be pointed out, however, that cases of

brittle failures of hold-downs loaded in tension were

observed in the past [3]. In that case, failure of the net

cross-section of the metal part of hold-down occurred,

as the load-carrying capacity of this part was lower

than the strength of the nailed connection to the CLT

wall panel. This information is relevant for the design

procedure of metal connectors that is discussed later

on in this paper.

3.2 Cyclic tests of angle brackets

The main load-carrying direction for the angle brack-

ets is shear, however the tests showed also a relatively

good behaviour in axial direction (tension). In case of

wall-foundation connection, failures in tension

occurred in the steel base around the anchor bolt

(Fig. 8) which was pulled through the angle bracket.

That resulted in a sudden significant drop of resistance

Table 2 Mechanical properties of hold-down connections according to EN 12512 [11] with overstrength factors (cRd)

Mechanical property Test configuration no.

1 2 3 4

xmean COV xmean COV xmean COV xmean COV

kel (kN/mm) 4.51 14.31 2.65 19.27 3.40 33.25 1.56 16.20

kpl (kN/mm) 0.75 14.24 0.44 19.13 0.28 7.27 0.21 5.43

Fy (kN) 40.46 8.11 32.21 4.52 3.61 35.59 2.72 28.13

vy (mm) 8.81 21.76 11.91 19.83 1.13 42.30 1.71 17.23

Fmax (kN) 48.33 5.37 36.21 5.45 – – – –

Fmax(3rd) (kN) 41.33 7.53 31.27 7.68 – – – –

vmax (mm) 20.30 14.17 21.52 11.00 – – – –

Fu (kN) 38.79 5.31 30.22 12.80 – – – –

vu (mm) 23.75 13.82 22.99 9.51 – – – –

D [–] 2.76 16.21 1.97 13.75 – – – –

F30 (kN) – – – – 9.98 7.03 7.88 4.78

D30 (–) – – – – 31.26 43.50 18.73 20.00

DF1-3 (%) 15.90 34.59 66.30 9.05 12.45 25.48 10.98 20.18

meq(1st) (%) 8.50 5.70 8.11 9.98 19.84 13.91 21.38 6.93

meq(3rd) (%) 2.78 17.29 3.60 24.40 14.63 17.79 16.09 14.86

F0.05 (kN) 42.40 31.59 8.48 7.04

F0.95 (kN) 54.95 41.40 11.70 8.80

cRd (–) 1.30 1.31 1.38 1.25


(Fig. 9). The upper part of the angle bracket, which

was nailed to the wooden panel, was hardly deformed.

This brittle type of failure should be avoided in

dissipative CLT connections by providing capacity

based design rules [21].

A different cyclic behaviour was observed under

shear loading. The failure mode was yielding of the

nails with formation of two plastic hinges (mode ‘e’

according to Eurocode 5 [9] Part 1), as shown in

Fig. 10. A slight ovalisation of the brackets around the

bolt resulted in a local drop of strength in each cycle

(Fig. 11), which however did not affect significantly

the overall connection behaviour. Figure 11 shows the

typical hysteresis loops for angle brackets loaded in

shear (Test configuration #7).

The behaviour in tension and in shear direction of

angle brackets connecting wall and floor panels (Test

#6 and Test #8) was similar to the behaviour of Tests

#5 and #7. In tension, a brittle type of failure was

observed, as the nails and screws which were

connected to the floor panel were extracted before

the nails, which were fastened to the wall panel, started

to deform (Fig. 12).

This means that the full dissipative capacity of the

bracket was not achieved, as the ductility and energy

dissipated through nail withdrawal is significantly

lower than it would be in the case of a ductile shear

deformation with plastic hinge formation of nails

connecting the angle bracket to the wall panel. In shear

direction (Test #8), most of the deformation resulted

from yielding of the nails in both wall and floor panel.

Here, only the deformation of nails in the wall panel is

desirable, whereas the connection to the floor panel

should remain elastic. In both cases the importance of

a proper capacity-based design of CLT connections is

shown again.

In terms of strength, angle brackets in tension

direction exhibited a relatively high load-carrying

capacity considering that their primary load-carrying

direction is shear. In the case of wall-foundation

connection, the average uplift resistance of an angle

bracket was 87 % of its shear resistance (Table 3).

This percentage could have been even higher if the

brittle failure in tension due to pull-through of the bolt

had been avoided, for example by using more than one

bolt. In case of wall-panel angle bracket connection,

the axial resistance was 63 % of the shear resistance

(Table 3). Again, this percentage could have been

higher if the brittle failure mode of nail withdrawals

had been avoided, for example using self-drilling

screws instead of slender nails. The difference

between the maximum load at the 1st and 3rd loading

cycles in Test #5 and Test #6 was less than 20 % while

it was less than 30 % in Test #7 and Test #8.

The deformation of nails and steel bracket in shear

resulted in an equivalent viscous damping

meq = 11–14 % for the 3rd cycles. In tension the

equivalent viscous damping was on average 7.1 % for

wall-foundation connections (due to steel deforma-

tion), whereas in the case of wall-floor connection meq

reached only 1.7 % for the 3rd cycles.

It can be concluded that angle brackets contribute

quite significantly to the strength and stiffness of the

joint not only with their shear resistance but also with

their uplift resistance, which should therefore not be

neglected in CLT building design.

Fig. 8 Wall-foundation angle bracket connection loaded in

tension (Test configuration #5)


configuration #5


3.3 Overstrength factors

Brittle members in timber structures must be designed

for the overstrength related to the strength of the

ductile connections to ensure the ductile failure

mechanism will take place before the failure of the

brittle members [22]. Dissipative zones shall be

located in joints and connections, whereas the timber

members themselves shall be regarded as behaving

elastically. Thus, overstrength values of typical CLT

connection have been derived on the basis of statistical

analysis of the experimental test results presented in

this paper.

The ovestrength ratio cRd is defined as the ratio of

the 95th percentile of the connection strength distri-

bution, F0.95, to the design connection strength, Fd

[22]:

cRd ¼F0:95

Fd

: ð2Þ

Based on the statistical analysis of the six cyclic

tests on each of the eight CLT metal connectors

configurations presented earlier in this section, the

design strength capacity Fd was calculated by dividing

the characteristic experimental strength F0.05 by the

strength partial factor cM, assumed equal to one

according to Eurocode 8 [10] for dissipative timber

structures. The experimental characteristic strength

values from tests F0.05 were based on the lower 5th

percentile values assuming the log-normal distribution

with a 75 % confidence level [12], following the

standard procedure for calculation of characteristic

5-percentile values from the EN 14358 standard [12].

The upper 95th percentile values were calculated

using the same standard. The resulting overstrength

factors cRd of CLT metal connectors are presented in

Tables 2 and 3.

The overstrength factors for hold-downs loaded in

tension are on average 1.3, while in shear direction,

hold-downs were found to have overstrength ratios of

1.25 and 1.38, depending on the type of configuration.

For steel angle brackets connecting foundations to

Fig. 10 Wall-foundation angle bracket connection loaded in shear (Test configuration #7)


configuration #7


CLT wall panel (Test #5 and Test #7), the overstrength

factors range from 1.16 to 1.23, depending on the

direction of loading. These results are consistent with

the values of 1.26 in shear and 1.18 in axial direction

proposed by Fragiacomo et al. [17]. Flatscher and

Schickhofer [15] also found overstrength values of

angle brackets below the value of 1.3 in both

directions, shear and uplift. Angle brackets connecting

CLT wall to CLT floor (Test #6 and Test #8) were

found to have slightly higher overstrength ratios,

namely 1.44 in uplift direction and 1.40 in shear

direction due to larger scatter of the experimental

results.

It should be pointed out that these values of the

overstrength factors can be used only for connections

which were experimentally tested and for which the

characteristic value of the strength is provided by the

producers. For connections that were not experimen-

tally tested, higher values of the overstrength factors

shall be used to allow for the difference between the

Fig. 12 Wall-floor angle bracket connection loaded in tension (Test configuration #6, left) and in shear (Test configuration #8, right)

Table 3 Mechanical properties of angle bracket connections according to EN 12512 [11] with overstrength factors (cRd)

Mechanical property Test configuration no.

5 6 7 8

xmean COV xmean COV xmean COV xmean COV

kel (kN/mm) 2.53 9.72 2.98 22.05 2.09 16.41 1.10 12.34

kpl (kN/mm) 0.42 10.11 0.50 22.01 0.35 16.56 0.18 11.95

Fy (kN) 19.22 2.73 11.12 9.69 22.98 5.19 16.61 7.46

vy (mm) 7.26 9.04 3.97 28.12 11.74 5.87 13.73 7.27

Fmax (kN) 23.47 4.32 12.57 7.71 26.85 3.15 19.91 6.95

Fmax(3rd) (kN) 18.59 9.72 10.96 7.80 18.80 10.25 14.26 9.75

vmax (mm) 17.69 9.62 7.10 10.62 28.51 14.75 29.09 7.01

Fu (kN) 18.74 4.32 10.06 7.68 21.48 3.15 15.86 7.59

vu (mm) 23.19 6.14 20.01 46.39 31.86 0.33 52.26 2.21

D (–) 3.21 6.86 5.40 54.20 2.63 6.03 3.97 10.81

DF1–3 (%) 22.85 32.43 9.26 9.32 32.59 9.31 28.20 6.79

meq(1st) (%) 12.33 5.43 7.40 11.71 22.75 5.93 17.49 5.59

meq(3rd) (%) 7.07 19.17 1.74 10.90 14.01 6.84 11.17 14.15

F0.05 (kN) 21.16 10.45 24.89 16.80

F0.95 (kN) 26.00 15.05 28.93 23.49

cRd (–) 1.23 1.44 1.16 1.40


predictions using the analytical formulas and the

actual experimental values.

4 Analytical models and comparisons

with experimental results

Comparisons among the experimental values obtained

from CLT connection tests (Sect. 3), the characteristic

values calculated following Eurocode 5 [9] provisions,

and the characteristic values determined using the

embedment formulas proposed by Uibel and Blaß [28]

are presented in this section.

4.1 Strength capacity of typical CLT metal

connectors

A typical connector (hold-down, angle bracket) used

in CLT construction consists of a metal part connected

to two CLT panels (or a CLT panel and the foundation)

using fasteners such as nails and screws. The metal

part is usually a linear member and, as such, cannot

transfer significant bending moment. CLT connectors

therefore transfer axial and shear forces between the

two connected panels.

The shear strength capacity FRd,x and axial strength

capacity FRd,y of a CLT metal connector is the minimum

among: (i) the shear and axial capacities of the upper

connection to the CLT panel FRd,upper,x and FRd,upper,y;

(ii) the metal part of the connection, FRd,metal,x and

FRd,metal,y; and (iii) the lower connection to the CLT

panel or foundation FRd,lower,x and FRd,lower,y:

FRd;x ¼ min FRd;upper;x; FRd;metal;x; FRd;lower;x

� �; ð3Þ

FRd;y ¼ min FRd;upper;y; FRd;metal;y; FRd;lower;y

� �: ð4Þ

For hold-downs and angle brackets loaded in shear

direction several failure modes are possible:

• shear failure of nails, attached to the CLT wall

panel (FRd,upper,x)

• group tear out of the fasteners (FRd,upper,x)

• net section failure of the steel part of the angle

bracket (FRd,metal,x)

• buckling of the steel part of the hold-down

between nails and bolt (FRd,metal,x)

• shear failure of nails, attached to the CLT floor

panel (FRd,lower,x)

• ovalisation of the hole of a bolt in the angle bracket

due to embedment of the bolt in steel (FRd,lower,x)

• tear out or block shear of the steel part of the angle

bracket (FRd,lower,x)

In the case of axial loading of metal connectors

there are also several possible failure mechanisms:

• shear failure of nails, attached to the CLT wall

panel (FRd,upper,y)

• block shear or tension failure of CLT wall panel

(FRd,upper,y)

• fracture in tension of the metal connectors in the

net section of the steel part (FRd,metal,y)

• yielding of the steel part of the connector around

the anchoring bolt (with possible pull-through)

(FRd,lower,y)

• tension failure of the anchoring bolt (FRd,lower,y)

• withdrawal of the anchoring bolt (FRd,lower,y)

• withdrawal of nails or screws from the floor panel

(FRd,lower,y)

In the cases where the shear capacity of the nailed

connection of the hold-downs with the CLT wall

panel does not exceed the strength capacities of the

steel net section, of the anchor bolt and of the other

possible failure modes discussed before, the total load

carrying capacity of the connection is the sum of the

resistances of all fasteners connecting the metal

connector to the CLT wall. Otherwise, some of the

aforementioned failure modes, which are undesirable

in the seismic design of timber buildings being brittle,

may occur. To achieve the desired ductile behaviour

of metal connectors loaded in axial and shear

directions, namely the plasticization of the nailed

connection with the CLT wall panel, capacity-based

design principles have to be followed [21].

To predict the shear and axial strength capacity of

CLT connectors is therefore necessary to derive

formulas for the resistance of fasteners such as screws,

nails and other dowel-type fasteners in CLT, taking

into account the nature of lamination, lay-up, wood

species, edge-gluing, and possible presence of grooves

sawn into the boards to relieve drying stresses.

Eurocode 5 formulas for the prediction of the shear

and withdrawal capacities of timber connections were

derived for sawn timber and engineered wood pro-

ducts different from CLT, and therefore their validity

for CLT must be checked.


Extensive research for the determination of embed-

ment strength capacity and withdrawal strength

capacity of dowel-type fasteners was conducted by

Uibel and Blaß [28, 29]. The proposed design

procedures deal only with ductile failure modes to

determine the lateral load resistance of such connec-

tions. Empirical models were developed based on test

results to establish the strength capacity of dowel-type

fasteners under lateral loading. For dowel-type

fasteners driven perpendicular to the plane of the

CLT panel loaded in shear Uibel and Blaß derived

formulas for the characteristic embedment strength fh,k

(in N/mm2) of 4 mm diameter nails and screws up to

12 mm diameter in CLT panels with laminations less

than 7 mm thick. The formula is expressed as a

function of the characteristic density of the wood qk

(in kg/m3) and the diameter of the fastener d (in mm),

as shown in Eq. (5):

fh;k ¼ 0:112d�0:5q1:05k ðN=mm2Þ: ð5Þ

The proposed characteristic embedment strength

equation is independent of the loading direction with

respect to the grain orientation of the layers. Once the

characteristic embedment strength fh,k is calculated, it

can be used in Johansen equations for load carrying

capacity of fasteners in [9]. It should be noted that the

common thickness of currently used laminations can

be up to 40 mm, which is greater than the lamination

thickness of the CLT tested by Uibel and Blass. The

validity of Eq. (5) must therefore be checked against

experimental tests carried out on CLT panels with

different lamination thickness. Verifications of char-

acteristic strength capacities of steel parts of the

connectors will be performed using the Eurocode 3

(EC3) formulas [8].

4.2 Stiffness of typical CLT metal connectors

In seismic design of timber structures, information on

stiffness of connections is crucial for predicting the

properties of the structures in terms of dynamic

behaviour. Numerous experimental, analytical and

numerical studies [3, 6, 16–20] concluded that

connections in CLT construction strongly influence

the deformability properties of CLT buildings. In the

existing European timber design codes [9] no provi-

sions for prediction of CLT connections’ elastic

stiffness are currently available. Thus, there is an

obvious need for analytical formulations of CLT

connectors.

The elastic stiffness of CLT metal connectors (kel,i)

in shear and axial directions (kel,x and kel,y, respec-

tively) can be calculated by taking into account

various deformability components: (i) the elastic

stiffness of nails by summing the elastic stiffness of

all n nails in the connector (n 9 kn) using EC5 (2004)

empirical formulae; (ii) the elastic stiffness of steel

plate of the connector ks; and (iii) the elastic stiffness

of the connection with the foundation or floor panel ka

(for example, the bolted or nailed connection). In the

case of steel-timber connections loaded in axial

direction, the deformability of the CLT panel perpen-

dicular to grain also has to be taken into account. The

elastic stiffness of CLT metal connectors in shear and

axial directions can be calculated by modelling the

connector as series of springs:

kel;i ¼1

n� kn

þ 1

ks

þ 1

ka

� ��1

: ð6Þ

4.3 Experimental-analytical comparison

Material and geometrical properties of the hold-downs

and angle brackets used in the tests were obtained from

European Technical Approval documents (ETA-11/

086 [14] for hold-downs, and ETA-09/323 [13] for

angle brackets). The properties of nails, used to

connect the metal connectors to CLT panels, were

obtained from the producer’s catalogue [24]. The

characteristic and mean wood densities were deter-

mined from 230 CLT specimens used in the experi-

mental tests following EN 384 [7]. A minimum 5th

percentile density q0.05 = 364 kg/m3 and mean den-

sity qm = 450 kg/m3 were obtained [18].

In shear direction, the strength and stiffness of hold-

downs is relatively weak (see Sect. 3.1), thus their

contribution to the strength and stiffness of CLT wall

systems and buildings is minor. For that reason,

establishing load-carrying capacity design methods

for hold-downs in shear direction is irrelevant. On the

other hand, the contribution of angle brackets to the

lateral resistance of CLT wall systems in their weaker

axial direction is significant (Sect. 3.2), thus their

analytical characteristic axial strength will be deter-

mined. Stiffness values will be calculated for hold-

downs in tension and angle brackets in shear as in the


weaker directions undesired failure modes occurred

(local buckling in hold-downs loaded in shear and

brittle nail withdrawal failures in angle brackets

loaded in tension).

In terms of strength capacities, the analytical

characteristic values of CLT metal connectors using

the formulas in EC5 and by Uibel and Blaß [28] will be

compared to the experimental 5th percentile strength

values obtained in the experimental tests. On the other

hand, the mean experimental elastic stiffness values

will be compared to the analytical stiffness predictions

at serviceability limit state, see EC5 (2004).

Tests #1 and #2 (hold-downs loaded in tension) had

a typical failure due to yielding of nails (failure mode

‘‘c’’ according to EC5). The estimated load-carrying

capacity is quite conservative in comparison with the

experimental 5th percentile values, namely 2.6 times

lower using the EC5 equations and 2.2 times lower

using Uibel and Blaß’ procedure (Table 4). Tests #5

and #6 (angle brackets loaded in tension) failed in a

brittle way. In Test #5, all test specimens failed due to

pull-through of the bolt (bolt punching), while all

specimens in Test #6 failed due to withdrawal of nails

anchoring the angle brackets to the floor panel. The

analytical predictions of failure modes correspond

well with the experimental failure modes (Table 4),

but the predicted values are very conservative

(Fig. 13). On the other hand, failures in the brackets

in shear direction (Test #7 and Test #8) occurred in the

nailed connection, which is considered as a ductile

failure mode. The calculated strengths for tests #5 to

#8 are also conservative (F0.05,exp/FRk ratios range

from 1.3 to 2.8). Uibel and Blaß’ embedment formulas

are more accurate than Eurocode 5 [9] formulas,

leading to about 18 % increase in strength.

Analytical predictions of the elastic stiffness of

CLT metal connectors indicate that most of the

deformation is due to nails deformations, namely

72–76 % in the case of hold-downs in tension and

90 % and more in the case of angle brackets loaded in

shear (Table 5). Note that the total deformation of

nails in the case of Test #8 is considered as the sum of

the deformation of the upper (bracket-CLT wall) and

lower (bracket-CLT floor) parts altogether. The stiff-

ness in the case of Test #2 (CLT wall-CLT floor) is

significantly lower than in Test #1 (CLT wall-steel

foundation). The contributions of individual stiffness

components to the total stiffness of the connection are

inversely proportional to their stiffness values (see

Eq. 6). As opposed to the analytical strength predic-

tions, which proved to be conservative in all cases, the

analytical predictions of elastic stiffness of CLT metal

connectors are in all cases significantly overestimated.

Table 4 Analytical characteristic strength capacities of CLT metal connectors with predicted failure modes compared to the

experimental 5th percentile values (in brackets: values obtained using the Uibel and Blaß’ formulas)

Characteristic strength

capacity

Test #1 Test #2 Test #5 Test #6 Test #7 Test #8

FRk,metal (kN) 59.5 59.5 81.1 70.5 34.6 32.6

FRk,upper (kN) 16.2 (19.1) 12.2 (14.3) 14.9 (17.5) 10.8 (12.7) 14.9 (17.5) 10.8 (12.7)

FRk,lower (kN) 72.4 72.4 7.6 5.8 19.8 10.8 (12.7)

F0.05,exp (kN) 42.4 31.6 21.2 10.5 24.9 16.8

F0.05,exp/FRk 2.6 (2.2) 2.6 (2.2) 2.8 1.8 1.7 (1.4) 1.6 (1.3)

Failure mode Nails in shear Nails in shear Bolt punching Nails

withdrawal

Nails in shear Nails in shear

Fig. 13 Experimental results of tension and shear tests on CLT

metal connectors compared to the analytical predictions of their

shear and axial characteristic strength capacities


The ratios of the analytical predictions to the exper-

imental values (kel,i/km,exp) range from 3.9 to 9.1

(Table 5).

This means that a very high percentage of the

connection total deformation comes from different

components. The eccentricity between the bolted

connection to the foundation and the connection to

the wall panel is a significant source of additional

flexibility, as the hold-down is subjected to bending in

the metal plate between the bottom of the nailed

connection and the top of the stiffening plates

(Fig. 14) where the resisting cross-section is very

weak, and local buckling occurs even during the

loading in the elastic range. In addition, local defor-

mations of the steel part between the anchorage bolt

and the vertical flange may also contribute to the total

deformation (Fig. 14 middle).

In the case of angle brackets loaded in shear,

flexural deformations and rotational displacements of

the metal part due to eccentric loads in two directions

are quite significant [27]. Thus, taking into account

only nails’ deformations, axial deformation of the steel

Table 5 Analytical elastic stiffness of CLT metal connectors compared to the mean experimental stiffness values

Stiffness capacity Test #1 Test #2 Test #7 Test #8

(kN/mm) (% kel,i) (kN/mm) (% kel,i) (kN/mm) (% kel,i) (kN/mm) (% kel,i)

Nails: n 9 kn 23.2 76.1 17.4 72.1 21.2 89.5 15.4 49.3

Steel plate: ks 81.0 21.8 87.2 14.4 228.1 8.3 560.0 1.4

Anchorage: ka 844.5 2.1 92.8 13.5 844.6 2.2 15.4 49.3

Total stiffness: kel,i 17.6 100 12.5 100 19.0 100 7.6 100

Experimental: km,exp 4.5 – 2.65 – 2.09 – 1.10 –

Ratio kel,i/km,exp: 3.9 4.7 9.1 6.9

Fig. 14 Examples of local bending in the metal plate of hold-downs loaded in tension


plate, and anchorage deformation is not enough for a

conservative estimation of the elastic stiffness.

Strength and stiffness predictions of CLT metal

connectors correspond well with the conclusions of

Acler et al. [1] and Tomasi and Sartori [27]. In their

experimental-analytical research on metal connectors,

they found that the use of the EC5 formulas for the

strength capacity and stiffness of nailed CLT wall-

hold-down and CLT wall-angle bracket connections

leads to conservative results for the strength and non-

conservative results for the stiffness values. In

conclusion, experimental values of hold-down and

angle brackets should be used in seismic analyses.

4.4 Provisions to improve the CLT connector

behaviour

Non-optimal stiffness performance of hold-downs and

angle brackets suggests that some adjustments should

be made in the design of CLT metal connectors. In

addition, if capacity based design principles had been

taken into account [21], ductile failure modes could

have been attained in all cases. Some proposals for a

better mechanical performance of metal connectors

(hold-downs and angle brackets) are listed herein:

• use screws with larger diameters in the lower part

of the angle brackets in wall to floor connections

instead of using slender nails (due to the higher

withdrawal capacity of screws) so as to avoid

brittle failure for nails’ withdrawal;

• re-design the angle bracket using larger sections

(increased thicknesses) to avoid plasticization of

the metal part;

• provide better detailing to achieve higher moment

capacity so as to prevent local buckling and

excessive deformation of the metal connector

(for example, extend the stiffening plates of

hold-downs to reduce bending deflection due to

eccentricity);

• choose a number and diameter of holes for

fasteners to ensure brittle failure for fracture in

tension of the metal connection will always follow

the failure of the nailed connection between the

metal connector and the CLT panel;

• use slender nails or screws to ensure plasticization

of the fastener and achieve the ductile connection

between the steel connector and the CLT wall

panel;

• use longer nails for a better ductility performance

of CLT wall panels and decrease the risk of brittle

shear plug failure.

5 Conclusions

An experimental testing programme on cyclic behav-

iour of typical CLT metal connectors such as hold-

downs and angle brackets (wall-foundation and wall-

floor connections) was conducted with the aim to

better understand their performance when subjected to

seismic actions. Hold-downs exhibited relatively high

strength and stiffness capacity in their primary (ten-

sion) direction, while in shear direction due to

buckling of their metal part, they did not reach either

high strength or significant stiffness values. On the

other hand, angle brackets proved to have significant

strength and stiffness capacity in both directions.

Thus, their axial stiffness and strength should not be

neglected in the assessment of CLT walls resistance

due to overturning moments caused by horizontal

forces. In some cases undesired brittle failure modes

were detected. To attain ductile failure modes, capac-

ity based design principles should be taken into

account.

All important design quantities of connections such

as stiffness, strength, impairment of strength, ductility,

overstrength, and equivalent damping ratios were

calculated according to the EN 12512 procedure. This

information is important in the assessment of the

seismic performance of CLT buildings, as most of the

forces and energy dissipation due to seismic excitation

are concentrated in the connections between CLT

panels and with the foundation. Experimental results

also serve for calibration of advanced component FE

models for non-linear static and dynamic analyses of

CLT walls and buildings. In addition, overstrength

factors were determined, which are important missing

information in current seismic design codes (Eurocode

8). A conservative proposal would be to use an

overstrength factor of 1.3 for hold-downs and

1.25–1.45 for angle brackets, depending on the type

of the connection (panel-foundation or panel–panel).

A comparison of the experimental results with

analytical equations proved that the design models

provide conservative estimations of the characteristic

strength values of CLT connections. Comparisons


among analytical procedures showed that the Uibel

and Blaß’ method provides slightly more accurate

predictions than the existing embedding formulas in

Eurocode 5. Analytical models for CLT metal con-

nectors stiffness prediction, which take into account

contribution of nail deformation, steel plate elongation

and anchorage deformation, significantly overesti-

mates the experimental stiffness values. Suggestions

for a better mechanical performance of metal connec-

tors in terms of strength and stiffness performance are

also given based on the experimental and analytical

study.

Acknowledgments The authors would like to acknowledge

the contribution to this research of IVALSA laboratory staff

Mario Pinna, Diego Magnago and Paolo Dellantonio, who

provided technical expertise for the experimental testing.

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