inference of geotechnical property values

7/30/2019 Inference of Geotechnical Property Values

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INFERENCE OF

PROPERTY VALUES

FROM IN SITU TEST

DATA

5.1 INTRODUCTORY COMMENTS

A feature of geotechnical engineering over the last several decades has been agrowing interest in the uses of in situ testing as a site investigation technique.Such methods eliminate much the uncertainty associated with recovery of andtesting of "undisturbed" samples. An indication of the enthusiasm with which thisapproach has been embraced is the proliferation of ever more sophisticateddevices. This tendency is expected to continue. These notes are restricted to themore traditional and simpler tests. What is required is the ability to translate data

acquired from a given in situ test result to something else. On occasion it may bethe conversion of one in situ test value to the equivalent for another in situ test,for example conversion from CPT (Cone Penetration Test) to SPT (StandardPenetration Test) and vice versa. More often what is required is the conversionfrom an in situ test result to a numerical value for some soil property, for exampleconversion of the CPT value to the small strain shear modulus (although withadditional instrumentation the CPT can measure shear wave velocity from whichthe shear modulus can be obtained). The underlying concept is that the SPT andCPT are relatively simple tests to execute whereas the small strain shear modulusis comparatively expensive to determine in situ, hence the attractiveness ofobtaining the G value from the CPT via a correlation.


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Design of Earthquake Resistant Foundations

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A good reference for many of the headings in this chapter is Kulhawy and Mayne(1990) which gathers together information from widely dispersed sources.Recently a survey book, Schnaid (2009), has been published. A problem withdealing with data from various sources is the different systems of units (pressures

in kPa and MPa for SI units, psi (pounds per square inch) for sources from theUnited States and older Commonwealth practice, and kg/cm2 for oldercontinental publications). Kulhawy and Mayne deal with this by normalisation ofstresses so that their diagrams are independent of units. This is achieved bydividing quanities that are expressed in stress units by atmospheric pressure in thesame units. Consequently in Figures 5.22, 5.24, and 5.25 the ordinate is qc/pa

where qc is the cone tip resistance and pa is atmospheric pressure.

A word of warning: All the relationships between one property and anotherdiscussed in these notes have been derived empirically by comparing observationson the soil at a particular site or some soil with a similar geological history. The

point of caution is that it may not be appropriate to employ a correlation derivedfor one particular soil condition at some other site where the origin of the soil andgeological history is quite different. In the NZ environment soil formingprocesses have often been different from those in North America and Europe.

Weathering and volcanic action have been important here whereas glacial actionhas been very significant in North America and Europe. However, the 20thcentury development of soil mechanics took place in Europe and North Americaand so not in environments where residual and volcanic soils are dominant.Because of these differences the well-known correlations between liquid limit andcompression index for normally consolidated clays and between plasticity index

and su /'v may not be applicable to soils derived from in situ weathering

processes. Thus we need to be careful about importing correlations fromother countries.

This difference between the origin of our soils and the Northern Hemisphere soilsis reflected in NZS4402: "Testing of Soils for Civil Engineering Purposes", theprocedures laid down therein are significantly different from those in Britishstandard (BS1377) and the relevant ASTM standard.

5.2 SOME COMPLEX SOIL PROFILES

On this and the following pages several soil profiles are discussed. The first isfrom offshore Hong Kong which consists of a complex sequence of marinedeposits overlying the insitu bedrock. The remainder is from various NewZealand sites from Auckland, Tauranga and Canterbury. These profiles areincluded to make the point that natural soil deposits are nearly always complex.Even in situations where the soil type appears to be homogeneous there will be

variations in water content and cone resistance that are not apparent to the nakedeye. Design methods for foundations need to take account of this complexity in arealistic manner. Although some idealisation of the actual soil profile is alwaysnecessary to represent the main features in the model that is adopted for design.Clearly an important part of the foundation design process is to reach an

understanding of the types of soil present, the details of the layering, and thenature of the variability at the site in question.


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Chapter 5: Inference of property values from in situ test data

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5.2.1 Chek Lap Kok Island (near Hong Kong)

The profile in Figure 5.1 is offshore from Chek Lap Kok Island near Hong Kong,Koutsoftas et al (1987). In this case the profile consists of recent sediments with

relatively high water contents and consequent low undrained shear strengths.Apart from layers of upper and lower marine clays there is an alluvial crust layer inbetween. Below these there is more alluvium and weathered granite (labeled asdecomposed granite) followed by the underlying granite bedrock.

Figure 5.1Soil profile offshore Hong Kong (after Koutsoftas et al (1987)).

Figure 5.2Profiles of the upper marine clay at Chep Lak Kok Island


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Figure 5.3Profiles of the lower marine clay at Chep Lak Kok Island

Figures 5.2 and 5.3 present results from a thorough site investigation process. The

left hand part of the diagram plots the natural water content with depth along withbars indicating the water contents corresponding to the liquid and plastic limitvalues for the soils. In Figure 5.2 it is seen that the natural water content is greaterthan the liquid limit value, a sure sign that the soil will be soft with low shearstrength. This is confirmed with the undrained shear strength results in the righthad diagram of the figure from which it is apparent that at depths of 10 m theundrained shear strength is only about 20 kPa. The roughly linear increase inundrained shear strength with depth in the right hand diagram and the fact thatthe natural water content is higher than the liquid limit suggests that the soft soilsare normally consolidated or only lightly overconsolidated. This is confirmed inthe middle diagram where it is apparent that the maximum past vertical stress (the

so-called preconsolidation pressure) is typically less than twice the vertical effectivestress.

The deeper soil from the lower marine clay presents a different picture. Here thenatural water contents are about half way between the liquid and plastic limit andthe undrained shear strengths obtained from vane testing are rather larger than forthe upper marine clay. Undrained shear strengths greater than, say, 50 kParepresent material well capable of sustaining shallow foundation loads. Values inexcess of 100 kPa are representative of stiff clays. The middle diagram indicatesthat the inferred previous maximum vertical effective stresses tend to be severaltimes the in situ vertical effective stresses, again indicating soil with good shear

strength properties.


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Figure 5.4 Soil profile data obtained with a piezometric cone penetrometer

(piezo-CPT).

Diagrams such Figures 5.2 and 5.3 are a very useful means of supplementing theinformation in a borehole log and assist one to come to a useful understanding ofthe nature of the soil present.

Figure 5.4 presents another view of the nature of the soil profile at Chep Lak KokIsland obtained by profiling with a cone penetrometer equipped with a transducerfor measuring the pore water pressure adjacent to the cone tip. The CPT (conepenetration test) produces a continuous profile of soil data in this case the tipresistance, the sleeve friction (not plotted in the figure) and the pore waterpressure at the tip. With this information there are a number of interpretativetools for identifying soil type and values for geotechnical properties. Thecombination of tip resistance and tip pore water


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Figure 5.5Coefficient of consolidation values plotted against Liquid Limit

Figure 5.6 Soil compression coefficient values plotted against natural watercontent.


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pressure profiles in Figure 5.4 indicates the variability of the soil profile; even inthe various stratigraphic units identified in Figure 5.1 there is variability.

Figure 5.5 has the coefficients of consolidation plotted against the liquid limit

values for the specimens. These are representative of many correlations betweenthe liquid and plastic limit values and values for other geotechnical properties,liquid and plastic limit tests being standard classification tests for cohesive soils.

Figure 5.6 shows another attempted correlation; this time between the naturalwater content and compression coefficient. The rationale for this being thatgreater water content gives more compressible soil. Although a plausible rule ofthumb the data in Figure 5.6 shows that there is a very large amount of scatteramong the data points.

Thus both Figures 5.5 and 5.6 indicate that correlations between simple

classification variables for cohesive soils, natural water content and liquid andplastic limit values, may be helpful at the level of the broad-brush approach butthey are no substitute for proper investigation.

5.2.2 Auckland cone penetration profiles

An intense CPT investigation was done at Albany to obtain insight into thevariability of the Auckland residual soil profile, Holland and Pender (2008). Asshown in Figure 5.7 a total of 29 CPT probings were done at quite close spacing.

Figure 5.7Layout of CPT probings at Albany.

08

17 2120

30

29

28

27

22

23

24

01

02

16

07

06

05

2503

26

1918

15

14

13

12

11

10

09

04

8.0m

8.0m

1.0m

1.0m


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Figure 5.8 gives some insight into the nature of the variability; at each level themean of the 20 values is plotted along with the extremes. In the right hand sideof the figure the coefficient of variation (standard deviation / mean value) isplotted from which it is apparent that general the coefficient of variation of the

sleeve resistance is smaller than that for the cone resistance.

0 2 4 6 8

Penetration resistance (MPa)

0 5 10 1510

9

8

7

6

5

4

3

2

1

0

Friction ratio (%)

0 0.25 0.5 0.7510

9

8

7

6

5

4

3

2

1

0

Coeffs. of variation

Figure 5.8 CPT data averaged at each depth. Left: qc, middle: fs, right: coefficientsof variation of qc and fs.

fs

qc


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5.2.3 Comparison of Auckland soft soil profile and a volcanic profile forTauranga

Figure 5.9 shows CPT profiles obtained in the Auckland area (one of many along

the SH20 route) in a sedimentary profile and from a volcanic soil profile in theTauranaga region (Pender et al 1999). This shows that the Auckland profile ismore complex that the volcanic one because of layering of sands and siltymaterials. At both sites settlement data was available and so efforts were made tocorrelate the inferred ground stiffness with the CPT qc values. It is of interest thatthe volcanic material appeared to be much stiffer than the sedimentary soilalthough the volcanic site was loaded to a smaller fraction of the bearing strength.

ChCh

5.2.3 Mangere bridge site (near Auckland)

The profile at the Mangere bridge site consists of a complicated mixture of softmarine soils (referred to as muds), gravels and tuffs as well other volcanicallyderived layers underlain by tertiary age sandstone and siltstone, Figure 5.8. Thelateral load data and strain gauge data from the test pile made it possible to inferthe coefficient of subgrade reaction for the various layers. It is apparent that thetuff layer between 5 and 7 metres depth is very much stiffer than any other layer inthe top 17 metres of the profile. The nature of the materials in this profile wouldpreclude the CPT and even the SPT (standard penetration test) as routineinvestigation tools.

The complexities of layering at these two sites are by no means unusual. Beforeany foundation design can proceed an understanding of the site is essential. Toachieve this the input from engineering geologists and geotechnical engineerings is

vital. However, this is not a once only provision of data. Rather the design

Figure 5.9 Comparison of volcanic soil profile CPT (left) with that of asedimentary profile near Auckland (right) (after Pender et al (1999)

0 1 2 3 4 5 6 7 8 9 10

qc (MPa)

-20

-18

-16

-14

-12

-10

-8

-6

-4

-2

0

D

epth(m)

0 1 2 3 4 5 6 7 8 9 10

qc (MPa)

-20

-18

-16

-14

-12

-10

-8

-6

-4

-2

0

D

epth(m)


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0 10 2015

10

5

0

qc (MPa)

Depth(m)

Figure 5.10CPT profiles from Spencerville 2010

5.2.4 CPT profiles from sandy soil profiles in Canterbury.

Much CPT profiling has been done in Christchurch and the surrounding areassince the earthquake of September 04 2010 with additional investigation followingthe events of 2011. Figure 5.10 shows profiles from several of a number that weredone in the Spencerville area; the spacing between these was tens of metres in areathat had extensive lateral spreading but comparatively modest amounts of ejectedsand. It is clear that the upper 5 m of the profile is more susceptible toliquefaction than the material below.

5.2.5 Mangere bridge site (near Auckland)

The profile at the Mangere bridge site consists of a complicated mixture of softmarine soils (referred to as muds), gravels and tuffs as well other volcanicallyderived layers underlain by tertiary age sandstone and siltstone, Figure 5.11. Thelateral load data and strain gauge data from the test pile made it possible to inferthe coefficient of subgrade reaction for the various layers. It was apparent that thetuff layer between 5 and 7 metres depth is very much stiffer than any other layer inthe top 17 metres of the profile. The nature of the materials in this profile wouldpreclude the CPT and even the SPT (standard penetration test) as routine

investigation tools.


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Figure 5.11Geotechnical profile at the Mangere bridge site.


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Figure 5.12 Details of the field vane apparatus that can be used at the base of

a borehole.

5.3 VANE SHEAR STRENGTH PROFILES

The vane shear test is a simple device that is useful for measuring the undrained

shear strength at shallow depths in saturated clay deposits. In soft ground it can beattached to drill rods and pushed into the ground with measurements being takenat regular intervals. In stronger soils it is pushed into the bottom of a borehole andthe measurement taken and the borehole advanced after the vane is withdrawn.

The device for this application is shown in Figure 5.12. The soil needs to becohesive as it is presumed that the vane test is done under undrained conditions,consequently it gives a meaningful result only in saturated ground.

5.3.1 Liquidity Index to Soil State

The liquidity index is a way of expressing the in situ water content of a cohesivesoil in relation to the Atterberg limits.

PL

LL PL

w-wLI=

w -w 5.1

where: LI is the liquidity index value (often expressed as a percentage)w is the water content of the soilwLL is the water content at the liquid limitwPL is the water content at the plastic limit.

As shown in Figures 5.2 and 5.3 it is common to plot the Atterberg limits as a

horizontal line with barred ends and the natural water content shown as a boldsymbol, an example can be seen in Figures 5.2 and 5.3. This is a descriptive


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representation of the liquidity index. As explained above with regard to the ChepLak Kok Island soil profiles, if the natural water content is close to the liquid limit

value, a liquidity index about unity, then the soil will be "soft" having a lowundrained shear strength and is most likely to be normally consolidated. If the

natural water content lies at the other end of the water content range, near theplastic limit, the liquidity index is close to zero and the soil will be "stiff" having ahigh undrained shear strength and will behave as a heavily overconsolidated.

5.3.2 Plasticity Index to soil stiffness

A common empirical relation for cohesive soils is to assume that the undrainedYoungs modulus is several hundred times the undrained shear strength. Thisconnection between stiffness and shear strength connects the behaviour at thestart of loading to the behaviour at failure; about the only justification one canoffer for such a crude approach is the common observation for laboratory soil

testing that stiffer is stronger. At the stress levels (or factor of safety) associatedwith conventional settlement estimates it has been suggested that the appropriateYoung's modulus is about 500 times the undrained shear strength, but the datapresented in Figure 5.13 show that the plasticity index and overconsolidation ratioare also factors involved.

Figure 5.13Undrained Youngs modulus as a function of su and PI (after Duncanand Buchignani (1976)).


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Figure 5.14The SPT penetrometer

5.4 SPT IN SITU PENETROMETER TESTING AND INTERPRETATION

The dimensions of the SPT penetrometer are shown in Figure 5.14.

The SPT has a long history and originated in the USA. The SPT value is thenumber of blows of a 140 pound weight (0.625 kN) falling through a distance of

30 inches (762 mm) that are required to advance the penetrometer 1 foot(305mm). The main disadvantage of the test is that the result is that formerly totest result was very operator and equipment sensitive, and also it does not providea continuous record. On the advantage side there is a vast amount of data thathas been accumulated from SPT testing and it can penetrate harder material thanthe CPT.

Two modifications are commonly applied to SPT values: one for overburdenstress and the other to "normalise" the SPT value to a common energy input.

The first of these is intended to reduce the SPT values all to the same vertical

effective stress. The commomly used chart is given in Figure 5.15.


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The next correction is an attempt to allow for the fact that different drilling rigsand test setups will deliver various proportions of the theoretical energy. It hasbeen suggested that the 'standard' configuration delivers 60% of the availableenergy and so the actual energy ratio is adjusted for a given type of rig is adjusted

using equation 5.2 according to the values in Table 5.1.

Table 5.1 Energy ratios and procedures for SPT testing.

Summary of energy ratios for SPT procedures (Seed et al 1984)

Country Hammertype

Hammer release Estimated rodenergy (%)

Correctionfactor for 60%rod energy

Japan DonutDonut

Free-fallRope and pully withspecial throwrelease

7867

78/60 = 1.3067/60 = 1.12

USA SafetyDonut

Rope & pulleyRope & pulley

6045

60/60 = 1.0045/60 = 0.75

Argentina Donut Rope & pulley 45 45/60 = 0.75Chine Donut

DonutFree-fallRope & pulley

6050

60/60 = 1.0050/60 = 0.83

Recommended SPT procedures for liquefaction assessment (seed at al 1984)

Factor Recommended procedure

Borehole For to five inch (100 to 125 mm) rotaryborehole with bentonite drilling mud forborehole stability.

Drill bit Upward deflection of drilling mud(tricone or baffled drag bit)

Sampler OD = 2.0 in (50 mm)ID = 1.38 in (35 mm) (constant no roomfor liners in the barrel)

Drill rods AW for depths less than 50 ft (15 m), N,BW or NW for greater depths.

Energy delivered to sampler (rod

energy)

2520 in-lb (60% of theoretical maximum)

Blowcount rate 30 to 40 blows per minutePenetration resistance rate Measures over a range of 6 to 18 in of

penetration into the ground.


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Figure 5.15 Factors for correction of SPT values for overburden pressure (after

Kulhawy and Mayne (1990)).

The following these suggestions the corrected N value is given by:

m1 N m60

ER)( =N C N

605.2

where: CN is the correction for overburden pressure (Figure 5.15) and Nm is themeasured blow count and ERm the corresponding energy ratio in per cent. Table5.1 gives various details for the Standard Penetration Test suggested standardprocedures. The use of normalised SPT values given in equation 5.2 is importantin the assessment of liquefaction; further details about the normalizationprocedure and alternatives are given by Youd et al (2000).

5.4.1 Correlations between SPT resistance and friction angle

Figure 5.16 gives some correlations of long standing between SPT N value and

relative density and friction angle. Note that the correlation depends on the in situeffective stress. Reference to Figure 3.6 shows that friction angle for sanddetermined in laboratory testing is also affected by the effective stress.


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Figure 5.16 Factors for correction of SPT values for overburden pressure (after

Kulhawy and Mayne (1990)).

5.4.9 Relation between SPT resistance and coefficient of

subgrade reaction

One approach used to predict the lateral response of piles, particularly in layeredground, is to use the Winkler spring-bed model (cf Chapter 14). Relations, again

of long standing, have been suggested between the SPT N value and thecoefficient of subgrade reaction.

The origin of correlations between the coefficient of subgrade reaction and N isthe method, given by Terzaghi and Peck (1948), for estimating the settlement offootings on sands; this implies that:

ks = 0.33N (kPa/mm) 5.3

This value was increased in subsequent work by Peck and Bazzara (1970).

The size of the loaded area, the diameter or width of the pile shaft, has an effecton the coefficient of subgrade reaction. Terzaghi (1955) explains how increasingthe size of the pile will reduce ks, he justifies this by considering sizes of stressbulbs. Thus correlations between N and ks are usually given for a circular loadedarea of diameter 300mm.

Meyerhof (1965) gives:

ks = 0.75N (kPa/mm) 5.4

Scott (1981) suggests that the values given by Terzaghi (1955) be increased by50% for submerged sand to give:

ks = 1.8N (kPa/mm) 5.5


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Some data illustrating the effect of the diameter of the loaded area on thecoefficient of subgrade reaction is given by Terzaghi and Peck (1948) whoproposed that ks be adjusted for the size of the loaded area using:

ks (B) / ks (0.3m) = [2B/(B + 0.3)]2

5.6

where: B is the width of the loaded area in metres.

According to Sugimura (1986) the Road Bridge Standard Regulation in Japangives the following estimate for ks:

ks = 0.8ESD-3/4 (kgf/cm3) 5.7

Note that the 0.8 term in this equation carries units of cm-0.25, that D (pilediameter) is in cm, and that ES is in kgf/cm2. The term D-3/4 serves the same

purpose as the term [2B/(B + 0.3)] in the Terzaghi and Peck relation. Note thatthese two relations imply that the observed size effect is not simply a consequenceof elastic behaviour of the soil as ks does not decrease inversely with thedimension of the loaded area.

Another approach to estimating the stiffness of sands and gravels is implied in thesettlement procedure of Burland and Burbidge (1985), this is also based on SPTdata.

Yet another approach, based on SPT values, has been proposed for estimating thesmall strain shear modulus and shear wave velocity in sands by Seed et al (1986).

Note that this modulus refers to very small elastic strains, thus it will beconsiderably larger than the modului given above which refer to the strain levelsassociated with settlement of foundations at working loads. For example Seed et al(1986) propose the following relation for the small strain shear modulus ofnormally consolidated sands:

11/3)= 3.6 [( ] (MPa)G Nmax m 60 5.8

where: Gmax denotes the small strain shear modulus (the maximum value that it

may take for a given material and effective stress), m is the mean principaleffective stress (kPa) and (N1)60 is a corrected N value, refer to the Seed et al paperfor more details.

5.5 CPT IN SITU TESTING AND INTERPRETATION

The CPT originated in Holland, hence the frequent name: "Dutch Cone". It isalso sometimes called the Static Cone to distinguish it from the dynamic mode ofadvancement of the SPT. The original CPT measured both the point resistanceand the sleeve friction, Figure 5.17. More recently there has been muchdevelopment aimed at placing more instrumentation at the tip. The piezoconecan measure the pore pressure response. Other developments includeinstrumentation to record lateral pressures on the sleeve and a miniatureaccelerometer to allow shear wave velocity measurement. The chief advantages of


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the CPT are the continuous penetration record, so that very thin layers can bedetected, and much less operator sensitivity.

For sands the cone resistance is greater than for clays, but the unit sleeve friction

is comparatively lower for sands than for clays. These two observations are thebasis of a method of soil classification based on CPT results. An early version ofsuch a classification diagram is reproduced in Figure 5.16. (The friction ratioreferred to in the diagram is defined in equation 5.10.)

An alternative to Figure 5.16 was suggested by Robertson et al (1986) andupdated by Robertson (2010); it is shown in Figure 5.17. More recent work hasmoved towards normalisation of the variables used in the diagrams. Firstthough it is necessary to make a small correction for the effect the porepressure generated at the intersection of the sleeve and the cone as shown inFigure 5.17. The following terms are defined:

2(1 ) t cq q u a 5.9

where: a is the area ratio affected by the pore pressure u2 indicated in Figure5.17 (it is usually determined from laboratory calibration, a typical values being0.70 to 0.85).

The friction ratio is expressed in two ways:

100 100

s sr f

t vo t

f fF R

q q 5.10

Figure 5.17 CPT details. Left: Basic device. Right: correction of pore pressureeffects


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Figure 5.18 CPT soil type identification scheme (after Douglas and Olsen

(1981))

There are two versions of the normalised penetration resistance:

t votl

vo

qQ

5.11

and:n

t vo atn

a vo

q pQp

5.12

where: pa is atmospheric pressure and the value of n is discussed below. Thesecond of these parameters is used to calculate a soil behavior index:


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Zone Soil Behavior Type1 Sensitive, fine grained2 Organic soils - clay3 Clay silty clay to clay4 Silt mixtures clayey silt to silty clay5 Sand mixtures silty sand to sandy silt6 Sands clean sand to silty sand7 Gravelly sand to dense sand8 Very stiff sand to clayey sand*9 Very stiff fine grained*

* Heavily overconsolidated or cemented

Figure 5.19 CPT Soil Behavior Type (SBT) chart (Robertson et al., 1986,

updated by Robertson, 2010).

0.52 2

3.47 log( ) log( ) 1.22c tn rI Q F 5.13

0.381 0.05 0.15 1.0voca

n I np

5.14

Some iteration is required to achieve a satisfactory value for n, contours for avertical effective stress of one atmosphere are plotted in Figure 5.20.


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Figure 5.20 Contours of stress exponent, n, (for 'vo/pa = 1.0) on normalizedSBTn Qtn- Fr chart. (after Robertson (2009)).

Figure 5.21 Normalised CPT Soil Behaviour Type chart with contours of Ic (after

Robertson (1990) and Robertson (2010)).


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Table 5.2 Classification of soil type based on numerical values

for the soil type behaviour index, Ic

1 Sensitive, fine grained N/A2 Organic soils clay >3.63 Clays silty clay to clay 2.95 3.64 Silt mixtures clayey silt to silty clay 2.60 2.955 Sand mixtures silty sand to sandy silt 2.05 2.66 Sands clean sand to silty sand 1.31 2.057 Gravelly sand to dense sand


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Table 5.3 Suggested (qc/pa)/N60 ratios (Robertson (2010)

1 Sensitive fine grained 2.02 Organic soils clay 1.03 Clays: clay to silty clay 1.54 Silt mixtures: clayey silt & silty clay 2.05 Sand mixtures: silty sand to sandy silt 3.06 Sands: clean sands to silty sands 5.07 Dense sand to gravelly sand 6.08 Very stiff sand to clayey sand* 5.09 Very stiff fine-grained* 1.0

Table 5.3 gives the suggestions of Robertson (2010) for the CPT/SPT ratio forvarious soil types.

5.5 COMMON CPT EMPIRICAL RELATIONSHIPS

5.5.1 Cone resistance to undrained shear strength

As explained above, a frequently employed in situ means of measuring theundrained shear strength of a saturated clay at is the vane test, at least for shallowdepths. The CPT has the advantage of a continuous record of penetrationresistance over greater depths but does not give the shear strength, so the questionof conversion of cone resistance to su arises. Initially this conversion appeals to

bearing capacity theory for the = 0 case which gives the result that the net coneresistance will be 9 times the undrained shear strength.

Reality shows that this factor, usually denoted by Nk, depends on the plasticityindex of the clay. Clays with low PI have been found to have a ratio of about 20

whilst a plasticity index of 50 or more is required to obtain a value near thetheoretical value of 9. Figure 5.23 gives a collection of results from one source,Lunne et al (1976); there are a number of other publications that give similarresults. Although the data is scattered, the trend for Nkto decrease with plasticityindex is apparent. A typical starting value for Nk , in the absence of AtterbergLimit data, is to use a number about 15.


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Figure 5.23Variation of Nk with plasticity index (after Lunne et al (1976)).

5.5.2 Relation between CPT resistance and relative density

Several methods have been developed to convert cone resistance to relativedensity for sands. A selection of these given by Kulhawy and Mayne (1990) arepresented in Figure 5.24. Included in the figure is a diagram in which relativedensity is expressed in terms of the CPT resistance by means of a simple equationdeveloped by Lancellotta (1983).

5.5.3 Relation between CPT and friction angle

There have been a number of studies of cone resistance based on bearing capacitytheories. The concensus is that since the bearing capacity theories do not includethe effect of compressibility they are at best an approximation. Shown in Figure5.25 is a diagram with selections from the correlations given by Kulhawy andMayne (1990). They suggest that the scatter of data is at least partly a consequenceof differing overconsolidation ratios for the various sands.

Figure 5.26 gives the diagram from Robertson (2009) which has contours of peakfriction angle for sand deposits.


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Figure 5.24Various relative density - terms of penetration resistance relations.Top left: CPT data for CPT uncemented and unaged quartz sands. Top right:

effect of compressibility of the sand particles. Bottom: an alternative method

of Italian origin for estimating relative density from CPT values. (after Kulhawy

and Mayne (1990)).

Figure 5.25 Various correlations between penetration resistance and friction

angle of sands. Top: two correlations based on SPT data. Bottom: a correlation

for triaxial compression friction angle based on CPT data. (after Kulhawy and

Mayne (1990)).


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Figure 5.26 Contours of peak friction angle, , on normalized Q tn Fr chartfor uncemented Holocene age sandy soils. (after Robertson (2009)).

5.5.4 Relation between penetration resistance and small strainshear stiffness and shear wave velocity

The above comments relate the stiffness associated with estimating the settlementof a shallow foundation. Another important question is the small strain modulusof sand, the modulus associated with the passage of elastic waves through the soil.

This is very important with respect to earthquake site response and also because itgives the upper limit of stiffness of a given soil.

The CPT resistance has been correlated with shear wave velocity by Baldi et al(1989). They give the following equation:

0.13 0.27s c

q )= 277 ( (m/sec)voV 5.15

where: qc and vo are in MPa.

Robertson (2009) gives contours of a normalized shear wave velocity forpleistocene and holocene soils as shown in Figure 5.27. The normalized shear

wave velocity is defined by:0.25

1a

s s

vo

pV V

5.16


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Figure 5.27 Contours of normalized shear wave velocity Vs1 on normalized

SBTn Qtn Fr chart for uncemented, Holocene and Pleistocene age soils.

(after Robertson (2009)).

5.5.5 Relation between penetration resistance and stiffness

The conversion of CPT and SPT values to a modulus of some sort is a frequentoperation.

D'Appolonia et al (1968) observed the settlement of a large number of footingson sand at a steel mill in Illinois. They then backfigured the modulus for the sand,the results showed that overconsolidated sands are stiffer than normally

consolidated. In addition to overconsolidation aging is known to increase thestiffness of sand as shown in Figure 5.28.

Schmertmann (1970), in his method for estimating settlement from cone profilessuggested that the appropriate modulus is twice the cone resistance. TheNorwegian Geotechnical Institute has reviewed the various correlations based onCPT values, Lunne and Christofferson (1983). These results also show thatoverconsolidated sands are stiffer than normally consolidated sands. Theirrecommendations are given below in equation 5.17.


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Figure 5.28 Influence of overconsolidation and aging on the stiffness of sandsat 0.1% strain. (after Baldi (1989)).

Figure 5.29Correlation between constrained modulus and net cone resistance.

Influence (after Kulhawy and Mayne (1990)).


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:

4 10 ,

20 2 10 50

120 50:

5 50 ,

250 50

c c

c c

c

c c

c

for normally consolidated sands

E q for q MPa

q for q MPa

MPa for q MPafor overconsolidated sands

E q for q MPa

MPa for q MPa

5.17

Yet another aspect of the relation between soil stiffness and penetration resistanceis a correlation with the constrained soil modulus (that is the stiffness in one-dimensional stiffness). Figure 5.29 presents a collection of data gathered byKulhawy and Mayne. A well defined trend is apparent, but note that at small netcone resistance there is a considerable scatter in the correlation. Robertson (2009)offers an alternative to Figures 5.28 and 5.29. Figure 5.30 presents a relationbetween the normalized cone penetration data and a normalized Youngs modulus

expressed in terms of a modulus number KE or a modulus factor E. These aredefined by:

0.5

voE a E t vo

a

E K p E qp

5.18

Figure 5.30 Contours of Youngs modulus number, KE, and modulus factor,

E, on normalized SBTn Qtn Fr chart for uncemented, Holocene andPleistocene age soils. (after Robertson (2009)).


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5.6 APPLICATION OF THE MARCHETTI DILATOMETERPenetrometer testing has been the mainstay of geotechnical investigation for manydecades. A more recent device, that has gradually gained a place in the collection

of tools available, is the Marchetti dilatometer (DMT). The device consists of aspade shaped probe which is pushed into the ground. A diaphragm on the side isused to measure the lateral stress acting against the spade. The concept of thedevice is shown in Figure 5.31 and a photograph of the spade is in Figure 5.32 thedimensions of which are about 95 mm wide and 15 mm thick. In operation of thedevice is pushed to the required depth and the membrane is inflated from whichthe liftoff pressure (po) and the pressure to displace the diaphragm 1.1 mm (p1) arenoted (not explained here are some calibration steps needed in estimating thesetwo pressures). From these measurements values for various soil properties can beinferred.

In another variation the device can be equipped to measure the shear wavevelocity of the soil.

Figure 5.31The concept of the Marchetti dilatometer.


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Figure 5.32Photograph of the flat dilatometer spade.

Three parameters are calculated from the results obtained. These are the MaterialIndex ID, defined as:

1 oD

o o

p pI p u

5.19

where: uo is the water pressure at the position where the measurement is taken(before the spade is inserted).

The material index is thought to provide a reasonable estimate of soil typefollowing Marchetti (1980) who observed that the difference between po and p1 issmall for clay and large for sand.

The Horizontal Stress Index KD, is defined as:

o oD

vo

p uK

5.20

where: vo is the in situ vertical effective stress at the position in the soil profilewhere the measurement is made. KD is related to Ko but is greater than Kobecause of the effect of blade penetration.

The Dilatometer Modulus ED, is defined as

135D oE p p 5.21


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This is obtained by assuming that the soil either side of the spade is an elastic halfspace and that at the lateral displacement of 1.1 mm the soil can still be idealizedas elastic.

A profile of DMT data for a clay soil profile is shown in Figure 5.33 and for asand profile in Figure 5.34.

Figure 5.33DMT results for a clay soil profile.

Figure 5.34DMT results for a sand soil profile.


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Figure 5.35DMT results and soil type c lassification.

5.6.1 DMT correlations with soil property values

As with the SPT and CPT there are a whole suite of correlations between DMTmeasurement values and soil property values. Figure 5.35 has a plot of dilatometermodulus, ED, against material index, ID. This diagram shows how various soiltypes fall into well-defined zones, the dilatometer equivalent of Figures 5.18 and5.19.


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Figure 5.36 Comparison between Ko values measured in a soft soil profile using

the DMT and self-boring pressuremeter (see section 5.7) (a) from a site in

Skotland and (b) from a site in Italy.

Figure 5.36 compares Ko values for a soft soil profile obtained from the DMTtesting and the use of the self-boring pressuremeter (discussed in section 5.7). Anobvious objection to the DMT estimation of the Ko is that there must beconsiderable disturbance involved in pushing the spade into the ground so that po,must be greater than the in situ horizontal effective stress. The self-boringpressuremeter, on the other hand, was specifically developed to introduce minimal

disturbance into the soil profile. Thus self-boring pressure values of Ko areregarded as being true. The comparison in Figure 5.36 shows that, even so, thatKo values from the DMT are in good accord with those from the self-boringdevice.

Figure 5.37 shows data giving the undrained shear strength, normalized withrespect to the vertical effective stress, in relation to the values obtained for thehorizontal stress index. The regression line shown on the diagram has thefollowing equation:

125

022 2

.

. Du voK

s

5.22


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Figure 5.37 Relation between undrained shear strength and the DMT horizontal

stress index.

Figure 5.38 Comparison between constrained modulus values estimated fromDMT data and those obtained from high-quality oedometer test data.

Finally in this section Figure 5.38 compares constrained modulus values estimatedfrom DMT testing with results from high quality oedometer testing.


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Thus we can conclude that, even though the DMT test might not be expected togive values for undisturbed soil properties the above diagrams show thecorrelations that have been developed seem to be able to account for theseproblems. However, one needs to note that Figures 5.36 and 5.38 give data only

for soft cohesive soils.

5.7 PRESSUREMETER TESTING

Another in situ testing device is the pressuremeter. This is a way of applying apressure to the side of a borehole and through measurement of the radialdisplacement of the side of the borehole to infer values for soil properties. Thereare two versions of the device. The first illustrated in Figure 5.39 is the Menardpressuremeter which is used extensively in France for the design of foundations.

An obvious criticism of the Menard pressuremeter is that the device needs to beinserted into a pre-drilled borehole. The problem is that by the time the device is

inserted into the borehole there will be complete release of the in situ stresses inthe ground, which is a serious form of soil disturbance. Nevertheless, by means ofappropriate correlations it appears that the French geotechnical community hasbeen able to make good use of the device another illustration of the power ofcareful correlation work.

Pressuremeter of similar design to the Menard device can be used effectively inboreholes in rock in which case the disturbance because of the release of in situstress are not so serious but it is necessary to have a more sensitive way ofmeasuring radial deformation as the stiffness of the rock is so much greater thanthat of soil.

Figure 5.39 Concept of the Menard pressuremeter.


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Figure 5.40Concept of the self-boring pressuremeter.

A second type of pressuremeter, especially developed for testing soil profiles is theself-boring device, Wroth and Hughes (1973), which is illustrated in Figure 5.40.

The idea behind the development of this device is that it can be drilled into softsoil layers with minimal disturbance so that the in situ horizontal effective stress aswell as the soil stiffness can be measured accurately. Attempts have been made touse this device around Auckland but there have been problems with shellfragments in sandy soils rupturing the thin rubber membrane.

5.8 GEOPHYSICAL METHODS OF SOIL PROFILING

An aim of site investigation work is to define the layering present in the soilpresent at a site. The CPT is clearly an effective way of achieving this. Figures 5.1

to 5.11 illustrate just how variable are natural soil deposits and so the questionarises about how to fill in the details between boreholes and CPT probings. Atsites where the point to point variability is considerable there is a need for amethod that is capable of giving a good indication of average properties for the

various layers. Geophysical prospecting provides just such a tool. There are twotechniques that are of use in site investigation: SASW (spectral analysis of surface

waves) and MASW (multichannel analysis of surface waves).

In the past the geotechnical community has been disappointed with geophysicalmethods, the promised performance has not often been realized. However, inrecent years there has been better instrumentation available and also much more

capable software for processing results of field testing. So it seems that thesemethods need to be considered anew.


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Figure 5.41 Shear strain ranges associated with various foundation engineeringapplications.

Basically what is being measured is the shear wave velocity of the soil. Figure 5.41complements Figures 3.24 and 4.41 and shows how various foundationengineering applications are associated with certain ranges of shear strain. All thedesign applications are well beyond the strains associated with elastic shear waves.

The diagram, as do Figures 3.24 and 4.41, show clearly that the shear modulus

associated with the passage of shear waves gives an upper bound on the stiffnessof the soil. Even so, the application to the very small strains involved ingeophysical investigation work is not a difficulty as long as the method is capableof mapping the shear wave velocity distribution in the soil profile. The underlyingassumption of the use of these methods in site investigation work is that changesin the shear wave velocity are indicative of changes in other soil properties.

5.8.1 Spectral analysis of surface waves (SASW)

The concept of this method is illustrated in Figure 5.42 and explained in moredetail by Stokoe et al (2004). As can be seen a pair of geophones are placed on the

ground surface and at some point is a source of vibration which is intended togenerate surface waves. The vibration source may be either steady state orimpulsive (ie hammer blows). For the steady state excitation the sensing devices(usually geophones) remain at fixed positions and the excitation frequency isgradually changed. For the impulsive excitation multiple blows are required withthe spacing of the sensors changed. From the information recorded it is possibleto estimate the distribution of shear wave velocity with depth at the site. Figure5.43 shows some results for work done in Canterbury in 2010.


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Figure 5.42 SASW field set-up and data processing to estimate soil shear wavevelocity with depth (after Stokoe et al (2004)).

10

8

6

4

2

0

Depth,m

300250200150100500

Shear Wave Velocity, m/s

VSGWL

Figure 5.43SASW results obtained in Canterbury in 2010.


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5.8.2 Multichannel analysis of surface waves (MASW)

A further development of geophysical methods is the so-called MultichannelAnalysis of Surface Waves (MASW); the concept is shown in Figure 5.44. The key

difference with SASW is that there are more than two receivers, so, at the expenseof even more sophisticated numerical analysis, long sections with shear wavecontours are plotted. An example from recent work in Christchuch is reproducedin Figure 5.45. It needs to be realized that what is plotted in Figure 5.45 is amoving average of the shear wave velocity, not spot values.

Figure 5.44 MASW field set-up and illustrative processed results.

Figure 5.45An MASW section from Christchurch.


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References

Baldi, G., Bellotti, R., Ghionna, V.N., Jamiolkowski, M., and Lo Presti, D.F.C.,1989. Modulus of sands from CPTs and DMTs. InProceedings of the12th International Conference on Soil Mechanics and FoundationEngineering. Rio de Janeiro. Balkema Pub., Rotterdam, Vol.1, pp. 165-170.

Been, K., Crooks, D. E., Becker, D. E. and Jefferies, M. G. (1985) "A stateparameter for sands", Geotechnique.

Bjerrum, L. (1973) "Problems of soil mechanics and construction on soft clays",State-of-the-Art Report 8th ICSMFE, Moscow, vol. 3, pp. 111-159.

Burland, J. B. and Burbidge, M. C. (1985) "Settlement of foundations on sand andgravel", Proc. ICE, Part 1, pp. 1325-1381.

D'Appolonia, D. J., D'Appolonia, E. and Brissette, R. F. (1968) "Settlement ofspread footings on sand", Proc. ASCE, Jnl. Soil Mechanics Div., Vol. 94,SM3, pp. 735-760.

Douglas, B.J., and Olsen, R.S., 1981. Soil classification using electric conepenetrometer. In Proceedings of Symposium on Cone Penetration

Testing and Experience, Geotechnical Engineering Division, ASCE. St.Louis, Missouri, October 1981, pp. 209-227.

Duncan, J. M. and Buchignani, A. L. (1976) "An engineering manual forsettlement studies", Geotech. Eng. Report, Dept. Civil Eng., Univ. Calif.Berkeley.

Holland, A. and Pender, M. J. (2008) Variability of an Auckland residual soil

profile obtained from closely spaced CPT soundings, Proc. of the NZGeotechnical Society Geotechnical Symposium 2008, July, pp. 63-68.

Koutsoftas, D. C., Foott, R. and Handfelt, L. D. (1987) "Geotechnicalinvestigations offshore Hong Kong", Proc. ASCE, Jnl. Geotech. Eng.,

Vol. 113 No. 2, pp. 87-105.Kulhawy, F. H. and Mayne, P. W. (1990) Manual on estimating soil proiperties

for foundation design, EPRI EL-6800 Project 1493-6 Final Report.Lancellotta, R. (1983), cited by Jamiolkowski, M., Ladd, C. C., Germaine, J. T. and

Lancellotta, R. (1985) "New developments in field and laboratory testingof soils", Proc. 11th. ICSMFE, San Francisco, Vol. 1, pp. 57-153.

Lunne, T and Christoffersen, H. P. (1983) "Interpretation of Cone Penetrometer

data for Offshore Sands", Proc. 15th. OTC Conference, Houston, Vol. 1,pp. 181-192.

Lunne, T. H., Eide, O. and De Ruiter, J. (1976) "Correlations between coneresistance and vane shear strength in some Scandanavian soft to mediumstiff clays", Canadian Geotechnical Journal, Vol. 13, pp. 430-441.

Lunne, T., Robertson, P.K., and Powell, J.J.M., 1997. Cone penetrationtesting in geotechnical practice. Blackie Academic, EFSpon/Routledge Publ., New York.

Maarchetti, S. (1997) The flat dilatometer: design applications. Keynotelecture. Proc 3rd International Geotechnical Engineering Conference,Cairo. Pages 421-448.


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Park, Choon B, Richard D. Miller, Jianghai Xia, and Julian IvanovMultichannel analysis of surface waves (MASW); active and passivemethods, Leading Edge (Tulsa, OK) (January 2007), 26(1):60-64.

Park, C. B., and Miller, R.D., 2008, Roadside passive multichannel analysis of

surface waves (MASW): Journal of Environmental & EngineeringGeophysics, v. 13, no. 1, p. 1-11.

Pender, M. J., Jennings, D. N. & Crawford, S. A. (1999) Relation betweensettlement and cone penetration resistance at two sites Proc. 5thInternational Symposium on Field Measurements in Geomechanics FMGM99, Singapore, Balkema, pp 601-608.

Robertson, P. K. (2009) Interpretation of cone penetration tests a unifiedapproach. Canadian Geotechnical Journal.

Robertson, P. K. and Cabal (Robertson), K. L. (2010) Guide to cone penetrationtesting for geotechnical engineering Gregg Drilling and Testing, 4thedition.

Schmertmann, J. H. (1970) "Static cone to compute static settlement over sand",Proc. ASCE, Jnl. Soil Mechanics Div., Vol. 96 SM3, pp. 1011-1043.

Schnaid, F. (2009) In situ testing in geomechanics: the main tests. Taylor &Francis.

Seed, H. B., Tokimatsu, K., Harder, L. F. and Chung, R. M. (1984) "The influenceof SPT procedures in soil liquefaction resistance evaluations", Report No.UCB/EERC-84/15, Earthquake Engineering Research Center, Universityof California, Berkeley.

Seed, H. B., Wong R. T., Idriss, I. M. and Tokimatsu, K. (1986) "Moduli andDamping factors for Dynamic analyses of Cohesionless Soils", Proc.

ASCE, Jnl. Geotech. Eng. Div., Vol. 112, No. 11, 1016-1032.

Stokoe, K. H., Wright, S. G., Bay, J. A. & Roesset, J. M. (1994). Characterization ofgeotechical sites by SASW method. Geophysical characterization of sites,pp. 15-25.

Stokoe, K. H., Sung-Ho, Joh. And Woods, R. D. (2004). Some contributions to in situgeophysical measurements to solving geotechnical engineering problems.Proc. International Conference on Site Characterisation (ICS-2), PortoPortugal, September.

Wroth, C.P., and Hughes, J.M.O. (1973). An instrument for the in situ measurementof the properties of soft clays. In Proceeding of the 8th InternationalConference on Soil Mechanics and Foundation Engineering, Moscow, Vol. 1,pp. 487494


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